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Graphical representation of crack widths

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Volume 45 Number 4 December 2015  i


ii  Volume 45

Number 4 December 2015

The Bridge and Structural Engineer


The Bridge & Structural Engineer

Indian National Group of the International Association for Bridge and Structural Engineering ING - IABSE

Contents :

Volume 45, Number 4 : December 2015

Editorial • From the Desk of Chairman, Editorial Board : Mr. Alok Bhowmick vi • From the desk of Guest Editor : Mr. Ravi Sundaram viii

Highlights of ING-IABSE Event

• Highlights of the ING-IABSE Workshop on “Code of Practice for Concrete Road Bridges IRC:112:2011” held at Hyderabad on 30th & 31st October 2015

xi

Special Topic : Geotechniques & Foundation Design of Structures

Indian Innovations in Foundation Engineering 1 Dr. N.V. Nayak Pile Load Testing in India - Current Practices and Recent Developments 7 Sunil S. Basarkar, Ravikiran Vaidya Recent Advancements in Ground Engineering Techniques in India: An Overview 18 Dr. V.R. Raju, Madan Kumar Annam Evaluation of Liquefaction Potential for Design of Deep Foundation 31 Mahesh Tandon, Navneet Gupta, Harsimran Singh Grewal Liquefaction Potential of Delhi and Some Mitigation Options 39 Chandan Ghosh Prefabricated Vertical Drains – Recent Developments 54 Dr. G. Venkatappa Rao, Dr. M.V.S. Sreedhar A Case Study on Assessment of Liquefaction Potential for Basal Reinforced Embankment on Soft Soil 64 Minimol Korulla, Anusha Nandavaram, Saurabh Chaurasia, Meenu P.S. 8. Geocell Basal Mattress for High Embankments and Retaining Walls 72 Dr. G. Venkatappa Rao, S. Jaswant Kumar 9. Improvement of soft clay Foundation bed for Embankments using Geocell - An Experimental Study 79 Sefali Biswas, R. B. Sahu, Satyendra Mittal, G. Bhandari 10. Geotechnical Investigations in Gravel-boulder deposits 87 Ravi Sundaram, Sanjay Gupta 11. Rectification Scheme of Collapsed Sheet Pile adjacent to Existing Railway Line –A Case Study 97 Dr. D. N. Naresh, Jitendra Kumar, Mohit Jhalani 12. Engineering Aspects of Storage Caverns 102 Altaf Usmani, Chandan Kumar, Atul Nanda 13. Deformability of Rock Mass for Dam Foundation 110 Dr. Rajbal Singh 14. Disaster Mitigation and Role of Civil Engineers 118 Dr. R. Kuberan

Contents

1. 2. 3. 4. 5. 6. 7.

Research Paper

1. An Experimental study on the Behaviour of Steel Plate-Anchor Assembly Embedded in Concrete under Cyclic Loading 128 Deepak K. Sahu, Saiwal Krishna, S.K. Chakrabarti

Panorama • Office Bearers and Managing Committee – 2015

145

• List of ING-IABSE Publications

148

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Volume 45 Number 4 December 2015  iii


The Bridge & Structural Engineer JOURNAL OF THE INDIAN NATIONAL GROUP OF THE INTERNATIONAL ASSOCIATION OF BRIDGE & STRUCTURAL ENGINEERING

March 2016 Issue of the Journal will be a Special Issue with focus on ENABLING WORKS, FORMWORKS & SACAFFOLDING SYSTEMS–Principles of Design and Construction SALIENT TOPICS TO BE COVERED ARE : 1. Modern Formwork System 2. Enabling & Temporary Works 3. Lifting, Transportation, Handling & Erection 4. Scaffolding Systems 5. Design & Codal Provisions 6. Safety & Precautions 7. Formwork Failures & Case Studies

The Bridge & Structural Engineer JOURNAL OF THE INDIAN NATIONAL GROUP OF THE INTERNATIONAL ASSOCIATION OF BRIDGE & STRUCTURAL ENGINEERING

June 2016 Issue will be a Special Issue with focus on TALL STRUCTURES (Tall Buildings, Chimneys, Silos, TV Towers, Cooling Towers, Transmission Towers)

SALIENT TOPICS TO BE COVERED ARE : 1. 2. 3. 4. 5. 6.

Structural System & Forms Green Building & Smart Cities Wind induced response & EQ resistant design of tall structures Critical Appraisal of Existing Codes & Standards (Indian as well as International) New Construction Materials and Techniques. Case studies for Design, Construction and Rehabilitation

Those interested to contribute Technical Papers on above themes shall submit the abstract by 30th April 2016 and full paper latest by 30th May 2016 in a prescribed format, at email id : ingiabse@bol.net.in

iv  Volume 45

Number 4 December 2015

The Bridge and Structural Engineer


B&SE: The Bridge and Structural Engineer, is a quarterly journal published by ING-IABSE. It is one of the oldest and the foremost structural engineering Journal of its kind and repute in India. It was founded way back in 1957 and since then the journal is relentlessly disseminating latest technological progress in the spheres of structural engineering and bridging the gap between professionals and academics. Articles in this journal are written by practicing engineers as well as academia from around the world.

Disclaimer :

Editorial Board

All material published in this B&SE journal undergoes peer review to ensure fair balance, objectivity, independence and relevance. The Contents of this journal are however contributions of individual authors and reflect their independent opinions. Neither the members of the editorial board, nor its publishers will be liable for any direct, indirect, consequential, special, exemplary, or other damages arising from any misrepresentation in the papers.

Chair :

The advertisers & the advertisement in this Journal have no influence on editorial content or presentation. The posting of particular advertisement in this journal does not imply endorsement of the product or the company selling them by ING-IABSE, the B&SE Journal or its Editors.

Jose Kurian, Former Chief Engineer, DTTDC Ltd., New Delhi

Alok Bhowmick, Managing Director, B&S Engineering Consultants Pvt. Ltd., Noida

Members : Mahesh Tandon, Managing Director, Tandon Consultants Pvt. Ltd., New Delhi A K Banerjee, Former Member (Tech) NHAI, New Delhi Harshavardhan Subbarao, Chairman & MD, Construma Consultancy Pvt. Ltd., Mumbai Nirmalya Bandyopadhyay, Director, STUP Consultants Pvt. Ltd., New Delhi S C Mehrotra, Chief Executive, Mehro Consultants, New Delhi

Advisors : A D Narain, Former DG (RD) & Additional Secretary to the GOI N K Sinha, Former DG (RD) & Special Secretary to the GOI G Sharan, Former DG (RD) & Special Secretary to the GOI A V Sinha, Former DG (RD) & Special Secretary to the GOI S K Puri, Former DG (RD) & Special Secretary to the GOI R P Indoria, Former DG (RD) & Special Secretary to the GOI

Front Cover :

Top Left : Triaxial test in soil sample in progress. The pneumatic panel measures the pore-water pressure and volume change in the sample. Bottom Left : Pile load test for the Commonwealth Games Project in progress. Akshardham complex in the background. Right : Field SPT in progress using automatic trip hammer.

• Price: ` 500

S S Chakraborty, Former Chairman, CES (I ) Pvt. Ltd., New Delhi B C Roy, Former Senior Executive Director, JACOBS-CES, Gurgaon Published : Quarterly : March, June, September and December Publisher : ING-IABSE C/o Secretary, Indian National Group of the IABSE IDA Building, Ground Floor (Room Nos. 11 and 12) Jamnagar House, Shahjahan Road New Delhi-110011, India Tel: 91+011+23388132, 91+011+23386724 E-mail: ingiabse@bol.net.in, ingiabse@hotmail.com, secy.ingiabse@bol.net.in Submission of Papers : All editorial communications should be addressed to Chairman, Editorial Board of Indian National Group of the IABSE, IDA Building, Ground Floor, Jamnagar House, Shahjahan Road, New Delhi – 110011. Advertising: All enquiries and correspondence in connection with advertising and the Equipments/Materials and Industry News Sections, should be addressed to Shri RK Pandey, Secretary, Indian National Group of the IABSE, IDA Building, Ground Floor, Jamnagar House, Shahjahan Road, New Delhi-110011.

The Bridge and Structural Engineer

Journal of the Indian National Group of the International Association for Bridge & Structural Engineering

December 2015

The Bridge & Structural Engineer, December 2015

The Bridge and Structural Engineer

Volume 45 Number 4 December 2015  v


From the Desk of Chairman, Editorial Board

To maintain the high quality of this esteemed journal, there is a dire need to maintain a strong flow of comprehensive papers on seminal theoretical as well as practical topics of interest, case studies highlighting important innovations in the field, professional issues that a Civil Engineer is beleaguered with and technical papers pertaining to critical review and codes and standards. Preparing such high quality papers generally requires significant time commitments from senior level academics and practicing engineers. I find that there is a reduced willingness of industry now a days to invest time and energy in preparing their most significant projects for journal publications. We in the editorial board do recognize the fact that there is a time constraints due to a quantum increase in number of infrastructure projects in the country, which is keeping all the competent engineers busy. However it is to be realised by all that ‘publication is vital to a profession’ and I would urge for a greater commitment to contribute papers, from senior structural and geotechnical engineers of the industry, in the larger interest of this civil engineering fraternity. The journal “The Bridge & Structural Engineer” is one amongst very few committed and collaborative linkage between academia and

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industry and would continue to make efforts in bridging this gap. This issue of the journal is focussed on the theme of “GEOTECHNIQUES & FOUNDATION DESIGN OF STRUCTURES". Interaction always takes place between a structure and its foundation, whether one likes it or not. If structure-ground interaction is to be taken into account in the design properly, structural and geotechnical engineers have to themselves interact closely. In my professional career, I have often encountered profound differences in approach between structural and geotechnical engineers, leading to a lack of understanding of the structural behaviour by the geotechnical engineer and lack of understanding of the soilstructure interaction by the structural engineer. These differences in approach are not only restricted to academics, but also between practicing structural and geotechnical engineers. The Editorial Board thought prudent to take steps in improving the understanding of soilstructure behaviour between structural as well as geotechnical engineers. The Editorial Board of ING-IABSE therefore decided to dedicate this issue on the topic of “GEOTECHNIQUES & FOUNDATION DESIGN OF STRUCTURES".

The Bridge and Structural Engineer


Our Guest Editor for this issue is Mr Ravi Sundaram, who is a well-known personality in the field of Geotechnical Engineering. He has been involved in developing engineering solutions to problems of collapsible soils, soft clays, expansive soils, weak rocks, etc. on major projects in India and abroad. He has published several technical papers that reflect his experience in geotechnical engineering practice. Mr Ravi Sundaram has taken considerable pains in carefully selecting the sub-theme topics and

The Bridge and Structural Engineer

the expert authors suitable for the sub-theme, so as to give broad coverage to all aspects of geotechnical and foundation engineering. It is hoped that the readers will find this issue of interest and value. Happy Reading

(ALOK BHOWMICK)

Volume 45 Number 4 December 2015  vii


From the Desk of Guest Editor

I feel extremely privileged and honoured to serve as Guest Editor of this issue of “The Bridge and Structural Engineer”. This special issue on Geotechniques and Foundation Design of Structures covers not only bridges but a whole range of civil engineering structures including metro projects, underground storage caverns, dam foundations, etc. The papers received from eminent experts cover a wide variety of topics of current interest such as liquefaction, ground improvement, pile foundations, sheet piles, soft soils, gravel-boulder deposits, deformability of rocks and disaster mitigation. Geo-cells, a new development in geotechnical engineering is exciting researchers and field engineers alike. The importance of designing foundations in liquefiable sands has generated two very interesting papers. State-of-the-practice papers on Indian innovations, ground improvement and pile load tests by the industry leaders have set the tone for geotechnical engineering excellence in the Indian construction industry. With the rapid pace of development in the infrastructure sector, fast-track projects have become the norm and engineers are expected to meet tough deadlines. This puts a lot of pressure on engineers, not just to complete the work on time but also to ensure that all

viii  Volume 45

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technical issues are effectively resolved. Also, quality assurance and reliability of design and construction have to be ensured. Geotechnical engineers have a vital role to play in solving some of the world’s most pressing problems of space utilization, transportation sector, construction in areas with difficult soil conditions and developments in the Himalayan region. In doing so, they have to expand their practice well beyond traditional soil mechanics and earth sciences into newer areas and develop an understanding of related allied subjects. It is becoming increasingly important to relate engineering solutions not only based upon overall stability, but also have acceptance criterion based upon its anticipated performance with the developments in the infrastructure sector, construction is being taken up in areas / lands which were considered as “not buildable” in the past. Foundations being the most important element of the structure, geotechnical aspects play a very vital role in the design and construction of the substructure. Any failure could jeopardize the project, resulting in time and cost over-runs in addition to the all too known blame-game and litigation. It is often said that “Geotechnical Engineering is a science but its practice is an art”. Soil being a natural deposition which inherently

The Bridge and Structural Engineer


exhibits variations, the art of interpreting the strata conditions and assessing foundation behaviour has to be tempered by knowledge of local conditions, geology of the area as well as performance of other structures in the vicinity. This emphasises the need for “observational method of design” in which the design may have to be modified based on the site conditions encountered and field observations. Expecting the “unexpected to occur”, the geotechnical engineer has to be prepared to change the analysis / design as per the site conditions exposed at the construction stage. The challenge in such design concept cannot be over-emphasized and involves proactive participation of all stakeholders for a successful execution of the project. This calls for a change in the mindset of structural engineers, designers, architects, clients / owners, contractors, etc. who are used to receiving geotechnical inputs / investigation reports in the early stages of the project. The design is usually done on the basis of the soil reports and there is no further reference to the geotechnical engineer unless there is an unusual ground condition or

The Bridge and Structural Engineer

failure or a “geotechnical surprise”. This can be well avoided if the contract calls for including the services of the geotechnical engineer till the foundation construction is over. I am extremely thankful to all the authors who immediately responded to my request to contribute papers and have very kindly taken time out of their busy schedule to write such excellent papers. These invaluable “gems” have made this issue of the journal a collector’s item, worth preserving and referring to whenever the need arises. I do hope that practicing engineers will find the papers in this issue useful and will find answers to questions they were hesitant to ask. I thank Mr. Alok Bhowmick and the team of the ING-IABSE for giving me this opportunity to edit this special issue and connect to all members to make this effort a success. Their cooperation and guidance in this effort is gratefully acknowledged.

(RAVI SUNDARAM)

Volume 45 Number 4 December 2015  ix


Brief Profile of Mr Ravi Sundaram Mr. Ravi Sundaram completed his M.Tech. in 1980 from IIT Delhi and has 35 years experience as a consulting engineer in India and abroad. He started his career with McClelland International Limited (USA), a worldwide reputed geotechnical company. He was based in the company’s Office at Dammam, Saudi Arabia and worked on the company’s projects in Saudi Arabia and Qatar. His international experience includes offshore & onshore soil investigations, geophysical tests, pile installation monitoring and pile drivability studies. He is a founder director of Cengrs Geotechnica Pvt. Ltd. and directs all geotechnical consultancy services and geotechnical construction activities. He is in charge of soil investigation works including field works, laboratory testing and engineering studies. His field of activities also includes pile foundations, foundation-strengthening and ground improvement. Mr. Ravi Sundaram’s expertise includes boreholes through soil and rock, resistivity and seismic surveys, dynamic testing of soils, rammed stone columns, bored piles and groundwater studies. He has been involved in developing engineering solutions to problems of collapsible soils, soft clays, expansive soils, weak rocks, artesian conditions and liquefaction of soils on major projects in India and abroad. His special interests include installation and testing of piles, foundations for bridges, high rise buildings, large-diameter tanks, power plants and heavy industrial structures. He has published over 40 technical papers that reflect his experience in geotechnical engineering practice. He is a fellow of the Indian Geotechnical Society and member of the International Society of Soil Mechanics & Geotechnical Engineering. He is also member of the Indian Roads Congress and the Association of Consulting Civil Engineers. Mr. Ravi Sundaram was Chairman of the Indian Geotechnical Society Delhi Chapter during 2009-11 and was responsible for getting it registered with the Registrar of Societies. He is actively associated with Bhoovigyan Vikas Foundation, a non-government organization working towards Earth-Care and advising the Indian Government on making their policies Earth-friendly. He has published a book titled “Retrieving the Lost Paradise – Celebrating the Earth Day”.

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HIGHLIGHTS OF THE ING-IABSE WORKSHOP ON “CODE OF PRACTICE FOR CONCRETE ROAD BRIDGES IRC:112:2011” HELD AT HYDERABAD ON 30TH & 31ST OCTOBER 2015 The Indian National Group of the IABSE in co-operation with Govt of Telangana, Roads and Buildings Department (NH Wing) successfully organised two day Workshop on “Code of Practice for Concrete Road Bridges IRC:112:2011” at Hyderabad on 30th and 31st October 2015. The Workshop was well attended by more than 200 delegates from various Govt Departments as well as other private and public organizations. The aim of the workshop was to provide a unique opportunity to the practicing engineers and students to interact with experts for dissemination of knowledge and experiences relating to the latest techniques in design of bridges and other structures using the “Code of Practice for Concrete Road Bridges IRC:112: 2011”. Participation of delegates in floor intervention and discussions was very encouraging. The Workshop was inaugurated by Shri T Nageshwar Rao, Hon’ble Minister of Public Works Department, Govt of Telangana by lighting the traditional lamp in the presence of S/Shri DO Tawade, Chairman, INGIABSE, RK Pandey, Secretary, ING-IABSE and Mahesh Tandon, Chairman, Scientific Committee as well as other dignitaries. Shri IG Reddy, Chief Engineer, R&B NH & PPP, Govt of Telangana extended warm welcome to the participants of the Workshop. Shri DO Tawade, delivered his address during the Inauguration. Shri RK Pandey, Secretary, ING-IABSE proposed Vote of Thanks. The Workshop on "Code of Practice for Concrete Road Bridges IRC:112:2011” was addressed by the following eminent experts covering the following themes who were either involved in the preparation of the Code or have used in extensively since its publication.

Shri T Nageshwar Rao, Hon’ble Minister, lighting the traditional Inaugural Lamp along with high dignitaries

A view of the Dais during the Inaugural Function

Shri DO Tawade, Chairman, ING-IABSE Delivering his address

Shri RK Pandey, Secretary, ING-IABSE Delivering his address

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Volume 45 Number 4 December 2015  xi


Shri T Nageshwar Rao, Hon’ble Minister, Minister of State, Telangana, PWD Delivering his address during Inaugural Function

A view of the Dais during the Valedictory Function

30th October, 2015 Session-1 1 Prof Mahesh Tandon

- Overview & Scope - Basis of Design

2 Shri Alok Bhowmick

- Actions and their Combinations

3 Prof Mahesh Tandon

- Material Properties and their Design Values

Session-2 4 Shri Vinay Gupta

- Analysis

5 Shri Umesh Rajeshirke 6 Shri Vinay Gupta

- ULS of Linear Elements for Bending and Axial Forces - Worked Example for Bridge Design with IRC:112 - Serviceability Limit State

31st October, 2015 Session-3 7 Shri JS Pahuja

- ULS of Shear, Punching Shear and Torsion

8 Shri VN Heggade

- ULS of Induced Deformations

- ULS of Two and Three Dimensional Elements for Out of Plane and In-Plane Loading Effects

9 Shri Vinay Gupta

- Durability and Deterioration of Concrete Structures

Session-4 10 Shri Alok Bhowmick

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- Prestressing System - Detailing Requirements Including Ductility Detailing

Number 4 December 2015

The Bridge and Structural Engineer


The concluding remarks of the Workshop were presented by Prof Mahesh Tandon, Chairman, Scientific Committee on 31st October 2015. He expressed the hope that the outcome of the Workshop would have enriched the delegates. The delegates who attended the Workshop mentioned that the subject matter of the Workshop is very timely. Shri IG Reddy, Chief Engineer, R&B NH & PPP, Govt of Telangana proposed Vote of Thanks. A cultural programme was organized in the evening of 30th October 2015 for the participants who rejoiced the evening. The Workshop was a great success.

Call for Papers – Seminar on “Urban Transport Corridors” The Indian National Group of the International Association for Bridge and Structural Engineering (ING-IABSE) is organising a Seminar on “Urban Transport Corridors” in association with Govt. of Andhra Pradesh at Vishakhapatnam in the 3rd week of June, 2016. The Seminar will have four Technical Sessions covering each theme in one Session as per the following: i) ii)

Policy and Planning Unified Urban Transport Development Authority  Planning for Multi-modal Transport for Urban Corridors  Transit Oriented Development including Land Use Planning 

System and Engineering Demand and Supply Management in Urban Transport  Infrastructure Requirement for Integrated Urban Transport  Use of ITS – Coordination, Efficiency, Monitoring, etc. in Urban Transport.  Safety and Security 

iii) Financing  Innovative Financing for Urban Transport Corridor.  Congestion Charging for Demand Management (including Parking) iv)

Case Studies Metro  Mono-Rail/LRT  BRTS  Intermediate Public Transport (Auto, Taxi etc.) 

Technical papers under various themes are invited for inclusion in the Seminar Report. The paper should be neatly printed including figures, tables etc. on A4 size paper with 25 mm margin on all side using 11 size Font (Times New Roman). Those who are interested to contribute a paper, kindly send their paper (maximum 9 pages plus one cover sheet) by 6th May 2016 at the following address. Selected authors will be invited to present their papers in the Seminar. Shri RK Pandey Secretary Indian National Group of the IABSE IDA Building, Ground Floor, Room No.12 Jamnagar House, Shahjahan Road New Delhi-110011 Tel:

011-23388132, 23386724

E-mail: ingiabse@bol.net.in, ingiabse@hotmail.com

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Volume 45 Number 4 December 2015  xiii


With Best Compliments From :

Reliance Infrastructure Ltd Mumbai

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The Bridge and Structural Engineer


INDIAN INNOVATIONS IN FOUNDATION ENGINEERING Dr. N.V. NAYAK Principal Advisor Gammon India Limited Mumbai – 400 025 nvn@gammonindia.com

Narayan V. Nayak is born in the year 1936 and graduated in Civil Engineering from the University of Bombay in the year 1959. He secured his M.Tech. in Civil Engineering from the Indian Institute of Technology, Bombay and his Ph.D from the University of Wisconsin, U.S.A. in the year 1970 (GPA, 4.0/4.0). He has 8 years of teaching, 8 years of consultancy & 35 years of practical experience.

1. Synopsis In this paper, few innovations developed by the author are illustrated. Some of them may look trivial, but will have significant effect in improving speed, quality, cost reduction and sustainability. Such innovations developed by the author are numerous but only few of them are illustrated in this paper. For more innovations, one may refer to author’s book “Foundation Design Manual” – 7th Edition.

2.

Removing Concrete above Cut off

Majority adopt the percussion method of removing concrete above the pile cut off as illustrated in Fig. 1. Very recently some have introduced pile breaker (Fig. 2 & 3) to remove the concrete above cut off. In both these cases, minor to mega cracks develop in the process and this permits entry of water into pile concrete, thereby leading to early corrosion, etc. Further chipping as above is extra cost and involves extra time.

Fig. 2 : Chipping by Pile Breaker

Fig. 3 : Pile breaker Chipping

Fig. 1 : Chipping of Concrete above the pile cut-off by jack hammer and Getting it ready for Wedge-on-method of removing concrete above cut off

The Bridge and Structural Engineer

The author developed a simple method, which consists of removing green concrete. This removal of green concrete can be done by tumbler when cut off is near the ground level or by a special tool, when cut off is at greater depth from the ground, developed by the author. (Fig. 4). Volume 45 Number 4 December 2015  1


of Piling Specialists the compaction by the rammer proposed by the author is preferable to compaction by vibrator. When the concrete is compacted by vibrator, arching action may take place between the concrete and guide case.

Fig. 4 : Removal of Green Concrete by Scooping Tool

Once the green concrete is removed as above, it should be compacted very well by vibrator or by ramming by a special tool, once again developed the author. (Fig. 5). In absence of such compaction, strength of concrete near cut off can get greatly reduced because

Fig. 6 : Air Voids

3. ‘L’ Bend to Reinforcement Cage at the Pile Tip

Fig. 5 : Rammer

of the presence of air voids/loose pockets/honey comb. In general, 1% air voids reduces the strength of concrete by 5% (Fig.6). As per the Federation 2  Volume 45

Number 4 December 2015

Providing ‘L’ bend (Fig.7) is an old practice. Inspite of many limitations to this practice, the same is continued by many, Roughly around 50% consultants/ contractors continue with such a practice. Provision of such an ‘L’ bend prevents cleaning of muck at the pile tip (resulting in soft toe) and obstructing the flow of concrete which may cause soft toe and even non formation of concrete shaft at toe (Figs. 8 & 9). The Bridge and Structural Engineer


Under ream No concrete mainly due to L - Bend Fig. 9 : Complete Concrete Shaft is missing at Pile Tip because of L Bend (Pile is under ream Pile) Fig. 7 : Reinforcement Cage with L bend at Pile Tip- Such L bends shall be eliminated

4. Socketing of Piles in Weathered/Sound Rock At present, the tendency is to specify more or sometime significantly more socketing than essential in weathered/sound rock. This means extra cost and time. Excessive chiseling of rock sometimes can cause distress/damage to the adjacent structure.

Fig. .8 : Tip of pile, cast with L bend in rebar cage. Note the taper and soil within the Reinforcement cage

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Widely adopted rational analysis is to determine socket length based on Cole and Stroud approach which is also now recommended in IS 2911 Part 1 Section 2 (2010). Cole and Stroud approach derives the strength of weathered rocks based on extrapolated N Values of Standard Penetration Test. When this approach is adopted in designing the pile socket length in rock, some consultants/clients request to verify the N values of standard penetration test by carrying out the test. This involves considerable loss of time and production. Author has developed a system by which torque needed to cut/penetrate the rock in socket length is related to pile penetration ratio which in turn related to N value of SPT (Table 1). Pile penetration ratio is the energy in tm required to advance the bore of one sq.m. by 1 cm (penetration). Torque utilized in penetrating rock can be easily read by torque meter fitted on the machine. Such torque meters are not Volume 45 Number 4 December 2015  3


provided on majority of machines supplied in India. However, the author got it fixed on all machines available with Gammon India Ltd. Table 1 : N of SPT vs PPR Values Sl. N Value PPR Value Remarks No. of SPT tm/m2/cm 1.

50

37.35

2.

100

74.70

3.

150

112.05

4.

200

149.40

5.

250

186.75

6.

300

224.10

7.

350

261.45

8.

400

298.80

5.

For Values in between, linear interpolation can be adopted

This table is provided to conveniently get PPR value corresponding N of SPT

Cole and Stroud approach is to be adopted for N of SPT upto 400

Piling Under Water

Whenever piling is to be carried out under water such as in backwater of sea, rivers and lakes, it is always advisable, if possible convert into land piling. This will result in tremendous reduction in cost and time.

Author has executed many such water piling works for Konkan Railway project and road projects. Maximum depth of water where such piling works were carried out was 14m on Godavari River.

6.

Cover Blocks to Reinforcement

Quite often cover blocks are made by hand mixing the concrete mix. This practice should be discontinued. The grade of concrete of cover blocks should be same as grade of concrete in structural element in consideration or higher. Even if it is same grade of concrete, it is hardly compacted. Very few organisations use vibrating table for compaction. In such situation, it is desirable to use precast cover blocks. Such blocks of M60 concrete are readily available in market. M80 concrete can also be specifically made on request. Or such cover blocks can be made at site duly compacted using vibrating table. Some adopt PVC cover blocks instead of concrete. Such PVC blocks are not desirable from strain compatibility. Cover blocks are often tide incorrectly with loose ends of binding wire projecting out rather than inside as shown in Fig.11. It is essential that loose ends of binding wires only project inside of cover and not outside. Otherwise these loose ends will be initiator for corrosion.

Converting to land piling is very convenient if sediment over the river/lake/backwater of sea is sand. Using excavators such sand is utlilised in making the earthen bund. It may be noted that for such bunds, slope below the lowest low water is more safe than above. Slope above lowest low water needs to be protected by pitching or by turffing as shown in Fig. 10. Also it is incorrect practice to drive wooden bully piles near the toe presuming it will add to factor of safety against slope failure and it will minimize earth filling. Driving such bully piles needs to be discontinued as failure can occur with slip circle passing below pile tip as shown in Fig. 10.

Fig. 10: Converting Marine Piling to Land Piling by Bunding

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Fig. 11: Reinforcement cage showing binding wire protruding inside the cage to prevent corrosion

The Bridge and Structural Engineer


7.

Well Resting on Piles – Innovative Design and Execution

So far intentionally very simple innovations have been highlighted but which have got significant impact on increased durability, improved sustainability and cost reduction. However, many complicated innovations have also been developed by the author and these are highlighted in the author’s book of “Foundation Design Manual – 7th Edn.” Of these complicated innovations, one innovation viz; well resting on piles has been illustrated in this paper. This is probably for the first time that such an innovation has been carried out in any part of the world. The case in consideration, refer to Konkan Railway Bridge on Zuari River in Goa. For this bridge, practically all types of foundations have been adopted viz shallow foundations, pile foundations and well foundations. The well foundations were pneumatically sunk. One such well foundation was required to be founded at depth of 30m below the water level. This means people working under pneumatic conditions can hardly work for half hour as practically all the time is spent in compression/decompression chamber. Hence, a novel idea was developed and implemented. Ideas involved resting well foundation on bored cast-in-situ piles, 8 numbers, each of 1500 mm dia. These piles were founded 35 m below water level and embedded into bottom slab of well foundation as shown in Fig. 12. For the scheme to be successful it

was absolutely essential to ensure the piles are executed in correct position and vertical deviation beyond certain limit will make piles to be on cutting edge or outside. Secondly we had to ensure that concreting above the cut off is done to the required level. As per the Indian Codal requirement, height to which concreting is to be done above cut-off was working out more than 7 m. But the author would not subscribe to the same. However, to ensure that concrete does not fall below the cut off level, author decided to cast concrete 1.5 m above cut-off. Though the piling was done by tripod piling from a temporary platform, it is to the credit of site team that piles were installed to the correct locations and piles were concreted practically 1.5m above cut-off. Chipping of piles to the required level took 20 days under compressed air. Here the author would like to convey his heartfelt thanks to Mr. E. Sreedharan for having shown his confidence in the Author and his team and permitting such a novel idea to be implemented. Hardly any other Government officer may adopt such a path.

8.

Need for Revision of Codal Stipulations on Permissible Settlement Under Safe Pile Load

Bureau of Indian Stands (BIS) has been highly conservative in stipulating permissible settlement of a pile under safe load. As per IS 2911 Part 4 (1985) which was prevailing till revision in 2013, the permissible settlement at 1.5 time the design/safe allowable load was 12 mm, independent of pile diameter (this was for uniform diameter of the pile). As per world wise practice including India, ultimate load capacity of the pile is taken as load at which pile settlement equal to 10% of pile diameter is reached. Ref. Fig.13). Safe load has to be fraction of ultimate load. Hence, permissible settlement at allowable/safe load or 1.5 times allowable/safe load has to be dependent on pile diameter. In the revision of IS 2911 Part 4 of 2013, the permissible settlements at 1.5 times the safe allowable load are :

Fig. 12 : Details of Caisson Resting on Piles

The Bridge and Structural Engineer

a.

12 mm upto and including pile diameter 600mm

b.

2% of pile diameter from 600mm to 900mm pile diameter.

c.

18 mm for pile diameter 900mm and above. Volume 45 Number 4 December 2015  5


The same is illustrated as curve (b) in Fig.13.

9. Conclusions

From the curve it can be easily see that earlier stated limitation applies here also for diameter greater than 900 mm and below 600 mm. Author strongly feels that we specify permissible settlement at 1.5 times the safe/allowable load as certain percent of pile diameter.

Innovations are “key” to success. Fortunately Indians are very capable of innovations in any field. Unfortunately they are not encouraged by the Management or Governments as is the case of developed countries particularly USA. In this paper, the author has illustrated many innovations developed by him in Foundation Engineering. Intentionally he has illustrated all but one very simple innovations to bring out how even these simple innovations can improve quality, sustainability, speed of execution and reduced the cost. He has illustrated only one innovation which is more thought provoking and needed extreme care during execution. The Author sincerely believes that this paper will encourage young students and engineers to the path of innovations.

10. References Fig.13: The Limitations of the revised BIS Code: IS 2911 Part 4:2013 on permissible settlement of the pile at 1.5 times the design load

It could be 3% pile diameter as in curve (c ) of Fig.13 or 2% pile diameter as in curve (d) of Fig.13. Still more rational approach would be to specify permissible settlement at design/safe load as “elastic settlement of pile shaft at design/ safe load (to be estimated) plus n (diameter of the pile)” where n could be 2% of pile diameter or conservatively even 1.5% of pile diameter.

6  Volume 45

Number 4 December 2015

1. Bureau of Indian Standards, IS 2911 Part 1 Section 2, 2010. 2.

Bureau of Indian Standards, IS 2911 Part 4 1985.

3. Bureau of Indian Standards, IS 2911 Part 4, 2013. 4.

Cole, K.W. and M.A. Stroud (1976) “Rock Socket Piles at Coventry” Geotechnique Vol. XXVI No.1.

5.

Nayak, N.V. (2015) “Foundation Design Manual- 7th Edn. Dhanpatrai Publications.

The Bridge and Structural Engineer


Pile Load Testing in India - Current Practices and Recent Developments

Sunil S. BASARKAR General Manager ITD Cementation India Limited, Mumbai, India Sunil.basarkar@itdcem.co.in

Ravikiran VAIDYA Principal Engineer Geo Dynamics, Vadodara, India ravikiran@geodynamics.net

Sunil S. Basarkar is the Head of Technical services and Quality at the Ground engineering division of his organization. He holds a specialization in Geotechnical engineering, with doctoral research in rock-socketed piles. He has 29 years of academic and field experience.

Ravikiran Vaidya has popularized various methods of Non Destructive Testing like high strain dynamic testing, integrity, sonic logging etc. in the country. He has 18 years of experience in this field and has worked extensively in India and Middle East. He holds a Masters in Structural Engg. from M.S. University of Baroda.

Summary Pile load tests are essential to confirm and validate the choice of safe design load and serve as a valuable tool in quality assurance during execution of working piles. Current practice in India still places reliance on conventional tests using kentledge and reaction anchors, considering their ease to test for loads upto 1500MT. There is a growing trend of adopting higher pile loads and this development has given a fillip to advances in testing methodologies and electronic monitoring for assessment of both, external and internal pile performances. Fast track projects, higher test loads and site conditions have given way to alternative methods like Dynamic pile load tests which are now common and Bi-directional loading methods, which since recent years is gaining wide acceptance. This paper presents applications of load tests in conveying vital pile behaviour, their critical assessment and the way ahead vis-Ă -vis Indian scenario. Keywords: Pile load test; Dynamic load test; Bi-directional load test; Working piles; Safe The Bridge and Structural Engineer

load;Maximum test load; Instrumentation; Load transfer; Maintained load; Cyclic load.

1. Introduction The performance and behaviour of pile foundations resting in a variety of geomaterial is well established. In spite of this fact, the behaviour of these foundations is dependent on local geology, construction methodology and workmanship. These factors make it mandatory to establish pile behaviour under field specific conditions through load tests. Pile load tests, apart from fulfilling traditional roles of design validation and routine quality control tool, nowadays is increasingly utilized as a tool for optimization and continuous improvisation of foundation design and construction practices. Current Indian practice places heavy reliance on traditional load test practices using kentledge or reaction piles. With increase in pile loads, advanced load testing systems have gained entry in India. Dynamic pile load tests as per ASTM D4945 [1] are well accepted after establishment of its correlation with static load test at the site, while Volume 45 Number 4 December 2015  7


Bi-directional testing, with rise of the testing loads, is slowly establishing its foothold in India. Attention has also increased on pile instrumentation to evaluate frictional resistances and to assess load transfer characteristics. This paper reviews current pile load testing methods, recent advances and their benefits in delivering economic testing solutions and giving valuable parameters hitherto not common in India.

2.

Load Test Classifications, Objectives and Strategies

Testing of piles by direct top static loading still remains one of the most accepted assessments of the pile load-displacement behaviour. Such tests are used to confirm the outcome of the fundamental pile design; and also form a part of quality assurance process on the working piles. Preliminary pile design is first carried out on the basis of site investigations, laboratory soil testing, and office study. Initial pile load tests are then carried out to refine and finalise the design load. In Initial load tests, performance of piles under ultimate conditions is intended. These piles are generally tested 2 to 3 times the estimated safe design capacity. Routine pile load tests are carried out on randomly selected working piles to confirm their intended performance. In these situations, the piles are generally tested to 1.5 times the design capacity. Such tests also serve as valuable quality assurance tool. Mode of load applications varies in pile load tests. In Maintained load method, application of load increment and displacement measurement at each stage is carried out till its rate is observed to reach a limiting value of around 0.2 mm/hr. Cyclic load application is used mostly during initial test to separate frictional and end bearing resistances of a single pile of uniform diameter. For most projects the main purpose of pile testing is to confirm the design choice of safe load before execution and, in addition, check for compliances to the contract specifications during construction. A broad categorization of objectives may be: (i) design validation, (ii) quality control, (iii) design development and, (iv) research. Basic objectives of load tests are well conveyed by FHWA-NHI [2]: 8  Volume 45

Number 4 December 2015

(i) To obtain detailed information on load transfer in side and base resistance (or lateral soil resistance for a lateral load test) to allow for an improved design (Load transfer test), or (ii) To prove that the test shaft, as constructed, is capable of sustaining a load of a given magnitude and thus verifying the strength and/ or serviceability of the design (Proof test). In fact, the testing strategy for any load test programme should be aimed to optimise the pile design in terms of its geometric dimensions and marginalise the factor of safety to the best extent. Federation of Piling Specialists [3] identifies risk levels as High, Medium and Low depending on complexity of ground conditions and availability of performance history of piles. According to this risk level, guidelines for deciding the frequency of Initial and Routine load tests are fixed. Any interpretations of static load tests results require a thorough understanding of the load transfer behaviour of the piles. In this context, papers by Coyle and Reese [4] and Vijayvergiya [5] discuss intrinsic mechanism of load-movements resulting in mobilization of pile side shear and base resistances.

3. Conventional Reaction Sustaining Systems and Applications in India For vertical load tests, load application in conventional practice is by top down technique through a set of hydraulic jacks. Ground conditions and test load magnitude will generally govern the reaction sustaining methods and are discussed in following sections. Applications for pull out are also stated. 3.1 Top down load applications Common reaction sustaining technique is by Kentledge method, where reaction arrangement exists in form of concrete blocks or any counter weights that rests on set of secondary girders, which in turn are supported by primary girder. The primary girder is sandwiched between the hydraulic jack and the secondary girders. The counterweights have to be sufficiently higher than the test load to prevent cantilever action of the secondary girders during loading. As an illustration, reaction by Kentledge load of 18.75MN for a high rise residential tower at Kolkata can be seen in Fig.1. This arrangement was The Bridge and Structural Engineer


made for Initial load test on a pile to validate its 9.00 MN design capacity by adopting a low load factor.753 concrete cubes of 1 m3 size each were used with a loaded area admeasuring 12 m x 11 m. Test pile was 1000 mm diameter and 58 m length, passing through silty clay/clayey silt for top 51 m (SPT N varying from 15 to 75), with lower 7 m resting on dense sand with N>100.

Fig.1: Load test using kentledge arrangement

Fig. 2: Load test using anchor piles

Reaction Piles or Active Anchors are more convenient where rock and competent stratum are available at a shallow depth. Such arrangement consists of two or more reaction piles or active anchors located on either side of a test pile. In order to minimise the interference between test and anchor piles, a minimum distance of three times the pile diameter is maintained between these piles. A reaction beam is placed on top of the anchor piles and the test pile is loaded by utilising a hydraulic jack placed co-axially on top. This results in applying compressive load on the test pile and uplift load on The Bridge and Structural Engineer

the anchor piles. Since the equipment and girder requirements are small, foot print area is relatively small; for instance, a test load of 8.00MN with reaction arrangement comprising four active anchors of 2.20 MN capacity would require and area of approximately 6 m x 2.2 m area. In one of the sites at Karwar located on west coast of India, a vertical load test was conducted. Steel piles were driven on reclaimed area through a larger diameter MS Liner upto weathered rock level, so that contribution only from socketed rock was available to the pile. This arrangement modelled the behaviour of marine piles, designed for resistance arising solely from weathered rock. Safe design load on this driven steel pile 850 mm outer diameter and with a conical bottom shoe was 3.50 MN. A 9.00MN reaction arrangement was made comprising eight rock anchors, four on each side (Fig. 2), each of 1.16 MN safe capacity, deriving strength from 9 m fixed length in rock. A similar arrangement was also made at Tuna port near Kandla, Gujarat. A test pile 1500 mm diameter and 35.50 m depth, when loaded to 21.103 MN indicated displacement of 8.98 mm with observed rebound of 7.50 mm. Sixteen active anchors each of capacity 1.32 MN, with arrangement for housing four anchors at each corner, each with 10 m fixed length in amygdaloidal basalt were installed. The test pile, after 22 m of clayey silt/ silty clay layers, passed through 10m of highly weathered amygdaloidal basalt (average UCS of 25 MPa), and then socketed about 3.5 m in moderately weathered amygdaloidal basalt (core recovery excess of 97% and RQD in a range of 25 – 50%) with UCS of 100.0MPa. In another instance, at Noida, near Delhi, piled raft supported on bored cast-in-situ pile of 1200mm diameter was proposed for a high rise commercial tower. Load test program was planned to determine and maximize the safe pile capacity to achieve most economic pile-raft combination. The ground conditions were daunting, since the site was situated at close proximity of Yamuna river and comprised loose to medium dense sand to about 5 m (SPT N values between 4 to 34) followed by a medium dense to dense sand (N values 10 - 93) upto an investigated depth of 40 m. Pile bore stability was the critical factor identified and a combination of stability fluids were suggested for arriving at an economic and Volume 45 Number 4 December 2015  9


reliable piling proposal. Five test piles were planned, of which three load tests of relevance to this paper are summarized in Table 1. Table 1: Noida load test details Test Pile Pile Bore Pile diameter length hole no. (mm) (m) stability fluid

Computed Applied design load maximum (MN) test load (MN)

Pile top displacement (mm)

TP1

1200

40.00

Bentonite 5.40

7.82

131.58

TP3

1200

34.00

Polymer+ 5.40 Bentonite

17.00

127.642

TP4

1200

34.00

Polymer

17.00

21.40

5.40

A reaction arrangement for all these load tests comprised a 22.00 MN capacity kentledge, with maintained loading method as per provision of IS: 2911 (Part 4) [6]. Comparative load-displacement curves for these tests are reported in Fig. 3, and the safe loads revealed for TP1, TP3 and TP4 were 3.13, 6.17 and 9.80MN respectively.

to the manufacturer’s product literatures may improve frictional resistance at pile-soil interface, and becomes an important factor contributing to increase in pile capacity. The conclusions on use of polymer fluid are based on site specific performance and should be used with caution at this time. Before generalization, dedicated laboratory and field studies in various soil conditions are desired to conclusively prove a positive contribution of polymer based slurry in improving the frictional contributions. 3.2 Pull out load tests Pull out load tests are carried out by connecting the test pile reinforcement to a primary girder, which is upward loaded with hydraulic jacks. A single hydraulic jack of high capacity may be co-axially placed above the test pile; or alternatively, jacks are placed atop reaction piles. At a project site at Kalma, located near Raigad town at Central India, a pull out load test (Fig. 4) was conducted to confirm performance of a 900 mm diameter RCC bored raker pile with a safe uplift load of 2.02 MN. The pile passed through 4.50 m of sandy clay, and 2.50 m of soft rock, before being socketed 6m into hard basaltic rock. Summary information of this load test is reported in Table 2.

Fig. 3: Borehole stability fluid performance - comparative load-displacement curves

As seen from the comparative curves, test pile TP03 with engineered polymer fluid was able to deliver a very high pile capacity. With increase in pile capacity, a significant reduction in the raft size and hence overall foundation cost became evident. At this point it is important to note that despite best engineering attention in field, for large diameter piles stabilized with bentonite slurries, complete cleaning of pile bores is still a matter concern. Under such situation, ability of polymer fluids, which according 10  Volume 45

Number 4 December 2015

Fig. 4: Pull-out arrangement on a working pile at Kalma

A 5.00 MN capacity hydraulic jack was used, and pullout load was applied by lifting the main rebars of the pile through the jack load, and transferring the load to the primary girder by 22 nos. of 32 mm diameter MS bolts (Fig. 4). Average rock-pile bond mobilized under test Load was 0.1265 MPa, which was very low indicating very high reserve strength in Pile. The Bridge and Structural Engineer


Table 2: Pull-out load test information – Kalma case study Pile No. Pile dia. (mm) P05

900

Pile depth (m)

Pile type Safe pull Test load out load (MN) (MN)

12.96

Bored castin-situ working raker pile (1H:10V)

2.02

3.03 (applied co-axially)

4. Measurement of External and Internal Pile Performances Two main types of movement measurements in a load test are pile head displacement and incremental strain measurements internally along the pile length. 4.1 Pile head displacement measurements Pile head displacements are required in all pile load tests and are measured with dial gauges or Linear Variable Displacement Transducers (LVDTs). Dial gauges are used in areas where the test pit requirements are shallow and easily accessible. In the areas requiring deep test pits where, due to safety reasons when access is restricted during loading, LVDTs are often deployed, with their read-out unit at safe and convenient location. A surveyor’s level may also be used for measuring the axial pile movement; more preferred as a matter of cross check rather than as a primary means of movement measurement. 4.2 Common instrumentations for incremental strain measurements

extended to steel end plates embedded inside a concrete pile or welded on the steel pipe at various locations along the pile length. Normally, telltale readings are referenced to the top of the pile. By noting the location of the specific telltale rod anchor plate and by measuring the relative movement of the individual rod, elastic shortening of pile at that location can be obtained. 4.3 Instrumentation case study Pile instrumentation is adopted particularly to understand the load transfer characteristics or when history of pile performance at the project site is not available. For a particular project site at Noida, instrumented pile load test was performed on one of the test pile TP02 of 1000 mm diameter and 35m depth below the cut-off-level. Instrumentation comprised four levels of vibrating wire strain sensors (Fig. 5), four at each level as indicated. The test pile was step loaded to 15.11 MN through 13 loading increments (Fig. 6). Strain readings were recorded and converted to appropriate load values with concrete elastic modulus, Ec 25000.00 MPa (for M25 grade pile concrete), using relationship Q=A.Ec.ε where, Q is the pile load at the strain gauge level, A= CS area of the pile & ε= measured strain. Using strain information and interpretations, curves for strain. vs. pile top load (Fig. 7), load distribution curves (Fig. 8) were plotted. Information furnished by the instrumentation is compiled in Table 3. Table 3: Unit load contributions in Test pile TP02 under 12.09 MN maximum pile top load

Incremental strain measurements are used to determine the distribution of load transfer from the pile to the soil and are generally considered as an optional measurement feature, but are finding increasing use in India.

Pile depth range (m)

Load transferred (MN)

Unit shear mobilized (KPa)

0.00 – 13.10

2.174

52.80

13.10 – 23.10

3.723

118.50

Electronic strain sensors of various forms are available which can be mounted along the pile length at various locations before the pile is installed. In piles, these sensors can be welded to the reinforcing bars and wires can be brought near from the top through a PVC casing. Strain sensor output are read electronically using portable readout boxes or with dataloggers.

23.10 – 32.10

4.755

168.18

32.10 – 35.00

1.436

-

Telltales or Strain rods also form one of the vital instruments and normally consist of PVC tubing The Bridge and Structural Engineer

Remarks

Includes shear and base contribution

Instrument observations indicated that strain gauge readings at -3.1 m could not be relied, probably due to lateral spread of the applied load along the pile cross section. Again, strain gauge readings beyond application of 12.09 MN load could not result in Volume 45 Number 4 December 2015  11


Fig. 8: Pile load distribution along depth

Fig. 5: Scheme for pile instrumentation

logical load interpretations, and hence were overlooked. With these exceptions, the instrumentations had been able to convey the unit shear contributions and had affirmatively confirmed predominant frictional pile with 88.12% of the load resisted through friction in top 32.1 m length of the shaft.

5. Dynamic Pile Load Test – Advent and Practices in India

Fig. 6: Pile top load-displacement curve

Fig. 7: Strain .vs. pile top load curve

12  Volume 45

Number 4 December 2015

High Strain Dynamic Load Tests (HSDPT) are the most common method to test the piles after static load tests. The method has now been routinely accepted across sectors like highways, railways, power, marine, oil & gas and real estate projects. The most common practice is to conduct some reliability studies by loading the same pile statically and dynamically and compare the findings before allowing high strain dynamic tests at the project site. Based on the criticality, number of piles and the stretch of each project, one or more reliability study is conducted at each project site to verify the HSDPT output. Correlations or reliability studies provides checks on the reliability of HSDPT in unknown stratum before allowing its use. Since the method requires documented expertise and high integrity of the personnel conducting the test, such studies also help confirm the expertise of the personnel or the agency conducting the tests. The test is conducted as per ASTM D4945 [1]. For bored piles, the method involves impact of a hammer of 1-3% of the test load and 7-10% of the self-weight The Bridge and Structural Engineer


of the pile, whichever is heavier. Sensor readings for strains and accelerations are measured and converted to forces and velocities in the pile. The field capacity is obtained by a series of impacts and the test is stopped. The data obtained is then modelled using proprietary software CAPWAPTM. This software discretizes the pile into a series of elements and helps in obtaining the soil resistances along the length and by end bearing. The software generates a Match Quality (MQ) number which indicates the quality of analysis. MQ of less than 3 or in some cases upto 5 are most preferred for acceptance of the test results. A recent technical paper by Vaidya, et al. [10] documents in detail the parameters to be monitored before acceptance of the test results. It should also be mentioned that the high strain dynamic test method correlates well the Davisson’s criterion of failure which is more conservative compared to the values derived from the current Indian code of piling practice [6].

A database of more than 150 such case studies in a variety of soils is now available across India and some of it is also available as published literature (Basarkar et al. [11], and Mhaiskar et al. [12]). With availability of large database is available, the authors have used the method for initial load tests at several project sites to obtain the ultimate load which is difficult to obtain with static load tests if design is conservative. Two reliability studies in rock and in soil are mentioned. Fig. 9 (a) shows a reliability study for a major infrastructure project in Mumbai. The pile was 1m in diameter and with a short depth of 9.4 m installed into rock for bottom 2 m. Fig. 9 (b) shows a reliability study for a project in Delhi for 1m diameter pile with depth of 27 m installed into a stratum that had fill for 2 m, followed by dense to very dense brown silty sand from 2 m to 12 m and further from 12 m to 30 m. 5.1 Pullout tests using HSDPT

(a)

Several published literatures including those of authors have reported good match between computed and measured values of soil resistances along pile length and end bearing when ultimate pile capacity is reached. The friction obtained from the static or dynamic load tests is basically related to the uplift capacity of the pile. After allowing for some ambiguity in modelling of friction in HSDPT, typically 80% of the friction obtained from the HSDPT test is considered as the uplift capacity of the pile [17]. The second author has used this approach at several projects successfully and a database is also available for driven piles.

6.

Bi-directional Pile Load Tests in India

Bi-directional technology employs Osterberg type Load Cell (O-cell). An O-cell is a sacrificial jack like device attached to the pile rebar cage [7]. Hydraulic lines and telltales extend from the O-cell to the top of the pile to monitor the movement of the pile base as the O-cell load is applied (Fig. 10). Since the O-cell after being loaded, derives reaction from the pile-soil and pile-base resistances, it is important that the location of placing of O-cell is correctly identified and the pile does not have any construction deficiencies particularly at the bottom. (b) Fig. 9: Reliability studies on dynamic load test. (a) Mumbai case study (b) Delhi case study

The Bridge and Structural Engineer

Developments and advantages of Bi-directional tests in India using O-cell are well brought about by Ayithi et al. [9]. According to Ayithi et al. [9], first O-cell tests were Volume 45 Number 4 December 2015  13


to facilitate measurement of pile compression above and below O-cell. Paired strain sensors, four above and three below the O-cell were installed to study load transfer. A maximum bi-directional load of 38.00 MN was applied through O-cell. For a socket rock comprising weathered Tuff, design value of side shear resistance was 0.1 MPa. The O-cell strain gauge data revealed a very high side shear resistance of 1 to 2 MPa. This load test output lead to reduction of the length of the working piles to 19m, thus contributing to a significant savings in foundation costs. All structural loads were taken entirely by side shear resistance of socketed piles. O-cell test carried out on two optimised working piles 2894-1 & 2894-2 (Table 5) indicated similar capacity as the test piles. Table 5:Worli Crest Tower case study:O-cell test results Fig. 10: Schematic of O-cell based pile testing [8]

conducted in India during 2001 for BandraWorli sea link project at Mumbai. Till date, about 15 O-cell tests have been performed in India and include four load tests at iconic World Crest Towers and Infra Wadala (2 nos), both at Mumbai. One case study is briefly outlined conveying the application to high capacity sockets for optimization. 6.1 World Crest Tower, Mumbai case history[9] At World Crest Towers, two tests were performed on dedicated piles with information summary reported in Table 4. Table 4:Pre-design O-cell test on pile Test pile no. 2873-1 2873-2

Designation Test pile Test pile

Diameter (mm) 1200 1200

Length (m) 26.70 26.70

The test piles after 3m fill layers were taken into weathered to slightly weathered Tuff layer. This single O-cell assembly system located 4m above the pile tip, had instrumentation consisting of paired compression telltales and pile toe telltales 14  Volume 45

Number 4 December 2015

Test pile no.

O-cell expansion (mm)

2873-1

Maximum load applied (MN) 38

3.97

Max. side shear mobilized (MPa) 1.55

2873-2 2894-1 2894-2

38 34 31

14.10 3.58 3.21

2.00 1.66 1.00

Max. end bearing mobilized (MPa) None transferred to toe

The above case study illustrates a safe and reliable application of O-cell for high capacity load application hitherto considered very tedious for conventional load tests. Through O-cell load application, complete mobilization of both side shear and end bearing are possible which can yield valuable insight on their resistances and lead to a highly economic pile design.

7. Critical Appraisal of Pile Load Test Practices in India In India safe vertical compression pile loads commonly vary from 0.60 MN to 15.00 MN, while uplift and lateral loads vary in the range of 0.20MN – 3.50 MN and 0.02 to 0.15 MN respectively. Kentledge loads beyond 20.00 MN requires large sized loading platform and invites objectionable safety concerns in The Bridge and Structural Engineer


India, though, there are reported instances of applying kentledge method for test load upto 27.50 MN. Under special cases abroad, reaction system using multiple reaction girders and anchor piles have been used to test upto 40.00 MN, with probably highest load of 57.00 MN performed in Taiwan [2]. Maintained load test is common and frequent practice to verify the safe loads, while cyclic load tests are generally preferred for an approximate assessment of side shear and end bearing contributions of a pile. Indian practice of this test is probably an offshoot of Slow maintained load method referred to as the standard loading procedure in the ASTM D-1143-81 [13]. Loading cycles are generally 20% of the computed ultimate load. Limitations of cyclic load interpretations are well eliminated by use of strain sensors along the pile. With the hardware and installation costs coming down, use of instrumentations is on upswing in India, though not at all at par with their use and reliance abroad. Among the strain sensors, the vibrating wire type is very common. The current practice is to install 2 or 4 strain sensors at the top, near the middle and then near the tip of the pile. This leads to a high degree of approximation. More sensors need to be placed and typically the distance should not be more than 3 to 5 m to facilitate better understanding of load transfer behaviour. In future, increasing use of instrumentation aided with better interpretation is expected to lead to optimization of pile design and installations. Bi-directional load tests have been used sparingly where project costs permits or when computing rock end bearing is essential. For initial tests where test loads higher than 30 MN are anticipated, Bi-directional tests can be preferred if the project cost permits. HSDPT offer speed, economy and reliability when site arrangements are proper and tests are done by expert personnel with good integrity. The method now is a part of contract documents of various government bodies and is well accepted in the real estate sector. Load tests from The Bridge and Structural Engineer

0.1 MN to 20 MN are commonly done across India. Typically 0.5-2% of the working piles are tested with HSDPT and at several projects now where past reference or database is available, the method also has been used for initial load tests. Since the method requires expertise, it is important that the end client knows basic monitoring during the test program, otherwise abuse of the tests in some instances have also been reported. Performance of large and small diameter piles as a separate entity needs to be appreciated. The Indian code does differentiates them by adopting an interface value of 600 mm and allocating separate displacement criteria of 12 mm and 18 mm respectively for small and large diameter piles to arrive at a safe load. Load test experiences on piles indicate that a better interface for large and small diameter piles would be 900 mm with displacement criteria for inference of safe loads as 15 and 20 mm respectively. Literatures on safe pile load criteria are abundant, but probably the best compilation with critical reviews is done by Fellenius [14]. The criterion proposed by Davisson [15] is more common in United States and several countries and also used for correlation with dynamic load tests. Chin [16] has proposed an extrapolation method to obtain approximate value of ultimate load although with high degree of variability. Load tests, unfortunately in India, has yet to see their frequent use as a tool for optimization and research, a reason probably attributed to fast track nature of the projects, where a green signal to proceed is accorded solely on the fact that the pile meets the minimum design load requirement, irrespective of the margins.

8. Concluding Remarks and Future Outlook The foregoing sections presented trends on pile load testing with specific reference to Indian practice. For a normal range of vertical compression pile loads, conventional test methods are commonly used, while high strain dynamic tests are increasingly used for routine piles. Volume 45 Number 4 December 2015  15


Future trend is to adopt large diameter piles, monopiles and rock socketed piles which will introduce a fresh demand for high capacity load tests. Considering the safety restrictions and reaction load feasibility, there is a scope for Bi-directional load tests with multiple use of O-cell which has a potential of applying loads even upto 300.00 MN. While, this load test has already gained acceptance in Indian conditions, albeit at higher cost, other technologies like Rapid load tests have not yet been performed in India due to various reasons that includes costs, permissions and database.In the current decade, embedded data collectors are finding applications in US and Europe particularly for driven piles. However, published literatures on their performances and experiences are scarce and therefore, more applied research is desired validating their performance for gaining global confidence and acceptance. Instrumented Pile Load tests with vibrating or sister bar strain sensors and tell tale extensometers as a tool for optimizing pile design will still take time, as with limited sensors only limited data is available, and there is hardly any discussion or emphasis on good interpretation. It is a known fact that if a combination of some of the methods mentioned in this paper is used at project sites, then the cost of such testing can be offset by the optimization that can be obtained with the results of these load tests. In addition, sizeable reduction in foundation costs is possible. The ability of pile load testing as a value engineering and geotechnical and structural optimization tool is yet to see light of the day in India. This practice would not only benefit in financial terms but has significant importance in sustainability.

References 1. ASTM D4945-00., Standard Test Method for High Strain Dynamic Testing of Piles, 2000. 2. FHWA-NHI-10-016., Drilled Shafts: Construction Procedures and LRFD Design Procedures, US Dept. of Transport and Federal Highways Association, 2010. 3. FEDERATION OF PILING SPECIALIST (FPS)., Handbook of Pile Load Testing, Kent, UK, 2006. 4. COYLE, H.M., and REESE, L.C., “Load Transfer of Axially Loaded Piles in Clay”, Journal of SM and FD, ASCE, Vol. 92 (SM2), 1966, pp 1 – 26. 5. VIJAYVERGIYA, V.N., “Load Movement Characteristics of Piles”. Proc. 4th Symp. of Waterway, Port, Coastal and Ocean Division, ASCE, Long Beach, California, Vol. 2, 1977, pp 269-284. 6. IS: 2911 (PART 4) (2013)., Indian Standard Code of Practice for Design and Construction of Pile Foundations, Part 4 – Load Test on Piles, 2013. 7. OSTERBERG, J.O., “The Osterberg Load Test Method for Bored and Driven Piles – the First Ten Years”, Proc. 7th International Conference and Exhibitions on Piling and Deep Foundation, DFI, Vienna, Austria, June 1998. 8. WWW.LOADTEST.COM

Acknowledgements

9. AYITHI, A., BULLOCK, P.J., KHOO, H.S., and RAMANA, G.V., “Technical and Economic Benefits of O-cell Testing for Deep Foundations in India”, Proc. Indian Geotechnical Conference, Roorkee, India, Dec. 22-24, 2013.

Developments in solid state electronics, computing power and internet has led to numerous advancements in pile load test technology, and authors salute the innovators for their contribution to this area of foundation engineering. Outcome of this paper is a result of mutual collaboration, and constructive interactions between authors’ firms, clients and consultants. Authors gratefully acknowledge this symbiosis.

11. BASARKAR, S., MANISH KUMAR, and VAIDYA, R., “High Strain Dynamic Pile Testing Practices in India – Favourable

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10. VAIDYA, R., and LINKINS, G., “Guidelines for successful High Strain Dynamic Load Tests & Low Strain Integrity Tests for Bored Piles”, Proc. Indian Geotechnical Conference, Kakinada, India, Dec., 2014.

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Situations and Correlation Studies”, Proc. of Indian Geotechnical Conference, Kochi, India. Dec. 2011. 12. MHAISKAR, S.Y., KHARE, M., and VAIDYA, R., “High Strain Dynamic Pile Testing & Static Load Test – A Correlation Study”, Proc. of Indian Geotechnical Conference, Mumbai, India, Dec. 2010. 13. ASTM D1143-81. Standard Method of Testing Piles under Static Axial Compressive Load, Vol. 04.08, pp 179-189, 1989. 14. FELLENIUS, B.H., “The Analysis of Result from Routine Pile Load Tests”, Ground Engineering, London, 13 (6), 1980, pp 19 – 31.

The Bridge and Structural Engineer

15. DAVISSON, M.T., “High Capacity Piles”, Proc. Lecture Series, Innovation in Foundation Construction, ASCE, Illinois Section, 1972, pp 52. 16. CHIN, F.K., “Estimation of the Ultimate Load of Piles not Carried to Failure”, Proc. 2nd South-East Conference in Soil Engng, 1970, pp 81-90. 17. TEFERRA, W., SAAVEDRA, M.R., and ECHANIZ, P., “Use of CAPWAP for Uplift Resistance Evaluation of Wind Energy Tower Piles”, Proc. 8th International Conference on the Application of Stress Wave Theory to Piles, Lisbon, 2008, pp 192.

Volume 45 Number 4 December 2015  17


Recent Advancements in Ground Engineering Techniques in India: An Overview

Dr V.R. RAJU M.D. Keller Asia 18, Boon Lay Way #04-104, Tradehub 21 Singapore 609966 vrraju@kellerasia.com

Madan Kumar ANNAM Tech. Manager, Keller India 7th Floor, Eastern Wing,Centennial Square, 6A, Dr Ambedkar Road, Kodambakkam, Chennai 600024 madankumar@kellerindia.com

A Geotechnical Engineer by training, Dr. V.R. Raju served as Managing Director, Keller Singapore & Malaysia in 1999 and as Business Unit Manager, Keller Far East in 2009, Managing Director, Asia and to the Group Executive Committee in 2012. Presently, Dr. Raju is Director of Engineering and Operations (Designate) in Keller Group.

Madan Kumar is a practicing Geotechnical Engineer having more than 20 years of professional experience in Geotechnical Engineering field in both Consultancy & Construction firms. Handled variety of Geotechnical Engineering problems during his career. His area of interest includes Heavy Foundations especially bored cast in-situ piles, deep excavations and retention systems.

Summary

1. Introduction

This paper summarises an overview of various types of soils needing improvement and the various ground improvement techniques available. Factors influencing the choice of technique are discussed and this is followed by some applications of these techniques to different types of structure and soil conditions. Structures and facilities that have utilized ground improvement include power plants, roads and highways, ports and airports, storage tanks, chemical plants, industrial structures and residential buildings. The basis for choosing a particular technique for the project is explored, be it time, cost, technical performance or environmental considerations. This paper will illustrate that ground improvement is often the ideal foundation solution for such structures.

The infrastructure projects such as power plants, highways, airports, ports and harbours, industrial plants cover large area of land often leads to highly variable soil conditions within the project boundary. A large portion of infrastructure projects located in coastal regions, where soils typically have low strengths and are highly compressible. Construction in increasingly urban environments means that sites with poor soil conditions and even landfills are being utilized for various structures and facilities. This construction activity on poor soils often leads to the necessity for ground improvement prior to start of construction. The type of ground improvement required depends very much on the type of structure to be built (and its sensitivity to ground improvement), the type of soil being treated (and its short and long term behaviour) and the types of tools and materials

Keywords: Ground Improvement, Infrastructure, Foundation, Buildings, Soft Soils. 18  Volume 45

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available. This paper covers an overview of the more common ground improvement methods in use especially in India. Specific projects are described to illustrate the various techniques. This paper will also cover the case studies where ground improvement using vibro techniques were implemented to mitigate liquefaction-induced damages to major infrastructure projects.

2.

Ground Engineering Techniques

Deep foundations like bored cast-in-situ piles and driven piles have historically been the foundation choice to address foundation of the structures constructed in the weak soil deposits. Alternatively, the weaker deposits can be handled with the commonly available ground improvement techniques such as

Fig. 2:Principle of vibro compaction technique

be rearranged into a denser state by means of vibration. Vibration is achieved by means of powerful vibrator at deeper depths. A Schematic showing Vibro Compaction technique is shown in Fig. 2. 2.2 Vibro Replacement The stabilization of weak deposits by displacing the soil radially with the help of a depth vibrator, refilling the resulting space with granular material and compacting the same with the vibrator is referred to as Vibro Replacement. The resulting matrix of compacted soil and stone columns will improve load bearing capacity and settlement characteristics. A schematic showing the basic principle of the vibro replacement technique, presented in Fig. 3. Keeping

Fig. 1: Application ranges of the deep vibro techniques

vibro replacement (vibro stone columns) and vibro compaction. These techniques use vibratory energy to densify loose soils at depth by backfilling. Vibro technique is one of the most suitable ground improvement techniques offers the weaker deposits to get compaction, drainage and increase in shear resistance in addition to preventing liquefaction during a seismic event. Fig. 1 shows transition zone of soils tends to liquefiable and possible techniques of ground improvement with deep vibro compaction or replacement. In addition, other ground engineering solutions such as grouting techniques and retention systems also discussed in this paper. 2.1 Vibro Compaction In order to address low bearing capacity, large amount of settlements and also to mitigate liquefaction potential during earthquakes, vibro compaction is the best suited ground improvement technique especially for loose sand deposits. The basic principle behind the process is that particles of non-cohesive soils can The Bridge and Structural Engineer

Fig. 3: Installation of stone columns (bottom feed method)

the site conditions in view vibro stone columns can be installed either wet method (top feed, Fig. 5) or dry method (bottom feed, Fig. 4). Technically and functionally, vibro stone columns installed in both methods serve similar. This method is suitable for pure silts/clays or mixed deposits of silts/clays/sands as they cannot be improved by vibro compaction because of their inability to respond to vibration. Vibro replacement techniques are widely adopted in various infrastructure projects to address bearing capacity, settlements and also to mitigate liquefaction potential across India. Volume 45 Number 4 December 2015  19


Fig. 4: Installation of stone columns (top feed method)

soils, reinforces fine-grained soils and stabilizes subsurface voids or sinkholes, by the staged injection of low-slump. Typically an injection pipe is first advanced to the maximum treatment depth. The low mobility grout is then injected as the pipe is slowly extracted in lifts, creating a column of overlapping grout bulbs. The process is illustrated in Fig. 6. The expansion of the low mobility grout bulbs displaces surrounding soils. When performed in granular soil, compaction grouting increases the surrounding soils density, friction angle and stiffness. In all soils, the high modulus grout column reinforces the soils within the treatment zone. Compaction grouting has been successfully used in India to increase bearing capacity, decrease settlement and liquefaction potential for some of the infrastructure projects.

Fig. 5: Schematic showing vibro replacement technique

2.3 Grouting Techniques Grouting can be defined as controlled injection of binder material, usually in a fluid state, into soil or rock, where it stiffens to improve the physical characteristics of the ground for geotechnical engineering applications. This technique takes place without disturbing the state of soil mass, via injection of grout under high pressure (from 20 to 40 MPa). Grouting techniques can cut off groundwater, strengthen soft in situ soils for tunnelling support, remediate settlement of structures caused by soft ground tunnelling, underpin structures, densify granular soils for liquefaction mitigation, construct or repair retaining walls, construct access shafts, etc.

Fig. 6: Process of compaction grouting

2.4 Bored Cast In-Situ Piles Piles are the most common type of deep foundation, that safely transfer loads from the super structure to the competent strata through soft soil stratum by means of friction, end bearing or a combination of both. A bored pile is a non-displacement form of foundation

2.3.1 Permeation Grouting Permeation grouting is a process of injecting a grout suspension to permeate through the intergranular voids of granular stratum. It produces an integrated mass of soil and grout suspension to lower the permeability and to improve strength. It is typically used for dam grouting applications and for groundwater control on tunnelling projects. 2.3.2 Compaction Grouting Compaction grouting also known as low mobility grouting that displaces and densifies loose granular 20  Volume 45

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Fig. 7: Installation of bored cast in-situ piles

that is built in situ and provides load bearing solutions suited to a wide range of ground conditions and applications. A hole is formed in the soil by boring and then filled with reinforcement and concrete. The sides of the hole can be unsupported or supported either permanent casing or temporary casing and The Bridge and Structural Engineer


drilling mud. Concreting is usually done using tremie pipes. The Schematic diagram of installation of Bored cast-in-situ pile is illustrated in Fig. 7. 2.5 Retention Systems Deep excavation or excavation pits is one of the challenging geotechnical problem in Urban environment. Several types of retention systems are constructed to support the earth where deep excavations are made. The design of the retention system considers movement control, stability during construction and wall drainage. In general, a retaining wall system (such as secant pile wall/ contiguous bored pile wall / soldier pile wall/ diaphragm wall/sheet pile wall) is combined with inclined ground anchors at different levels (depending on the depth of excavation) to support the intended excavations.

3.

Choice of technique

As can be seen from the above, several techniques are available for ground engineering and the choice of the appropriate one is important. The following sections describe some factors to be considered. 3.1 Suitability of the method Some methods lend themselves naturally to certain soils. Vibro compaction of reclaimed sand fills or alluvium deposits in North India is a good example. In reclaimed fills, the sands are relatively clean, and therefore the method is very fast and economical, even to large depths. 3.2 Technical compliance This is usually verified by design calculations to check for sufficient bearing capacity, factor of safety against slope failure or that the magnitude of settlements (total and differential) etc. are within limits. Some structures such as earth embankments and storage tanks are able to tolerate settlements in the order of decimetres during the construction stage. Therefore, ‘soft’ techniques relying on consolidation are often suitable. Other structures such as industrial plants require solutions which do not allow settlements of more than a few centimetres. In such cases ‘hard’ solutions such as densification of sands or some form of preload over the improved ground (to preclude long-term settlements) are required. The Bridge and Structural Engineer

3.3 Availability of quality assurance/quality control methods The availability of methods to ensure that quality is ensured during and after construction is important. Pre and post improvement testing by penetration methods (e.g. CPT), sampling etc. are essential. Post improvement load testing is possible for certain techniques (e.g. stone columns). In addition, real time monitoring during the improvement process using automated data loggers to inform the operator of what is happening below ground is very helpful to ensure quality. The data loggers can also be used to provide a printout of the construction process, a so-called birth certificate of the improvement point for daily review by the engineer. 3.4 Availability of materials Ground engineering methods usually required a range of materials, some natural (e.g. stone) and some manufactured (e.g. cement, geotextiles). The availability of these materials will influence the choice of technique. For example, several soft soil deposits are exists in coastal regions with nearby hilly terrain. That means stone is easily available and has led to extensive use of stone columns to treat the soft soils nearby coasts. 3.5 Time Methods which require long consolidation periods will obviously not be suitable for fast track projects. Installation/construction time is also important. Nowadays, however, modern high-production machinery allows a significant reduction in construction time. For example, the use of twin configurations in vibro methods has significantly increased production rates. 3.6 Cost Assuming that the solution satisfies technical requirements, cost often becomes the deciding factor. Methods which use less or cheaper added material are of course cheaper. However, if the cost of time or the risk of non-performance are added, then other apparently expensive solutions become economical. 3.7 Convenience Solutions which do not require other additional measures such as the placement of a large preload, or Volume 45 Number 4 December 2015  21


excavations (as in excavate and replace methods) are more convenient and practical. 3.8 Protection of environment Methods which produce large quantities of spoil are of course not desirable. In situ treatment methods which do not remove the soil or discharge excess cement/binder are preferred. For example stone columns installed by the ‘dry’ method only displace the in situ soil and often preferable method to protect environment. Similarly, in situ soil mixing would be preferable to jet grouting where possible. Another criterion whould be the influence of the method on sensitive structures nearby.

in order to meet design requirements. Based on the analysis of the post CPT results (Fig. 9), a relative density of more than 70% was achieved. Vibro compaction technique was proven cost effective in order to address low bearing capacity, improvement in settlements and also to mitigate liquefaction in the event of earthquakes.

4. Applications Variety of ground engineering techniques has been used successfully across India for various infrastructure projects for different structures with varying soil conditions satisfying performance requirements. The following sections describe briefly case histories of some of the sectors of the selected infrastructure projects.

Fig. 8: Vibro compaction using twin vibrators

4.1 Power Plant Projects 4.1.1 Vibro Compaction Ground conditions at one of the thermal power plant at Goindwal Sahib, Punjab has primarily sandy silt/ silty sand to about 1.5 m to 2 m depth, followed by fine sands with fines contents less than 10% to the final explored depth of about 30 m. The average SPT N value is 10 up to a depth 6 m and further SPT N values are in the range of 15 to 25 to a depth of about 15 m. Medium dense to dense sand layers were encountered beyond 15 m depth with SPT N values are generally > 25. The existing natural soils (fine sands) at the proposed site being loose the foundations were susceptible to liquefaction in an event of an earthquake. Hence, heavy foundations in the form of pile foundations were initially selected. After having critical review and keeping cost consciousness of the project, ground improvement using vibro compaction was selected to mitigate liquefaction and to enhance the bearing capacity. Vibro compaction for main works has been carried out to a depth of up to 10 m to satisfy design requirements (Fig. 8). Post penetration tests (CPT) were conducted after completion of the vibro compaction, as part of QA/QC procedures 22  Volume 45

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Fig. 9: Pre and Post penetration resistance values

4.1.2 Vibro Stone Columns A thermal power plant in north Chennai proposed to be built over the presence of alternate layers of soft clays and stiff clay which posed geotechnical challenges to the foundation design and construction of the pile foundations. The proposed plant consists of main plant structures (Boilers, TG, ESP, etc.) and balance of plant structures (Cooling Tower, CHP, AHP). All these heavy structures were intended to be built on pile foundations however, the rest of plant structures having low bearing pressures were proposed on ground improvement technique using vibro stone columns. Ground improvement in such environment provides a commercially economical solution. It was The Bridge and Structural Engineer


proposed to adopt installation of stone columns using wet top feed method to achieve required bearing capacity and to address post construction settlements. This application was adopted successfully for the lightly loaded structures such as Induced Cooling Towers (see Fig. 10), workshop buildings and other balance of plant structures.

adopting international standard of practices using latest equipment ensures the required process for installation of deep bored piles. Operational excellence with best practices delivers the high quality whilst ensuring peak productivity by achieving safety goal of zero accidents is possible.

Fig. 10: Induced cooling towers on vibro stone columns

Fig.11: Bored piling works in progress

4.2 LNG Storage Tanks

4.2.2 Vibro compaction and vibro stone columns

In an increasingly complex and competitive Indian foundation industry, creating real value for the client is essential. “Value� is not only providing the lowest price for the project, but also the best solution for the problem and reliable, on-schedule delivery. At two LNG tank projects in Gujarat State, where the foundations were delivered, in two different ways.

In the other location within the state of Gujarat, two 90 m diameter LNG storage tanks and regasification facilities are proposed and the project site is located in seismic Zone 5. Therefore liquefaction had to be accounted for in the design in addition to the low bearing capacity and control of settlements. The traditional piling solution was converted into ground improvement techniques using vibro stone column and vibro compaction (Fig. 12 and Fig. 13). The

4.2.1 Bored cast in-situ pile foundation Over two phases in the past, four LNG storage tanks and regasification facilities in Gujarat, the plan was to add two more storage tanks, each 90 m in diameter. The design called for a total of 1,272 bored piles, 1,000 mm in diameter going to a depth of 37 m below the existing ground level. In addition, 2.7 m long reinforced concrete columns were to be constructed on top of each pile foundation. The client proposed a total duration of 8 months to complete piles including structural columns, 6 initial and routine pile load tests. Crucially, the 8-month contract period included the seasonal monsoon and ground preparation works, which tends to severely disrupt steel and concrete supply. To deliver the project on-time, extensive project planning was done by deploying a total equipment fleet of 60 nos. including an onsite concrete batching plant with sufficient storage (see set up in Fig. 11). The project was successfully delivered within the scheduled time. Execution of deep bored piles requires state-of-art process by The Bridge and Structural Engineer

Fig. 12: Typical cross section of tank over stone columns

vibro stone column solution not only dealt with the problem of liquefaction, but also comfortably met the settlement requirements under static loads. The alternative foundation solution successfully demonstrated significant savings in time and cost Volume 45 Number 4 December 2015  23


compared to bored pile foundations. An extensive soil investigation program was carried out post construction, and a field trial was carried out to optimise installation parameters and confirm the geotechnical parameters of the improved ground. The ground improvement and tank pad works were completed on time, and superstructure works are underway.

pile foundations leads to high construction cost and long construction time. Keeping these challenges in view, alternative foundation techniques using ground improvement by installation of vibro stone columns (Figure 15) were adopted in majority of the cases and successfully demonstrated across India. One of the case study of Paradeep oil tank is discussed. The subsoil conditions at the project site comprise of recently reclaimed material for a depth of 3 m followed by alternate layers of loose to medium dense fine silty sand. The purpose of the ground improvement was to limit the settlement of the tanks during hydrostatic tests and also to achieve the required bearing capacity. The treatment area extended 6 m beyond the tank footprint to ensure the edge stability of the tank. Ground improvement technique using vibro stone columns were used across India successfully (see completed oil storage facility in Fig. 14).

Fig. 13: Tank erection in progress

4.3 Oil Storage Tanks 4.3.1 Vibro Stone Columns Large numbers of oil storage tanks were constructed by HPCL, IOCL, MRPL, Shell across India. Majority of these storage tanks are situated in coastal regions where weak soil deposits pose geotechnical challenges in terms of bearing capacity, settlements and some cases liquefaction mitigation. A common solution to address these challenges is through deep foundations in terms of bored pile foundations. Construction of pile

Fig.15: Installation of vibro stone columns in progress

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Fig. 14: Storage tanks resting on vibro stone columns

4.4 Packaged Solutions 4.4.1 Vibro Stone Columns Unavailability of suitable construction site for expansion of existing power plants demands the use of fly ash deposit/pond which is the one of the best options. Ash ponds in general are not consistent with the depth and density characteristics. The traditional methods of foundation design in such situations may result in commercially costlier solution. Ground improvement in such case provides a technocommercially feasible solution. One of the power plant built in UP is such classic example. The depth of ash varies across the site and ranges between 3 m and 13 m and is loose to medium dense in condition. The site falls under seismic Zone –III according to the IS 1893 (Part 1):1982, making it susceptible to liquefaction in an event of an earthquake. An extensive research has been done to address geotechnical challenges and The Bridge and Structural Engineer


stone columns also helped in enhancing the lateral capacity of deep pile foundations. Similar application was adopted for other structures such as Switchyard and for the Water Treatment Plant structures. 4.4.2 Bored cast in-situ piles and vibro stone columns

Fig. 16: Lateral pile load test in progress

established that ground improvement using stone columns (dry bottom feed method) need to be adopted for not only to mitigate the liquefaction potential but also to enhance the bearing capacity of the hydraulically deposited fly ash deposits. The stone columns are also installed to enhance the lateral capacity of bored cast-in-situ piles.

Fig. 17: Stone columns surrounding the piles

The density characteristics of fly ash vary across the site as a result the net safe bearing capacity for open foundations is less than the desired value of 10T/ m2. It is proposed to install stone columns to at least 0.5 m into the underlying stiff clayey silt / silty clay layer to achieve the desired bearing capacity for open foundations. The existence of loose fly ash deposits resulted in less than the desired lateral capacity of bored cast-in-situ piles. Stone columns were installed around the bored cast-in-situ piles (Fig. 17) to enhance the density characteristics of the fly ash surrounding the piles there by improving the lateral capacity to 7T (working load, see Fig. 16). It was proposed to adopt dry bottom feed method for installation of stone columns to achieve above applications. Further, the The Bridge and Structural Engineer

A Chemical plant consisting of Sulphate of Potash, Bromine and Cogen plants, and other ancillary structures was proposed at northern part of Gujarat State. The subsoil profile consists of top 5 m soft to firm silty clay followed by silty sand to a depth of 20 m below EGL. A packaged solution was adopted for this project which involved bored cast in-situ piling to address settlement sensitive structures and dry vibro stone columns for improvement of bearing capacity and to control settlements of lightly loaded structures having bearing pressures less than 10 T/m2. Packaged solutions in combination of deep foundations and ground improvement using vibro stone columns (dry bottom feed method) provide comfort to the designers in selection of foundation system based on structural requirements resulting in savings in cost and time.

Fig. 18: Installation of bored piles and vibro stone columns at site

Initial and routine pile load tests to ensure design load carrying capacities (compression, lateral and uplift) and number of plate load tests were conducted to satisfy the performance requirements. This success of foundation solutions demonstrated early commissioning of the project. 4.5 Residential Buildings 4.5.1 Vibro Stone Columns (low rise buildings) Deep foundations driven piles have choice for major constructed in the

like bored cast-in-situ piles and historically been the foundation buildings and other structures weak soil deposits. Construction

Volume 45 Number 4 December 2015  25


of pile foundations is becoming a challenge due to their high cost, large construction time and also due to severe environmental issues (noise pollution, ground vibrations and carbon footprint). Since the proposed building complex is located within the close proximity of a well-developed residential locality, severe resistance by the neighbourhood to the pile driving activities called for a rethinking on the alternative foundation system. According to the soil investigations conducted at the proposed project site, the top 6 m to 8 m of soil profile comprises of silty clay / sandy clay having highly varying consistency.

lower than the predicted settlement, which proved the efficiency of the improved ground. Application of vibro replacement proved to be an effective ground improvement solution which was completed within 6 weeks (as against 6 months to that of pile foundations). The project is delivered (Fig. 19) to the end users ahead of time as a result of alternative foundation solution marking a milestone in ground modification. The savings in time is key to success of ground improvement benefitting the entire cycle involving End Users, Suppliers, Bankers and Developer.

Under the present scenario, after giving due consideration to the nature of the ground being improved and the type and sensitivity of the structures being built, ground improvement using vibro stone columns (dry bottom feed method) was selected as alternative foundation system. Keeping the importance of the post construction performance of the structure, real time settlements were monitored on the raft foundation for duration of more than 2 years (Fig. 20). The measured settlements are substantially

4.5.2 Vibro Stone Columns (high rise buildings)

Fig. 19: Completed structure resting on vibro stone columns

Fig. 20: Results of monitored settlements

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A residential complex G+14 floor of about 13 towers was proposed in an area of 12 acres in NCR region. Appropriate foundation system was to be adopted to support structural column load of approximately 150 T transmitted to soil which resulted in large settlements. In addition, it was required to address liquefaction mitigation as the project site falls in seismic zone IV according to IS 1893. Ground improvement using vibro stone columns (dry bottom feed method) in such case provides a techno-commercially feasible solution to address bearing capacity, reduction in settlements and to mitigate liquefaction during earthquakes. Routine Group Column Load test (Fig. 21) was conducted to ascertain the effectiveness of design and performance of the ground improvement works and observed that the results were within the acceptable limits. Further, keeping the importance of the post construction performance of the structure, real time settlements are being monitored at critical locations of the building structure to record actual

Fig. 21: Group column load test results

The Bridge and Structural Engineer


settlements during and post construction of the structure. The measured settlements are substantially lower than the predicted settlement, which proved the efficiency of the ground improvement. Application of vibro replacement proved to be an effective ground improvement solution. (see completed structure in Fig. 22).

Fig. 23: Work in progress and view of load test set up

Fig. 22: Completed structure resting over stone columns

4.5.3 Retention Systems A multi storeyed residential building (3B+G+19F) in Chennai having structure foot print area of 1500 m2 to be constructed on large diameter pile foundations and also to support the proposed basement excavation. The subsoil conditions within the project site consist of 4 m thick silty clay followed by clayey sand till 20 m below EGL. Ground water table was encountered at about 8 m below ground level during investigation. In order to support heavy structural loads, large diameter deep foundations using bored cast in-situ piles were designed and installed. Further, complete retention system package includes installation of Contiguous Bored Piles (CBP) in combination of vertical grout columns behind CBP wall, installation of inclined ground anchors, capping beam was economically The Bridge and Structural Engineer

Fig. 24: Typical cross section of retention system

designed and built. (see typical section in Fig. 24). Routine compression load tests were conducted on the executed working piles to assess theoretical load carrying capacity and to ensure workmanship of the installed piles. All ground anchors were tested prior to locking to the required locking loads. A complete foundation package with specialized foundation contractor with the state of the art construction techniques demonstrated the success of the project. Typical site set up can be seen in Fig. 23. 4.6 Metro Rail Projects 4.6.1 Compaction Grouting As a part of expansion of Metro network, New Austrian Tunnelling Method (NATM) was adopted to construct the proposed tunnels in Delhi Metro. The presence of loose filled up sandy soils over a stretch Volume 45 Number 4 December 2015  27


of 100 m near one of the station posed problems with effective soil arching which was required for NATM construction. Compaction grouting was adopted to enhance the densities of loose sandy soils to form effective arching. In general, site execution constituted of drilling, installation of stinger rods & pumping the grout mix. A truck mounted hydraulic drill rig was used to drill a nominal diameter hole of 90 mm to a depth of about 8 m through the overburden soils. After completion of drilling process, the grout mix was pumped through the stinger rods, to form a bulb like element in the loose soils, in stages from bottom to the top of the working platform. Compaction grouting was successfully executed to facilitate the soil arching effect for NATM construction. Field trials were carried out to establish suitable grid pattern prior to commencement of main works. Pre and Post treatment tests were performed to assess the performance of compaction grouting works which are

6,000 inclined soil anchors were installed for number of underground metro stations across the city. The design load of each anchor varied from 60T to 80T, to support the deep excavation and all anchors were tested prior to the required locking loads. Installed anchors can be seen in Fig. 26.

Fig. 26: Installed ground anchors

4.7 Hydro Projects 4.7.1 Permeation Grouting

Fig. 25: Layout of compaction grouting grid

are proved satisfactory for the design requirements. Figure 25 illustrates layout of the compaction grouting grid implemented at site. 4.6.2 Anchoring systems As part of the development of the underground package of a metro rail construction in Bangalore, excavations to a depth up to 20 m to 25 m were required to build underground metro rail station. The proposed retention system for the deep excavations was a combination of soldier pile walls with timber lagging and inclined ground anchors at different levels. Secant piling system was adopted at some of the other metro stations. The anchoring project site in the present case consists of 2 to 4 m of moorum fill followed by sandy silt, with bedrock (weathered rock) as shallow as 5m and as deep as 13 m. About 28  Volume 45

Number 4 December 2015

Construction of a dam was proposed to generate 160 MW hydro power in West Bengal. In this connection, a cofferdam needs to be built to divert the river and to allow 20 m deep excavation in the riverbed for construction of dam foundation. The subsoil profile of the riverbed comprise of highly permeable mix of sand, gravels, boulders followed by bedrock. The depth of bedrock varied between 5 m to 20 m from the river bed. Seepage control was the major geotechnical challenge after excavation for building the dam foundation. In such conditions, installation of grout curtain would be the ideal choice to reduce the permeability up to 10-6 m/s and to allow construction of dam foundation in a relatively dry condition. Permeation grouting technique was chosen to form the grout curtain in the river bed along the alignment

Fig. 27: Installation of permeation grouting in progress

The Bridge and Structural Engineer


of the coffer dam. The coffer dam with permeation grouting supported a maximum excavation depth of 20 m with a maximum water head difference of 19 m. Post construction in-situ permeability tests i.e. pumpin test (above water table) and pump out test (below water table) were conducted at regular intervals along the cofferdam. The average permeability achieved was in the range between 10-6 and 10-7 m/s. Construction arrangements can be seen in Fig. 27. 4.8 Air Ports 4.8.1 Vibro Stone Columns One of the first Greenfield airports in South India intended to expand its operations. The airport has a single runway with a supporting taxiway and functioning with the existing apron of 807 m x 125 m. As a part of expansion, about 41,000 m2 of additional apron needs to be developed. Based on the available soil information, it is understood that top 10 to 11 m soil comprising of silty clay (N~4 to 18) followed by clayey sand or sandy clay (N~8 to 25) up to 15 m depth. Weathered rock is encountered from 15 m to 18 m depth below the existing ground level. Based on the in-situ soil conditions and given loading conditions, ground improvement technique using vibro stone columns (dry bottom feed method) has been chosen as one of the foundation alternatives to meet the required bearing capacity and to satisfy performance requirements. Ground improvement using vibro stone columns ensured enhancement of the bearing capacity of apron foundation and satisfied the performance criteria. Besides improving the shear strength and compressibility parameters of the in-situ soil, the technique also provides effective drainage paths to ensure rapid consolidation. This technique is proved to be cost effective solution for treating soft soils. Site view can be seen Fig. 28.

Fig.28: Installation of vibro stone columns in progress

The Bridge and Structural Engineer

4.9 Ports and Harbours 4.9.1 Vibro Stone Columns An International Container Transhipment Terminal was proposed at Vallarpadam, Kochi. The development consists of a 600 m long quay wall and a backup yard for container stack to be constructed. According to the soil conditions, soft to stiff marine clay soils were encountered till the top 20 m to 25 m from the existing ground level. In such subsoil environment, a natural dredge slope of 1 in 8 to 1 in 10 would normally be formed when dredging is completed in front of the piled jetty. The natural

Fig. 29: Installation of vibro stone columns in progress

dredge slope in such soils would extend beyond the jetty which would not be comfortable for utilizing as a container stackyard. In order to restrict the dredge slope to 1V:3H, the weak subsoil need to be improved. Ground improvement using vibro stone columns were chosen to improve shear strength of the slope portion and also to improve bearing capacity and to limit settlements in the container stackyard. Typical arrangement can be seen in Figure 30 and installation

Fig. 30: Cross-section stone columns and quay

Volume 45 Number 4 December 2015  29


of vibro stone columns can be seen in Figure 29. Routine load tests were performed to satisfy quality requirements of the project. The container stack yard is under operation.

3. KLAUS KIRSCH AND SONDERMANN W (2003), “Ground improvement, Geotechnical Engineering Handbook, vol 2: Procedures”, Ernst & Sohn, Berlin, Germany, pp. 1 - 56.

5. Conclusions

4.

RAJU VR (2009), “Ground improvement – principles and applications in Asia”, Ground Improvement Technologies and Case Histories (Leung CF, Chu J and Shen RF)

5.

RAJU VR ET. AL. (2010). “Some Environmental Benefits of Dry Vibro Stone Columns in a Gas Based Power Plant Project”, Indian Geotechnical Conference, December 2010

Advanced ground engineering techniques which are practiced internationally have been successfully implemented and demonstrated with sound engineering practices in India. These techniques include ground improvement using deep vibro methods, grouting techniques, packaged foundation solutions, deep and large diameter bored piles and complete retention systems. Application of deep vibro techniques proved to be an effective ground improvement solution in varying soil conditions. In India, deep vibro techniques have been extensively used for the construction of wide range of infrastructure and building facilities as an alternative foundation system in place of deep foundations. In seismic zones with liquefiable soils, ground improvement technique provides technically sound and cost effective foundation solutions. Packaged foundation / complete retention system solutions offer speed in the overall construction schedule and enable the projects to be completed within stipulated duration. However, efficient and economic solutions to problems caused by soil conditions require a thorough evaluation of project conditions, project needs, specialist foundation contractor, method capabilities and thorough field testing program.

6. References 1.

PRIEBE, H.J (1995), “The Design of Vibro Replacement”, Ground Engineering, December 1995

2.

PRIEBE, H.J (1998), “Vibro Replacement to Prevent Earthquake Induced Liquefaction”, Ground Engineering, September 1998

30  Volume 45

Number 4 December 2015

6. KLAUS KIRSCH AND FABIAN KIRSCH (2010), “Ground improvement by deep vibratory methods”, Spon Press, London and New York 2010. 7.

RAJU VR (2011), “Ground Improvement Using Vibro Techniques in FlyAsh Deposits”, National Conference on Recent Advances in Ground Improvement Techniques, February 2425, 2011, CBRI Roorkee, India

8. RAJU VR AND MADAN KUMAR. (2012), “Ground improvement solutions to mitigate liquefaction: Case Studies”, Proceedings of Indian Geotechnical Conference Key Note lecture 6. 9.

HARI KRISHNA ET. AL (2012), “Application of ground anchors to support deep excavation and compaction grouting for NATM tunnel construction for Delhi Metro Rail Corporation (DMRC)” ICUST, New Delhi, India

10. ANIRUDHAN, MADAN KUMAR, HARI KRISHNA (2015), “Optimal foundations in soft ground: an innovative approach for economizing cost and time” International Conference on Soft Ground Engineering (ICSGE2015), December 2015, Singapore

The Bridge and Structural Engineer


EVALUATION OF LIQUEFACTION POTENTIAL FOR DESIGN OF DEEP FOUNDATION

Mahesh TANDON Managing Director Tandon Consultants Pvt. Ltd. New Delhi, India tandon@tcpl.com

Navneet GUPTA Principal Consultant Tandon Consultants Pvt. Ltd. New Delhi, India navneet.gupta@tcpl.com

Harsimran Singh GREWAL Senior Design Engineer Tandon Consultants Pvt. Ltd. New Delhi, India hsg@tcpl.com

Mahesh Tandon, born 1941 received his Bachelor degree in Civil Engineering IIT Roorkee & Masters from University of Hawai, USA. He is former Distinguished Visiting Professor at the IITs at Kanpur, Roorkee and Gandhinagar appointed by the Indian National Academy of Engineering (INAE) and the ALL India Council for Technical Education (AICTE). He was recently conferred Honorary Fellowship by the Indian Concrete Institute.

Navneet Gupta, received his B.Tech (Civil) degree from V.R.C.E., Nagpur (1992) & M.Tech (1996) from University of Roorkee, now IIT, Roorkee.

Harsimran Singh Grewal, received his B.Tech (Civil) degree from Indian Institute of Technology, Roorkee (2010). He has been working as a Senior Design Engineer in Tandon Consultants Pvt. Ltd., New Delhi and his areas of interest include designing Metro Viaduct bridges, steel structures and Underground Metro stations.

Working as a Principal Consultant in Tandon Consultants Pvt. Ltd, New Delhi with area of interest in designing and planning of long span prestressed and RCC bridges, steel bridges, vehicular underpasses, underground Metro structures, special structures and providing specialist advice for these.

Summary Evaluation of soil liquefaction potential is an important aspect of designing bridges in a seismic event. Soil liquefaction can induce large deformations in and around foundation soil during an earthquake. Depth of liquefaction is required by the structural designer to arrive at vertical loads, shear forces and bending moments in the deep foundation system. Load combinations and seismic parameters used for structural design are an important aspect to estimate liquefaction as the same cannot be assessed in isolation of the final design. Therefore, it is essential that structural designer must The Bridge and Structural Engineer

understand the phenomenon of liquefaction and different parameters assumed to evaluate liquefaction potential. In absence of appropriate guidelines or codal provisions, many structural engineers have to go through a major challenge while assessing liquefaction and, if needed, finding a remedy for it. This paper highlights the step by step procedure to evaluate liquefaction and final design forces in the deep foundation. Keywords: Liquefaction, earthquake, SPT, Deep foundation, RDSO Guidelines for seismic design of railway Bridges Volume 45 Number 4 December 2015  31


1.

What is Liquefaction?

Loose saturated sand is most liable to get liquefied under moderate to strong earthquake. The pore pressure increases to an extent that the effective stresses are virtually reduced to zero, leading the soil to lose all its strength and the ground to behave similar to a liquid. Liquefaction is a state in saturated granular (cohesion less) soil wherein the effective shear strength is reduced to negligible value during a seismic event for structural engineering purposes.

unless appropriate methods of soil compaction or stabilization are adopted.  Alternately, the foundation should be taken deeper below the liquefiable layers. 

Reference should be made to specialist literature for analysis of liquefaction potential.

IS 1893 : Part-1 2002 (Ref 2)  Submerged Loose Sand and classification SP may liquefy with

This has a direct effect on the foundation as under:

 Requirement of deeper/pile well foundations.

 Requirement of higher diameter or number of piles.  Requirement of more reinforcement.

 Foundation in such strata should be avoided unless appropriate methods of soil compaction or stabilization are adopted.

Fig. 1 depicts the modelling of pile foundations during seismic analysis for liquefiable soil.

 Alternately, the foundation should be taken deeper below the liquefiable layers.

N < 15 in Seismic zone III, IV V

N < 10 in Seismic zone II

 Specialist literature may be referred to determine liquefaction potential of site. RDSO Guidelines for seismic design of railway Bridges (Ref.3) Liquefaction analysis procedure is given in considerably more detail as compared to the other Indian Codes with regards to the parameters to be used and the overall procedure. However, there is no clarity with regards to the following parameters: PGA (Peak Ground Acceleration) to be considered: MCE or DBE.

Earthquake Magnitude.

Load combinations to be used for (MCE / DBE )

Vertical geotechnical capacity of pile under liquefaction condition

Structural design of pile

Fig. 1: 3D-Model of Pile & Pile cap during Seismic Analysis 

2.

Present Codal Provisions

The recently updated Indian codes have in relatively more detail, though not enough, the criteria to estimate the liquefaction potential of soil. The same has been reproduced as shown below: IRC : 6-2014 (Ref 1)  In loose and poorly graded sand with little to no fines, the earthquake may cause liquefaction.  Foundation in such strata should be avoided 32  Volume 45

Number 4 December 2015

3. Suggested Procedure for Liquefaction Calculations The procedure as given in RDSO can be used per se but not in isolation to the overall structural design as explained below. Liquefaction depth calculations get a bit restricted in their use to the DBE case only due to absence of load The Bridge and Structural Engineer


combinations in MCE case. Liquefaction, though an objective process during a specific seismic event for a specific ground water table, cannot be evaluated for design in an MCE condition with the worst water table due to the absence of well-defined load combinations for the same. Therefore, this paper sheds some light on the evaluation procedure to be adopted with the current codal provisions. Load Combinations Given in RDSO The liquefaction depth has to be incorporated for all load combinations as given in RDSO and reproduced below:

= Zone factor = Z = 0.24 (corresponding to zone IV for MCE) However, as per the RDSO guidelines, no load combinations have been given for the MCE case. For the DBE case, value of amax/g reduces to 0.12 for Zone IV. The RDSO guidelines give the value of amax/g as amax/g = (Sa/g) x Z x I Sa/g = 1, For zero period acceleration (Refer Fig. 2)

(A) Ultimate limit state design 1) 1.25DL + 1.5 DL(S) +1.5EQ + 1.4 PS+ 1.7 EP 2)

1.25DL + 1.5DL(S) + 0.5(LL + LL (F)) + 1.2EQ + 1.7 EP + 1.4PS + 1. 4HY + 1.4BO

3)

0.9DL + 0.8DL(S) + 1.5EQ + 1.4 PS + 1.7 EP

(B) Serviceability Limit State 1)

1.0 DL+1.2 DL(S) +1.0 EQ + 1.0 EP + 1.0PS + 1.0HY+ 1.0BO

2) 1.0 DL + 1.2 DL(S) + 0.5(LL+LL(F)) + 1.0EQ + 1.0 EP + 1.0PS + 1.0HY + 1.0 BY

4.

Fig. 2: Response Spectra

I

= Importance Factor = 1.5 for most projects as per Indian Codes for structural design.

Evaluation of Liquefaction Potential

Estimation of two parameters is required to evaluate liquefaction potential 1) CSR or Critical (or Cyclic) Stress Ratio – Demand on soil layers during the seismic event and, 2) CRR or Critical (or Cyclic) Resistance Ratio – Capacity of the soil to resist liquefaction.

Hence, the value of amax/g for Zone IV may be taken as 0.12x1.5 i.e. 0.18 (as per RDSO) for the purpose of design in DBE case. In the author’s opinion, importance factor should not be considered for liquefaction depth calculations.

(Total vertical σv/σv’ = vertical Stress)

Stress/Effective

A Factor of Safety, FOS (= CRR/CSR) of greater than 1 is usually associated with non-liquefiable soil. FOS >1 generally implies small earthquake related deformations.

which may vary from approximately 2 to 1 depending upon Ground Water Table considered. It would be equal to 2 for Ground Water Table at Ground Level and equal to 1 if Ground Water Table is considered at the level where liquefaction is to be determined.

4.1 Evaluation of Cyclic Stress Ratio (CSR)

Effective

CSR

rd

= 0.65 (amax /g) ( σv /σv’) rd = Stress Reduction Factor which depends on depth below ground level

amax/g = Ratio of Peak Horizontal Ground Acceleration to acceleration due to gravity The Bridge and Structural Engineer

Stress

=

Total Stress – Pore Water Pressure

From Fig. 3 it can be seen that stress reduction coefficient is inversely proportional to depth below ground level.

rd

= 1-0.00765z for z < 9.15 m and

rd

= 1.174 - 0.0267z for 9.15 < z < 23 m Volume 45 Number 4 December 2015  33


4.2.1. Evaluation of Standardized SPT (N60) and CRR7.5 (N60) is the standard penetration test value for hammer of efficiency 60%. N60

= N C60

Where,

N

= Observed field SPT value

C60 = CHTCHwCSSCRL CBD CHT

= Energy ratio correction

CHw = Hammer weight correction Fig. 3: Stress reduction Coefficient with depth

4.2 Evaluation of Cyclic Resistance Ratio (CRR) CRR = CRR7.5 kM kσ kα

CSS

= Sampling method correction

CRL

= Rod length correction

CBD

= Bore Hole diameter correction

Where kM

= Earthquake magnitude correction (See 4.2.2)

Normalized standardized SPT (N1)60 is the normalized N60 to effective overburden pressure at ground level i.e. 98 kPa.

= Overburden correction

(N1)60 = CN N60= N CNC60 = N CNCHTCHwCSSCRL CBD

= Sloping ground correction

Where CN =

The overburden and sloping ground corrections usually apply mostly to the liquefaction hazard of embankment dams and other large structures and are usually 1 for normal bridge structures. CRR7.5 is the CRR calculated for an earthquake magnitude of 7.5. The same is a characteristic of the composition of the soil and can be obtained from either of the following:

(Pa/σv’)1/2 = 9.79 (1/σv’)1/2

Pa is the atmospheric pressure

A total of six numbers of corrections are applied on observed N value to arrive at (N1)60. There is no provision in the Indian codes for these corrections based on SPT equipment and methods followed in the country. Fig. 4 & Fig. 5 depict that N value depends on the type of equipment and method followed for determination of N.

 Testing of retrieved undisturbed soil sample in lab. However, the specimens of granular soil retrieved with typical techniques are too disturbed to give meaningful results.  Field tests. To avoid difficulty with sampling and testing, the commonly used methods for evaluation of liquefaction potential include:

Standard penetration test (SPT): Generally preferred because of more extensive data base. Cone penetration test (CPT) Shear wave velocity (Vs) Becker penetration test (BPT)

34  Volume 45

Number 4 December 2015

Fig. 4: Type of Hammers

The Bridge and Structural Engineer


However, all the above specifications do not match completely with the standard equipment details as per the IS codes for the SPT since the above specifications have been reproduced from ASTM standards. In case of non-standard equipment, the corrections need to be applied as per the RDSO Guidelines as per the following: Corrections for non-standard Equipment: Element Nonstandard Hammer Type

Fig. 5: Method of taking SPT

Corrections to SPT (RDSO Guidelines) (Ref. 3) The details of the standard equipment to be used for SPT for the purpose of liquefaction have been given in RDSO Guidelines as shown below: Element Sampler

Drill Rods Hammer

Rope Borehole Drill Bit Blow Count Rate Penetration Resistant Count

Standard Specification Standard split-spoon sampler with: (a) Outside diameter = 51 mm, and Inside Diameter = 35 mm (constant – i.e., no room for liners in the barrel) A or AW-type for depths less than 15.2 m; N- or NW-type for greater depths Standard (safety) hammer: (a) drop hammer (b) weight = 65 kg; (c) drop = 750 mm (d) delivers 60% of the theoretical potential energy Two wraps of rope around the pulley 100 to 130 mm diameter borehole Upward deflection of drilling mud (tricone or baffled drag bit) 30 to 40 blows per minute Measured over range of 150 to 450 mm of penetration into the ground

If the test equipment meets the above specifications, N = N60 and only a correction for overburden is needed. The Bridge and Structural Engineer

(DH= doughnut hammer; ER = energy ratio) Nonstandard Hammer Weight or Height of fall (H = height of fall in mm; W = hammer weight in kg) Nonstandard Sampler Setup (standard samples with room for liners, but used without liners Nonstandard Sampler Setup (standard samples with room for liners, but liners are used) Short Rod Length Nonstandard Borehole Diameter

Standard Specification CHT =0.75 for DH with rope and pulley CHT =1.33 for DH with trip/auto and ER = 80 CHW = (H x W) / (63.4 x 762)

CSS =1.10 for loose sand CSS =1.20 for dense sand CSS =0.90 for loose sand CSS =0.80 for dense sand CRL =0.75 for rod length 0-3 m CBD =1.05 for 150 mm borehole diameter CBD =1.15 for 200 mm borehole diameter

Corrections to SPT Based on International Practice Table 2: Corrections to SPT (Modified from Skempton 1986) as Listed by Robertson and Wride (1998) Factor Overburden pressure Overburden pressure Energy ratio Energy ratio Energy ratio Borehole diameter Borehole diameter Borehole diameter Rod length Rod length Rod length Rod length Rod length Sampling method Sampling method

Equipment variable –

Term CN

Correction (Pa/σ’v0)0.5

CN

CN ≤ 1.7

Donut hammer Safety hammer Automatic-trip Donut-type hammer 65–115 mm 150 mm 200 mm <3 m 3–4 m 4–6 m 6–10 m 10–30 m Standard sampler Sampler without liners

CE CE CE

0.5–1.0 0.7–1.2 0.8–1.3

CE CE CE CR CR CR CR CR CS Cs

1.0 1.05 1.15 0.75 0.8 0.85 0.95 1.0 1.0 1.1–1.3

Commonly used in country. The RDSO has suggested the correction of 0.75. This may be verified with proper test

Other corrections are comparable with RDSO

In Fig. 6, the graph between CRR7.5 and (N1)60 is shown for sand with different fines content (reproduced Volume 45 Number 4 December 2015  35


from the RDSO guidelines) but the same is very cumbersome to read. Alternately, the following equations can be used for clean sand (Ref 4).

The above equation can be easily used in the spreadsheets. Above equation is applicable for (N1)60 < 30. For higher values of (N1)60, granular soils are too dense to liquefy and hence assumed to be non-liquefiable. Influence of Fine Content (FC) on CRR7.5 In the original development Seed et al. noted an apparent increase of CRR with increase in fine content. Curves are presented in the RDSO Guidelines for percentage of fines of 5, 15 & 35% but the same are very cumbersome to read. Alternatively, following equations developed by I.M. Idriss with the assistance of R.B. Seed (Ref. 4) for correction of (N1)60 to an equivalent clean sand value i.e. (N1)60CS (N1)60CS

=

Fig. 6: Relationship between CRR7.5 & (N1) for sand for Mw 7.5 earthquake

correction is required for the same. From Fig. 7, it is seen that CRR7.5 is 1 corresponding to an earthquake magnitude of 7.5.

α + β (N1)60

where α and β = coefficients determined from the following relationships:

α = 0

for FC <= 5%

α = exp [1.76 – (190/FC2)] for 5% < FC < 35%

α = 5.0 for FC >= 35%

β = 1.0 for FC <= 5%

β = [0.99 + (FC1.5/1000)] for 5% < FC < 35%

β = 1.2 for FC >= 35%

where FC is the percentage fine content by weight passing through IS standard sieve of 75 micron. The equation can be used in the spreadsheet to calculate (N1)60CS for various % of fine content. From Fig. 6 it can be seen that CRR7.5 is very much influenced by fine content. CRR7.5 for (N1)60 = 20, varies from 0.2 to 0.4 for fine content of 5% to 35%. 4.2.2 Magnitude Correction The original formulae and graphs derived were valid for an earthquake magnitude of 7.5 and hence no 36  Volume 45

Number 4 December 2015

Fig. 7: Earthquake Magnitude Correction

Based on actual earthquake magnitude, Km is obtained from Fig. 7. The value of Km varies from 2 to 0.8 for earthquake magnitude of 6 to 8. It may be noted here that the magnitude, as used in this procedure, is a measure of the duration of strong earth shaking. It is recommended to use the Moment magnitude scale for the purpose of determining Km. Earthquake magnitude to be used for a particular site is not clear in any of the codes. This paper suggests to use the same from Fig. 8 or Fig. 9 which are slightly different from each other. From the graph it can be seen that earthquake magnitude is very subjective for a particular site. Therefore, the same should be defined by a technical expert in this field before proceeding with liquefaction calculations. The Bridge and Structural Engineer


Sample = Standard Use of Liner = No Bore Hole Diameter = 150 mm Correction for Bore Hole Diameter CSD = 1.05 for 150 mm = 1.15 for 200 mm Correction for Sampler Setup CSS = 1 for Standard Setup Without Liner CSS = 1.1 for loose sand = 1.2 for dense sand

2.40

0.26

CE or CHT

Cyclic Stress ratio (CSR)

5.40

CN

Effective overburden (σo) t/m2

0.95 12.00 0.98

Total overburden pressure (σo) t/m2

1.95

Stress reduction coefficient (rd)

Saturated demsity (t/m3)

11

Fine Content (%)

Observed SPT Value

SP-SM

= 0.9 for loose sand = 0.8 for dense sand = 0.75 for depth 0 - 3 m = 0.8 for depth 3 - 4 m = 0.85 for depth 4 - 6 m = 0.95 for depth 6 - 10 m = 1 for depth greater than 10 m

Submerged Density t/m3)

Type of Strata

3.00

2.00 0.75

21

2.00

1.00 11.00 0.94

14.27

6.77

0.23

1.22 0.75

9.00

ML-SM

45

2.00

1.00 52.00 0.93

17.27

8.27

0.23

1.10 0.75

10.50

ML-SM

53

2.00

1.00 52.00 0.89

20.77

9.77

0.22

1.01 0.75

12.00

ML-SM

59

2.01 52.00 0.85 23.27 11.27

0.21

0.94

0.75

13.50

CI

50

2.06

1.06 77.00 0.81

26.28

12.78

0.20

0.88 0.76

15.00

SM-ML

43

2.00

1.00 34.00 0.77

29.36

14.36

0.19

0.83 0.75

15.50

SM-ML

46

2.00

1.00 34.00 0.73

32.36

15.86

0.18

0.79 0.75

18.00

SM-ML

48

2.00

1.00 34.00 0.69

35.36

17.36

0.17

0.76 0.75

Conclusion

SP-SM

FOS

7.50

CRR

1.62 0.75 1.38 0.75

CRRM=7.5

0.25 0.24

(Nt)4OCs

3.83 5.27

β

8.33 11.27

α

0.96 12.00 0.97 1.00 11.00 0.95

SPT corrected (N1)03

1.96 2.00

CS or CSS

14 20

CR or CRL

SP-SM SP-SM

CS or CSD

4.50 6.00

CH or CHW

Fig. 8: Earthquake Hazard Map Developed by Building Material & Technology and Promotion Council of India

Depth beloe EGL m

With Liner CSS Correction for Rod Length CRL

0.986 1.05

0.75 1.00

12.81

1.55 1.03 14.77 0.16 0.20 0.77

Liquefiable

0.986 1.05

0.85 1.00

14.94

1.55 1.03 16.97 0.18 0.23 0.92

Liquefiable

0.986 1.05

0.95 1.00

20.33

1.21 1.03 22.08 0.24 0.30 1.27 Non Liquefiable

5. Worked Example Using Spreadsheet on actual field data

0.986 1.05

0.95 1.00

18.83

1.21 1.03 20.54 0.22 0.28 1.19 Non Liquefiable

0.986 1.05

0.95 1.00

36.51

5.00 1.20 48.82

NA

NA

>1

Non Liquefiable

Input Data List of constants used for this project Peak Horizontal Ground Acceleration (amax/g) = 0.18 Earthquake Magnitude (Mw) = 7.0 Magnitude Scale Factor KM = 1.25 Actual Depth of water tabel = 0 Design Depth of water table = 0 Stoped Stratigraphy coeff. Kα = 1 Effective Stress Coeff. Kα = 1 Hammer Energy Correction CHT = 0.75

0.986 1.05

1

1.00

41.65

5.00 1.20 54.97

NA

NA

>1

Non Liquefiable

0.986 1.05

1

1.00

43.16

5.00 1.20 56.80

NA

NA

>1

Non Liquefiable

Fig. 9: Earthquake Magnitude IS 1893

(From RDSO Fig G-1) m m

Input Data from GI Report Weight of Hammer

=

63.5 Kg

Height of Fall

=

750 mm

The Bridge and Structural Engineer

0.986 1.05

1

1.00

34.34

5.00 1.20 46.21

NA

NA

>1

Non Liquefiable

0.986 1.05

1

1.00

27.87

4.93 1.19 38.04

NA

NA

>1

Non Liquefiable

0.986 1.05

1

1.00

28.37

4.93 1.19 38.64

NA

NA

>1

Non Liquefiable

0.986 1.05

1

1.00

28.29

4.93 1.19 38.55

NA

NA

>1

Non Liquefiable

It is noticed that liquefaction calculations are very much dependent on the observed N. Therefore, reliability of same is very important.

Volume 45 Number 4 December 2015  37


The liquefaction calculations may vary from bore hole to bore hole which cannot be relied upon. This paper suggests to plot the observed “N” value and if required, carry out more bore holes to further supplement the N value data. Any odd value due to human error or small loose pocket should be ignored to arrive at N values to be used for evaluating the liquefaction potential of the particular stretch.

6. Conclusion Liquefaction calculations are simple but subjective. Therefore, various parameters such as  PGA to be considered  Magnitude of earthquake  Water table to be considered  Load combination to be used  Corrections to be applied on SPT Shall be clearly defined in consultation with an expert

38  Volume 45

Number 4 December 2015

to arrive at overall economy and safety of the project.

7. References 1.

IRC: 6-2014.

2.

IS: 1893 (Part-1)-2002.

3.

RDSO Guidelines for seismic design of railway Bridges.

4. YOUD, T.L., IDRISS, I.M., ANDRUS, R.D., ARANGO, I., CASTRO, G., CHTRISTIAN, J.T., DOBRY, R., FINN, W.D.L., HARDER, L.F., HYNES, M.E., ISHIHARA, K., KOESTER, J.P., LIAO, S.S.C., MARCUSON III, W.F., MARTIN, G.R., MITCHELL, J.K., MORIWAKI, Y., POWER, M.S., ROBERTSON, P.K., SEED, R.B., STOKOE II, K.H. 2001. “Liquefaction resistance of soils: Summary report from the 1996 NCEER and 1998 NCEER/NSF workshops on evaluation of liquefaction resistance of soils”, Journal of Geotechnical and Geoenvironmental Engineering, ASCE. 127(10): 817-833.

The Bridge and Structural Engineer


LIQUEFACTION POTENTIAL OF DELHI AND SOME MITIGATION OPTIONS Chandan GHOSH Professor & Head (Geohazards Division) National Institute of Disaster Management Ministry of Home Affairs, Govt. of India New Delhi-110002, India cghosh24@gmail.com

Summary National Capital Region-Delhi has been subjected to both local and distant earthquakes. Recent earthquakes in Nepal and Afghanistan have shaken our understanding of the microzonation studies conducted so far, with special mention about the varying impact of M6.9 Sikkim-Nepal border (2011) earthquake in the north Sikkim and adjoining areas even beyond 100km from the epicentre. Some studies such as checking of building vulnerability and retrofitting, strong motion recording, micro-earthquake survey; have been undertaken for Delhi. The liquefaction susceptibility within Delhi has been initially assessed based on Standard Penetration Test data from 491 soil investigation project reports containing 2500 borehole locations, which latter on carried out extensively for Delhi by National Centre for Seismology (2015). The liquefaction potential of Delhi has been evaluated using simplified procedures based on SPT results from the earlier data. The results have been classified into two groups according to the extent of liquefaction observed, namely, as to whether the site is liquefiable or non-liquefiable. This paper suggests possible option for identifying liquefaction induced damages vis-à-vis some of the mitigation measures for foundation improvement. Keywords: Liquefaction, Standard Penetration Test (SPT), Earthquake, Peak ground acceleration, foundation, fault, intensity scale, cyclic stress ratio, simple shear, shear wave velocity, magnitude scaling factor

1.

Introduction

National Capital Territory (NCT) of Delhi is The Bridge and Structural Engineer

Prof. Chandan Ghosh, a double doctorate degree holder, has been working in various field such as Seismic microzonation & Risk mapping, Reinforced earth, Application of geosynthetics for hill area development, bioengineering by Vetiver Grass for erosion and landslides control, documentry disaster resistant housing technologies, retrofitting and earthquake risk mitigation & management, etc. He has published more than 100 research papers in reputed journals and proceedings.

frequently affected with distant earthquakes occurring in the Himalaya. Several earthquakes that occurred in and around NCR have injured or claimed lives in Delhi, e.g., October 10, 1956 (M 6.7), August 27, 1960 (M 6.0), August 15, 1966 (M 5.8) but there has been no such local shock >M5 in the last 50 yrs that might bring significant impact on the non specific real estate developments taking place since thereafter. With burgeoning population surge in the megacity, uncontrolled infrastructure development and non-engineered construction practices the city built environment (Bilham et al 2003, Bilham and England 2001, Ambraseys and Bilham 2000) is becoming increasingly vulnerable to earthquakes and related risk. Some of the significant issues created by collapse of buildings (Ghosh, 2011), even without any earthquake shaking, have drawn attention of the city planners in terms of the importance of vulnerability studies. Delhi, being in zone IV, has experienced many earthquakes in the recent past and now detail microzonation report (NSC, 2015) identifies the likelihood of liquefaction hazards. Though, in this report liquefaction hazards is shown extensively without any consideration of the over ground built up facilities, the real risk due to earthquake in Delhi is incomprehensible. More so, buildings constructions are not regulated as per codal guidelines and constructions are hardly checked at site where there is wide gap between provisions given in the code since 1962 and actual implementation of the same. Even moderate earthquakes (M<7) occurring in the Himalaya have shaken many establishments in Delhi. Delhi’s geological set-up is vulnerable to large amplified shaking due to earthquakes. A Volume 45 Number 4 December 2015  39


detailed study of the sub-soil profile based on 2500 borehole records has been interpreted in terms of liquefaction susceptibility maps and major findings in the NCS (2015) report are discussed. Liquefaction induced ground failures, causes extensive structural and lifeline damage during an earthquake in urbanized areas around the world. Some examples of significant disaster caused by soil liquefaction due to earthquakes are 1964 Alaska and Nigata, 1989 Loma Prieta, 1994 Northridge, 1995 Kobe and 1999 Izmit. In India, during the Bhuj (2001, M 6.9) earthquake, several areas in Rann of Kutchh had experienced liquefaction, sand boiling and ground-cracking. Field-test-based methods (SPT, CPT) are likely the best suited to estimate liquefaction-induced ground deformations for low-to medium-risk projects. These methods are empirical and do not incorporate the extensive knowledge gained from laboratory studies of liquefaction. Generally, liquefaction-induced ground failures include flow slides, lateral spreads, ground settlements, ground oscillation, and sand boils. From the geotechnical engineering view-points, design issues concerning to liquefaction damages consist of three parts: (a) evaluation of liquefaction hazard, (b) evaluation of potential ground displacement, and (c) mitigating the hazard by designing to resist excessive ground displacement or strength loss, by reducing the potential for liquefaction, or by choosing an alternative site with less possible hazard. Although it is possible to identify areas that have the potential for liquefaction, its occurrence cannot be predicted any more accurately than a particular earthquake can be with a time, place, and degree of reliability assigned to it. Once these areas are defined in general terms, it is possible to conduct site investigations that provide very detailed information regarding liquefaction potential.

2. Geological setting of Delhi Delhi is located at the Northern end of Aravali Mountains and is mostly surrounded by IndoGangetic Alluvium. It is postulated that the Aravalis might have extended as far as upto the Himalaya, which makes Delhi susceptible to seismic events in the Himalaya (DST 2004, Valdiya 1976). An extension of the Aravali Hills enters the Delhi region from South, spreads out into a rocky tableland and traverses in a North-easterly direction 40â&#x20AC;&#x192; Volume 45

Number 4 December 2015

across the Delhi (Parvez et al, 2004, 2006). The rocks of Delhi have undergone multiple folding and metamorphism. The longitudinal ridge, trending NNE-SSW runs from the West of the capital city and terminates at the right bank of Yamuna in the north. The largest part of Delhi lies in the alluvium, but the small hills and ridges in and around Delhi consist of Alwar quartzites. The thickness of the alluvial deposits and the depth to the bedrock varies significantly throughout Delhi. The trends of the faults and major shear zones vary from NNE-SSW to ENE-WSW. The alluvial deposits belong to the Pleistocene period, which mainly consists of mostly inter-bedded lenticular and interfingering deposits of clay, silts and sand along with Kankar and of recent age, i.e.; newer alluvium (NCS, 2015). The updated geological profile of NCT, Delhi is shown in Fig. 1 and bed rock profile of Delhi is shown in Fig. 2.

3. Seismicity of Delhi The National Capital Region (NCR) is located between Mahendragarh-Dehradun fault in the West and Great Boundary fault in the East, at the junction of Aravalli Delhi fold belt and Indo-Gangetic foredeep. A few faults / lineaments of seismogenic nature have been inferred dissecting the fold belt (NCS 2015, Chauhan 1975, GSI 2000). Three earthquakes of Mâ&#x2030;Ľ 4 had their epicenter located in this region prior to the event of 26 November 2007. These were (i) Delhi earthquake of 15 July 1720 of epicentral intensity IX, in which some 1000 people perished, (ii) earthquake of 27 August 1960 of M 6.0 (VII) and (iii) earthquake of 28 July 1994 of M 4.0 (IV). The Bulandshahar earthquake of 10 October 1956 of M 6.5 and Moradabad earthquake of 15 August 1966 of M 5.6 had considerable effect in the NCR. Some of the damaging earthquakes originating in the Northwest Himalaya also affected the region (Rao, 2003, Rao and Satyam 2007). The Seismic Zone map of India, therefore, categorises NCR in Zone IV that is of High Hazard category (IS 1893 - 2002) with an expected basic peak ground acceleration of 0.24g. Iyenger and Ghosh (2004) prepared the seismo-tectonic map of Delhi. Based on the probabilistic seismic hazard assessment Iyenger and Ghosh (2004) estimated the maximum magnitude that can produce by each fault. They identified 13 faults from local sources and 7 faults from Himalayan source. The Bridge and Structural Engineer


thrust zone, just 200â&#x20AC;&#x201C;250 km north of Delhi, has been identified as a significant seismic gap in the Central Himalayas. Several tectonic features such as the Himalayan Main Boundary Thrust (MBT) and the Main Central Thrust (MCT), the Delhi-Hardwar ridge, the Delhi-Lahore ridge, the Aravalli-Delhi fold, the Sohna fault, the Mathura fault, the Rajasthan Great Boundary Fault and the Moradabad fault in addition to several other minor lineaments, dominate this region (GSI 2000, Rao 2003, Shukla et al 2007). Delhi is therefore quite vulnerable to Himalayan earthquakes and its burgeoning population and industrial works face increasing risk from seismic hazard.

Fig. 1: Geological map of national Capital Territory, Delhi (EREC, IMD 2005)

Fig. 3: Seismotectonic map of the region around Delhi showing all earthquake events and its epicentres (NSC, 2015)

Fig. 2 Updated bedrock profile of NCT, Delhi after collating borehole data from CGWB with GSI map (EREC, IMD 2005)

Earthquakes of magnitude from 3 to 7.4 have been observed in and around Delhi during the past three centuries. Figure 3 shows the epicentres of some moderate and large earthquakes. The Himalayan The Bridge and Structural Engineer

Dealing with earthquakes in and around Delhi that are inherently unpredictable calls for some of the fundamental questions (Rajendran et al, 2004): (a) Which are the potential faults that are likely to generate future earthquakes and what would be the expected magnitude? (b) Given that information (assumed in this study as a scenario earthquake of M7.2), the next obvious question is, what type of ground shaking is expected at a specific site? How would an RCC or a masonry structure situated on a rock surface or soil strata respond to the expected ground shaking? Sharma et al (2003) computed peak ground horizontal accelerations at bedrock level in the Delhi region due to the seismogenic sources present around Delhi. The Volume 45 Number 4 December 2015â&#x20AC;&#x192; 41


entire area is divided into six seismogenic sources for which seismic hazard analysis was carried out using the complete and extreme part of the seismicity data. Maximum likelihood estimates of hazard parameters viz., seismic activity rate Îť, b value and maximum probable earthquake Mmax were made for each zone. The return periods and the probabilities of occurrence of various magnitudes for return periods of 50, 100 and 1000 years were also computed for each zone. The peak ground acceleration (PGA) values for 20% exceedance in 50 years were then computed for each zone (Sharma and Wason, 2004). The maximum PGA value considering all the zones is 0.34 g, which is due to the Mathura fault zone. The Mathura fault zone and the Sohna fault zone were observed to be contributing maximum PGA values in the Delhi region governing the iso-acceleration contours computed for the region. The seismic zonation map for the PGA values at the bedrock level has been evolved. This has been used directly as input for the microzonation of ground motion at the surface by incorporating the local site conditions. Site amplification factor varying between 1 to 2.5 (Rao and Satyam, 2007) were used for computation of liquefaction potential. The detail of the evaluation of liquefaction potential will be discussed in the latter section.

4. Seismic Microzonation of Delhi Seismic microzonation is a procedure for estimating the total seismic hazard from ground shaking and related phenomena by taking into account the effects of local site conditions. The subsurface and topographic conditions can amplify or reduce the peak ground acceleration at a site with respect to what would be expected for firm ground at that location (Sharma and Wason, 2004). These local effects would then be incorporated in a seismic microzonation map. Seismic microzonation parameters can be used jointly with other scientific data banks, integrated in an expert system, to prepare land use and urban planning maps fully accounting for the complete interaction between the solid earth system (Shukla et al, 2007), the environmental system and the social, economic and political system, in addition to providing well estimated seismic inputs needed for seismically resilient building design. The construction of an integrated expert system will make it possible to tackle the problem at its widest level of generality and to maintain the dynamic updating of zoning models, 42â&#x20AC;&#x192; Volume 45

Number 4 December 2015

warranted by new data and the development of new strategies for model building (Andras et al 1999). Considering the situation in Delhi, a drastic change is required in the basic approach to hazard identification that must no longer be considered a post disaster activity. As clearly underlined by the losses of lives and properties suffered in the wake of the earthquakes events of Uttarkashi (1991), Latur (1993), Jabalpur (1997), Bhuj (2001), Kashmir (2005), Sikkim (2011) and most recently in Nepal (2015) and Manipur (2016) it is extremely important to anticipate as best as possible, the future probable earthquake threats in the country as well as areas of high vulnerability (Bilham et al 2003), so that effective measures are required to mitigate their impacts on built infrastructure. To initiate Seismic Hazard Microzonation of Delhi on 1:10000 scale (NCS, 2015), project specific toposheets were generated by Survey of India (SoI), representing NCT Delhi in 74 toposheets. Geological map on 1:50000 scale was upgraded on 1:10000 and using high resolution satellite imageries of recent origin. An intensive programme of geotechnical and geophysical investigations at more than 500 sites spread over NCT Delhi was taken up. Based on Probabilistic Seismic Hazard Analysis and subsequent ground response spectra of soil, several thematic and product maps have been developed. Based on these, an integrated hazard index map has been developed; classifying NCT Delhi in Low, Moderate and High hazard Zones (DST 2004, NCS 2015, Shukla et al 2007).

5. Earthquake Induced Liquefaction Liquefaction is a phenomenon in which shaking of a water-saturated sediment temporarily loses its strength due to the increases in the cyclic stress induced pore pressure and it behaves as a liquid. In qualitative terms, the cause of liquefaction was described very well by Seed and Idriss (1982): "If a saturated sand is subjected to ground vibrations, it tends to compact and decrease in volume; if drainage is unable to occur, the tendency to decrease in volume results in an increase in pore water pressure, and if the pore water pressure builds up to the point at which it is equal to the overburden pressure, the effective stress becomes zero, the sand loses its strength completely, and it develops a liquefied state. "Soils that liquefy tend to be young, loose, granular soils that are saturated with water (National Research Council, The Bridge and Structural Engineer


1985). Unsaturated soils do not liquefy, but they may undergo differential settlement (Kramer, 1996). If an earthquake induces liquefaction, several things can happen: (1) The liquefied layer and everything lying on top of it may move down slope; (2) The liquefied layer may oscillate with displacements large enough to rupture pipelines, move bridge abutments, or rupture building foundations; and (3) Light objects, such as underground storage tanks, can float toward the surface, and heavy objects, such as buildings, can sink. Typical displacements can range from few centimetres to meters. Thus, if the soil at a site liquefies, the total damage resulting from an earthquake can be dramatically increased from that caused by shaking alone. Three phenomena generally will be induced by ground shaking during a strong earthquake: (1) Amplification of ground shaking by a "soft" soil column; (2) Liquefaction of water-saturated sand, silt, or gravel, creating areas of "quicksand"; and (3) Landslides/ flow slides/lateral spreading, including rock falls and rock slides, triggered by shaking, even on relatively gentle slopes. The liquefaction of a soil under seismic excitation requires two types of conditions: (1) Permanent factors: soil characteristics and parameters describing the state of the soil. Soil is known to have susceptibility to liquefaction when it is relatively pulverulent and uncompacted, under little stress and water-logged. (2) Aggravating factors: the first is an earthquake acting as a trigger. There are a number of different ways to evaluate the liquefaction susceptibility of a soil deposit. They are organized as follows (Kramer, 1996), (1) Historical criteria: Soils that have liquefied in the past can liquefy again in future earthquakes. (2) Geological criteria: Saturated soil deposits that have been created by sedimentation in rivers and lakes, deposition of debris or eroded material, or deposits formed by wind action can be very liquefaction susceptible. (3) Compositional criteria: Liquefaction susceptibility depends on the soil type. Soils composed of particles that are all about the same size are more susceptible to liquefaction than soils with a wide range of particle sizes. (4) State criteria: At a given effective stress level, looser soils are more susceptible to liquefaction than dense soils. For a given density, soils at high effective stresses are generally more susceptible to liquefaction than soils at low effective stresses. The Bridge and Structural Engineer

6. Liquefaction Hazards Liquefaction is a physical process which occurs during some earthquakes (Fig. 4) and it may lead to ground failure. The shaking causes increased pore water pressure which reduces the effective stress, and therefore reduces the shear strength of the sand. If there is a dry soil crust or impermeable cap, the excess water will sometimes come to the surface through cracks in the confining layer, bringing liquefied sand with it, creating “sand boils”. The soil liquefaction depends on various factors such as the magnitude of earthquake; intensity and duration of ground motion; the distance from the source of the earthquake; site specific conditions; ground acceleration; type of soil and thickness of the soil deposit; soft; young; relative density; grain size distribution; fines content; plasticity of fines; degree of saturation; confining pressure; permeability characteristics of soil layer; position and fluctuations of the groundwater table; reduction of effective stress and shear modulus degradation

Fig. 4 :Ticketing Counter at the Jetty in Port Blair area collapsed during when the underlying soil layer reportedly liquefied and resulted in collapse of RC piles of the jetty structure during the 2004 Sumatra Earthquake

Liquefaction of saturated soil deposits during moderate to great earthquake is one of the most pervasive threats to the safety of structures. Fig. 5 shows the recreation of liquefaction in saturated Yamuna sand in 2 litre bottle having embedded round objects and overground objects, which is subjected to tamping as well as controlled vibration. The experiment showed that Yamuna sand in its loosest state, as shown in the experiment, are susceptible to liquefaction. Excess Pore Water pressure (PWP) fluctuations during liquefaction were noted visually and the same are plotted in Fig. 5. As the sand specimen was made at loosest possible state, the experiment showed significant reduction in height of sand as no. of tamping increased. Specimen ht and water column in the 1mm dia stand pipe were taken at every stage of tamping. Bottle cap was kept in open Volume 45 Number 4 December 2015  43


in the Air. As envisaged in the above experiment the embedded object came up to the surface and the over ground object came down, which is shown in Fig. 6 through another laboratory test.

Fig. 5: Relative changes of excess PWP (in terms of water column above static ht of water in the 2 litre bottle) captured by open stand pipe Vs. No. of Tamping. Spike in the water ht. vs. no. of tamping curve is the indicative of liquefaction occurrence in Yamuna sand (Delhi)

Fig. 6: A laboratory liquefaction test showing the buoyancy effect on embedded object and tilting of building block due loss of shear strength created by external excitation

Liquefaction resulted in large sand boils, excessive settlement, loss of bearing capacity, lateral spreading, landslides, mud flows and slope movements (Bray et al, 2003). The type of failure and amount of ground displacement are a function of several parameters including the looseness of the liquefied soil layer, the thickness and extent of the liquefied layer, the thickness and permeability of nonliquefied material overlying the liquefied layer, the ground slope, and the nearness of a free face (Liquefaction of both natural and artificial deposits has been observed in almost all major earthquakes, e.g. 1995 Kobe (Japan), 1999 Kocaeli (Turkey), 1999 Chi-Chi (Taiwan), 2001 Bhuj (India). Currently, several in situ techniques, such as SPT, CPT, shear wave velocity, BPT are used for assessing the liquefaction potential of soil deposits. These methods are based on the cyclic stress 44  Volume 45

Number 4 December 2015

approach, first proposed by Seed and Idriss (1971) and subsequently developed by Seed and Idriss (1982), NRC (1985), Idriss (1999), Youd and Idriss (1997) and Idriss and Boulanger (2004, 2006, 2008). In this approach, the liquefaction potential of a site is evaluated based on the estimation of: (1) the Cyclic Stress Ratio (CSR) induced by the earthquake; and (2) the Cyclic Resistance Ratio (CRR) of the soil, which represents the cyclic stress ratio required to initiate liquefaction in a given number of loading cycles. If the CSR is greater than the CRR, the soil is expected to liquefy during the earthquake. Additionally, several correction factors have been incorporated in the design procedure that account for earthquake magnitude scaling factor, high overburden pressures, and initial static shear stresses (Tokimatsu and Seed 1987, Youd 1989, seed et al 1984). Liquefaction hazard at a site is commonly expressed in terms of a factor of safety. This factor is defined as the ratio between the available liquefaction resistances, expressed in terms of the cyclic stresses required causing liquefaction, and the cyclic stresses generated by the design earthquake. Both of these stress parameters are commonly normalized with respect to the effective overburden stress at the depth in question to define a cyclic resistance ratio (CRR) and a cyclic stress ratio induced by the earthquake (CSR) (Seed and Idriss, 1971). In the Cyclic stress approach, parameters concerned are: (i) height of the soil column over the point of interest, (ii) in situ shear resistance of soil, (iii) soil density, (iv) granulometric composition of the soil (often noted as fines content, FC), (v) ground water level, and (vi) Moment magnitude of Earthquake (Mw) to be considered for the peak ground acceleration (amax).

7.

Evaluation of liquefaction based on SPT

Standard penetration test (SPT) is widely used for the determination of insitu shear resistance of noncohesive soils. The SPT consists of driving a 2-inch (5-cm) outside diameter (OD) “split barrel” sampler at the bottom of an open borehole with a 140-pound (63.6-kg) hammer dropped 30 inches (75 cm) (Seed and Harder 1990). The “N” value is the number of blows to drive the sampler the last 1 foot (30 cm), expressed in blows per foot. After the penetration test is completed, the sampler is retrieved from the hole. The split barrel is opened, the soil is classified, The Bridge and Structural Engineer


and a moisture specimen is obtained. After the test, theborehole is extended to the next test depth and the process is repeated usually at every 1.5 m interval. SPT soil samples are disturbed during the driving process and cannot be used as undisturbed specimens for laboratory testing. There are a number of advantages to the SPT: (1) The test is widely used, and often local experienceis well developed. (2) The test is simple, and many drillers can perform the test. (3) The SPT equipment is rugged, and the test can beperformed in a wide range of soil conditions. (4) There are numerous correlations for predicting engineering properties with a good degree of confidence. (5) The SPT is the only in place test that collects a soil sample. There are significant disadvantages, such as, this test does not provide continuous samples. Moreover, the samples cannot reproduce test conditions in the field Drilling disturbance, mechanical variability, and operator variability etc. can cause a significant variation in test results. The SPT should not be used unless the testing is observed and logged in detail. However, this method is quick, economical and complete information on the sub-soil profile, viz. insitu density, ground water level, thickness of soil layer that may undergo liquefaction, grain size distribution, void ratio, water content, fines content, are known (Robertson and Wride, 1998). The empirical method of liquefaction analysis relies on correlations between observed cases of liquefaction and measurementsmade in the field. Seed and Idriss (1971, 1982) first published thewidely used “simplified procedure” utilizing the Standard Penetration Test (SPT). In 1996, a workshop sponsored by the National Center for Earthquake Engineering Research (NCEER) was convened by Professors T.L. Youd and I.M. Idriss with 20 experts to review and update the simplified procedure which had last been updated in 1985. The update of the simplified procedure that resulted from the NCEER workshop is summarized in NCEER (1997) and Youd et al. (2001). The first step in the liquefaction hazard evaluation is to define the normalized cyclic shear stress ratio (CSR) from the peak horizontal ground acceleration expected at the site. The Bridge and Structural Engineer

CSR = 0.65 (amax/ g) ( σo/σ 'o)rd

(1)

where (amax/g) = peak horizontal acceleration at ground surface expressed as a decimal fraction of gravity, σ0 = the vertical total stress in the soil at the depth in question, σ0' = the vertical effective stress at the same depth, and rd = deformation-related stress reduction factor. The stress reduction factor, rd, was originally determined using a plot developed by Seed and Idriss (1971) showing the reduction factor versus depth. After consensus from the Liquefaction Workshop the modified expression for rd: rd = 1.0 – 0.00765z for z <= 9.15 m

(2a)

rd = 1.174 – 0.267z for 9.15 m < z <= 23 m

(2b)

The second step in the liquefaction hazard evaluation involves determination of the normalized cyclic resistance ratio (CRR). The most commonly used empirical relationship for determining CRR was originally compiled by Seed et al. (1985). This relationship compares CRR with corrected Standard Penetration Test (SPT) resistance, (N1)60, from sites where liquefaction did or did not develop during past earthquakes. Figure 7 shows this relationship for Magnitude 7.5 earthquakes, with an adjustment at low values of CRR recommended by the Liquefaction Workshop.

Fig. 7: CSR/CRR plot against SPT blow count (Seed and Idriss, 1981

Seed et al. (1985) proposed a "standard" blowcount, N60, which corresponds to a transfer of approximately 60% of the theoretical free-fall hammer energy. In Volume 45 Number 4 December 2015  45


Figure 7, CRRs calculated for various sites are plotted against (N1)60, where (N1)60 is the SPT blowcount normalized for an overburden stress of 100 kPa and for an energy ratio of 60 percent. Solid symbols represent sites where liquefaction occurred and open symbols represent sites where surface evidence of liquefaction was not found. Curves were drawn through the data to separate regions where liquefaction did and did not develop. As shown, curves are given for soils with fines contents (FC) ranging from less than 5 to 35 percent. While Figure 7 provides information about the variation in CRR with fines content, the preferred approach from the Liquefaction Workshop for adjusting for fines is to correct (N1)60 to an equivalent clean sand value, (N1)60cs using the following equations: (N1)60cs = α + β(N1)60

(3)

where α and β = coefficients determined from the following relationships: α = 0 for FC ≤5% α = exp[1.76 – (190/FC2)] for 5% < FC < 35% α = 5.0 for FC ≥35% β = 1.0 for FC ≤5% β = [0.99 + (FC1.5/1000) for 5% < FC < 35% Several other corrections are made to (N1)60, as represented in the following equation: (4)

Where Nfield = measured standard penetration resistance; CN = factor to normalize N into a common reference effective overburden stress; CE = correction for hammer energy ratio (ER); CB = correction factor for borehole diameter; CR = correction factor for rod length; and CS = correction for samples with or without liners. Values given in Youd, et al., (2001). The effective vertical stress, σ'vo, is the stress at the time of the SPT measurement. The energy calibration term, CE has a very significant effect on the (N1)60 used to compute CRR. The value of this correction factor can vary greatly depending on the SPT hammer system used in the field and on site conditions. For important sites where CE could result in changes from liquefied to non liquefied, energy ratio measurements should be made. These measurements are relatively inexpensive and represent a small increase in overall field exploration costs. 46  Volume 45

Number 4 December 2015

The final step in the liquefaction hazard evaluation is the computation of the factor of safety for level ground liquefaction resistance has been defined as fs = CSRliq / CSReq where CSReq is the cyclic stress ratio generated by the anticipated earthquake ground motions at the site, and CSRliq is the cyclic stress ratio required to generate liquefaction (Seed and Idriss 1982, Youd and Noble 1997). The factor of safety against liquefaction, FL can be estimated as FL = [(τav/α≠0 )1 for M = M]/τav/α≠0 (5) Ranges for FL for liquefaction are suggested as: FL < 1.0, Liquefaction occurs certainly; 2.0 ≤FL < 1.0, Liquefaction likely; 2.0 ≤FL, No liquefaction. A factor of safety in the range of about 1.1 is generally acceptable for single family dwellings, while a higher value in the range of 1.3 is appropriate for more critical structures.

8. Analysis of Borehole data

β = 1.2 for FC ≥35%

(N1)60 = Nfield CN CE CB CR CS

Before computing the factor of safety from liquefaction, the CRR result obtained (using the corrected SPT blow count identified in the equation for (N1)60 must be corrected for earthquake magnitude M if the magnitude differs from 7.5.

About 2500 bore hole data from 491 projects have been analysed. Sub-soil profile of each borehole has been checked keeping in view that: a) range of field SPT-N values were minimum and not exceeding 30, b) sub-soil layer having sandy-silty (SM, SP, CL-ML) of minimum 1m thick were present within 15 m below ground level, c)

the ground water table was present at the middle or above the selected soil layer in item (b),

d) in case no. of liquefying layers are more than one within explored depth then the weakest one, i.e. having minimum SPT-N value is taken as liquefying layer, and e)

all the above information are categorized in five layers, existing ground level to 3 m, 3-6 m, 6-10 m, 10-15 m and >15 m.

Atleast one borehole from each project site having less than 5 boreholes has been selected. Based on the above factor of Safety against liquefaction following The Bridge and Structural Engineer


tap per Idriss and Boulanger (2004, 2008) (Fig. 8) Liquefaction susceptibility of Delhi based on 491 project site bore log data collected till 2004 is shown in Fig. 9. Final microzonation report recently published by NSC, Delhi is also shown in Fig. 10 at depth range from 3-6 m. The report consists of systematic categorization of liquefaction hazards in NCT Delhi, along with many other parameters.

Fig. 9: Liquefaction susceptibility map at 3-6m below ground level (1st cut seismic microzonation (EREC, IMD, 2005)

Fig. 8: Comparison of the frozen sample-based Yoshimi et al. (1994) and Tokimatsu and Yoshimi (1983) correlations to the case history data and case history-based correlation by Idriss-Boulanger (2004, 2008)

9.

Mitigation of Liquefiable Soils

The risk of liquefaction and ground deformation can be reduced by the following types of ground improvement: densification, solidification, drainage, dewatering, and reinforcement. Conventional ground mitigation measures, such as removal and recompaction, or bypassing the weak strata by costly deep foundations, are not always practical. Consequently, alternative methods of ground treatment, including compaction grouting, vibro-compaction, stone column construction, The Bridge and Structural Engineer

Fig. 10: Liquefaction susceptibility map at 3-6m bgl (NSC, 2015)

Volume 45 Number 4 December 2015â&#x20AC;&#x192; 47


dynamic compaction, and geosynthetic reinforcement are often specified (Hausmann, 1990). These techniques allow for the remediation of potentially liquefiable soils at depth without the excavation procedures typical of those applied in densely developed areas.

Soil reinforcement provides resistance to ground deformation. Shake table tests indicate that continuous underground walls can control horizontal ground movement. Their effectiveness depends on such factors as quantity, orientation, shear resistance, and excitation direction.

The main goal of most soil improvement techniques used for reducing liquefaction hazards is to avoid large increases in pore water pressure during earthquake shaking. This can be achieved by densification of the soil and/or improvement of its drainage capacity. Soil densification is generally considered highly reliable, and the standard remedial measure against liquefaction (Andus et al, 1999). It reduces the void space of the soil, thereby decreasing the potential for volumetric change that would lead to liquefaction. Resistance to shear deformation also increases with increased density. Several sites improved by densification performed well â&#x20AC;˘during the 1964 Niigata, Japan, 1978 Miyagiken-oki, Japan, 1989 Loma Prieta, California, and 1994 Northridge, California, earthquakes (Iai 1994, JSSMFE 1995). Other methods such as the use of shallow or deep foundations designed to accommodate the occurrence of liquefaction and associated vertical and horizontal deformations may also achieve an acceptable level of risk. Mitigation should provide suitable levels of protection with regard to potential large lateral spread or flow failures, and more localized problems including bearing failure, settlements, and limited lateral displacements.

Vibro-compaction sometimes referred to as Vibroflotation, is the rearrangement of soil particles into a denser configuration by the use of powerful depth vibration. It involves the use of a vibrating probe that can penetrate granular soil to depths of over 30m. The vibrations of the probe cause the grain structure to collapse thereby densifying the soil surrounding the probe. To treat an area of potentially liquefiable soil, the vibroflot is raised and lowered in a grid pattern. Vibro Replacement is a combination of vibroflotation with a gravel backfill resulting in stone columns, which not only increases the amount of densification, but provides a degree of reinforcement and a potentially effective means of drainage.

Solidification is also considered a highly reliable remedial measure against liquefaction. It prevents soil particle movement and provides cohesive strength. During the 1989 Loma Prieta earthquake, the few sites improved by solidification techniques performed well (Mitchell and Wentz, 1991; Graf, 1992a). Lowering the ground water level by dewatering reduces the degree of saturation, thereby preventing the development of excess pore water pressure which would lead to liquefaction. Dewatering is a difficult and very expensive task, since both upstream and downstream seepage cutoffs are usually required, and pumps must be maintained constantly. 48â&#x20AC;&#x192; Volume 45

Number 4 December 2015

Densification by dynamic compaction Fig. 11 is performed by dropping a heavy weight of steel or concrete in a grid pattern from heights of 10 to 30 m. It provides an economical way of improving soil for mitigation of liquefaction hazards. Local liquefaction can be initiated beneath the drop point making it easier for the sand grains to densify. When the excess porewater pressure from the dynamic loading dissipates, additional densification occurs. However, the process is somewhat invasive; the surface of the soil may require shallow compaction with possible addition of granular fill following dynamic compaction. Hardening and/or mixing techniques seek to reduce the void space in the liquefiable soil by introducing grout materials either through permeation, mixing mechanically, or jetting. These techniques are known as permeation grouting, soil mixing, or jet grouting. Compaction grouting is a technique whereby a slowflowing water/sand/cement mix is injected under pressure into a granular soil. The grout forms a bulb that displaces and hence densifies, the surrounding soil. Compaction grouting is a good option if the foundation of an existing building requires improvement, since it is possible to inject the grout from the side or at an inclined angle to reach beneath the building. The Bridge and Structural Engineer


also the possible impact of vibration from future earthquake in Delhi-NCR. In some cases, engineers may decide to design mitigation measures to prevent liquefaction of certain soil types and allow limited deformations in others (i.e., allow some liquefaction). Liquefaction mitigation is likely to treat the ground underneath the structure to a sufficient depth, in most cases the bearing capacity reduction due to liquefiable ground outside the structure is not likely to govern the design. Instead, the propagation of excess pore pressures from liquefied to improved ground tends to determine the lateral extent of improvement required.

Fig. 11: Dynamic compaction of soils to increase density at a site

Liquefaction hazards can be reduced by increasing the drainage ability of the soil. If the porewater within the soil can drain freely, the build-up of excess pore water pressure will be reduced. Drainage techniques include installation of drains of gravel, sand or synthetic materials. Synthetic wick drains can be installed at various angles, in contrast to gravel or sand drains that are usually installed vertically. Drainage techniques are often used in combination with other types of soil improvement techniques for more effective liquefaction hazard reduction.Vertical drains are installed under a surcharge load to accelerate the drainage of impervious soils and thus speed up consolidation. These drains provide a shorter path for the water to flow through to get away from the soil. Time to drain clay layers can be reduced from years to a couple of months

10. Performance Assessment Implementation of mitigation measures should be designed to either eliminate all liquefaction potential or to allow partial improvement of the soils, provided the structure in question is designed to accommodate the resulting liquefaction-induced vertical and horizontal deformations. Fig. 12 shows one such cost effective ground improvement measures for shallow foundation that not only insures improvement of bearing capacity but The Bridge and Structural Engineer

Fig. 12: Shallow Foundation treatment in a residential building in NOIDA, Uttar Pradesh, India by manually driven micopiles, a) Holes made by driving pointed rod about 1m depth, b) filling the holes by cement-sand-brick chips, c) construction of foundation on the same

A number of methods can be used to verify the effectiveness of soil improvement. In-situ techniques (Fig. 13) are popular because of the limitations of many laboratory techniques. Usually, in-situ test are performed to evaluate the liquefaction potential of a soil deposit before the improvement was attempted. With the knowledge of the existing ground characteristics, one can then specify a necessary level of improvement in terms of insitu test parameters. Performing in-situ tests after improvement when completed allows one to decide if the degree of improvement was satisfactory. In some cases, the extent of the improvement is not reflected in in-situ test results until sometime after the improvement has been completed. Liquefaction mitigation and performance criteria vary according to the acceptable level of risk for each structure type and human occupation considerations. Volume 45 Number 4 December 2015â&#x20AC;&#x192; 49


Ajeet Pande, Dr H S Mandal, Dr R.K. Singh, Dr Rajesh Prakash and other staffs of EREC-IMD, who have been instrumental to go on with this work as mission mode project under DST. Fig. 1, 2 and 9 presented in the paper were prepared during 1st cut seismic microzonation work and hence the interpretation of liquefaction potential; for which due acknowledgement is on record to all co-researchers of EREC, IMD. Fig. 13: Improving liquefaction potential of the soils at the Gautam Buddha University campus site, UP, India (Source: Cengrs Geotechnica, Delhi)

13. References 1.

11. Conclusions This paper presented a series of probabilistic liquefaction opportunity maps for NCT, Delhi taking into account of (a) SPT energy transfer efficiency as 60%, (b) a scenario earthquake M7.2 and updated seismo-tectonic map, (c) peak ground acceleration (PGA) at bed rock level varying from 0.1 to 0.34g (Sharma et al, 2003) with maximum magnification factor at surface level as 2.5, (d) pre and post monsoon ground water table map from Central Ground Water Board (CGWB), (e) updated bed rock profile prepared from Geological Survey of India (GSI) and CGWB borehole records. Based on 2500 borehole data (6 to 30m depth, SPT-N value 2 to 50) from 491 geotechnical investigation reports procured from DDA, CPWD, DMRC and various other agencies, this paper presented liquefaction potential map of NCT, Delhi following semi-empirical approach of Idriss and Boulanger (2004). Results show that about 80% of the soil cover in Delhi are of liquefying type (SP, SM, CL-ML) and Yamuna river belt area is having high potential for liquefaction susceptibility. Taking notable ground improvement techniques followed across the world and their good performance against earthquakes, some measures for Delhi-NCR as per detail microzonation report published by National Seismological Centre (2015) are discussed.

12. Acknowledgement The author gratefully acknowledges the cooperative research work on 1st cut seismic microzonation of Delhi carried out at EREC, IMD (now National Seismological Centre, MoES) during 2005-2006 and the colleagues with special mention about Dr A.K. Shukla, Dr P.S. Mishra, Dr T.K. Roy, Dr 50  Volume 45

Number 4 December 2015

Ambraseys, N. and Bilham, R. (2000), A note on the Kangra Ms = 7.8 earthquake of 4 April 1905. Current Science, 2000, 79, 101–106.

2. Andrus, R., Stokoe, K.L., and Chung, R.M. (1999). “Draft Guidelines for Evaluating of Liquefaction Resistance using Shear Wave Velocity Measurements and Simplified Procedures,” Report No. NIST IR-6277, National Institute of Standards and Testing, Gaithersburg, MD. 3.

Bilham, R. and England, P. (2001), Plateau popup during the great 1897 Assam earthquake. Nature, 410, 806–809.

4. Bilham, R., Wallace, K., Blume, F. and Feldl, N. (2003), Flexure and fragmentation of the Indian plate: Mid-plate earthquakes, Indo-US Workshop on Seismicity and Geodynamics, National Geophysical Research Institute, Hyderabad, p. 27. 5. Bray, J. D., Kayen, R. E., and Faris, A. (2003). Recent Advances in Soil Liquefaction Engineering: a Unified and Consistent Framework, Keynote presentation, 26th Annual ASCE Los Angeles Geotechnical Spring Seminar, Long Beach, CA 6.

Chauhan R.K.S. (1975). “Seismotectonics of Delhi Region”. Proceedings INSA 41,pp. 429447.

7. Department of Science and Technology (DST) Report (2004), Geo-Scientific studies in and around Delhi, p. 74. 8. Geological Survey of India (2000), Seismotectonic Atlas of India and its Environs, published by Geological Survey of India. The Bridge and Structural Engineer


Ghosh, C. (2011) “Investigation of a building collapse in Delhi”, Proc. Indian Geotechnical Conference – 2011, Kochi, India, Vol. 2, pp.1064-67

17. Idriss, I. M., and Boulanger, R. W. (2008). Soil liquefaction during earthquakes. Monograph MNO-12, Earthquake Engineering Research Institute, Oakland, CA, 261 pp.

10. Graf, E.D. ( 1992). “Compaction Grouting, 1992,” Proceeding, Grouting, Soil Improvement and Geosvnthetics: Geotechnical SDecial Publication No. 30, held in New Orleans, Louisiana, on 25-28 February, R.H. Borden, R.D. Holtz and I. Juran, Eds., ASCE, New York, NY, Vol. 1, pp. 275-287.

18. IS: 1893-2002, Criteria for earthquake resistant design of structures.

11. Hausmann, M.R. (1990). Emzineering Principles of Ground Modification, McGraw-Hill, 632 p.

20. JSSMFE (1995). Remedial Measure Azainst Soil Liquefaction--From Investigation and Design to Implementation, Japanese Society for Soil Mechanics and Foundation Engineering (in Japanese; translation into English in progress).

9.

12. Iai, S., Matsunaga, Y., Morita, T., and Sakurai, H. (1994). “Effectiveness of Measures Against Liquefaction Confirrnedat aRecent Earthqualce-A Case History During the 1993 Kushiro-Oki Earthquake, Proceedim-zs, 26th Joint Meeting of United States-Jauan Panel on Wind and Seismic Effects: NIST SP 871, held in Gaithersburg, Maryland, on 17-20 May, N. Raufaste, Ed., U.S. Department of Commerce, National Institute of Standards and Technology, Gaithersburg, MD, pp. 213-325. 13. Idriss, I. M. (1985) “Evaluating Seismic Risk in Engineering Practice.” In Proceedings of the 11th International Conference on Soil Mechanics and Foundation Engineering, Vol. 1, p. 155. 14. Idriss, I. M. (1999). An update to the Seed-Idriss simplified procedure for evaluating liquefaction potential, in Proceedings, TRB Workshop on New Approaches to Liquefaction, Publication No. FHWA-RD-99-165, Federal Highway Administration, January. 15. Idriss, I. M., and Boulanger, R. W. (2004). Semi-empirical procedures for evaluating liquefaction potential during earthquakes, in Proceedings, 11th International Conference on Soil Dynamics and Earthquake Engineering, and 3rd International Conference on Earthquake Geotechnical Engineering, D. Doolin et al., eds., Stallion Press, Vol. 1, pp. 32–56. 16. Idriss, I. M., and Boulanger, R. W. (2006). Semiempirical procedures for evaluating liquefaction potential during earthquakes, J. Soil Dynamics and Earthquake Eng. 26, 115–30. The Bridge and Structural Engineer

19. Iyenger, R.N and Ghosh, S. (2004), Seiismic mapping hazards of Delhi city. Proc. 13th World Conf. on Earthquake Engineering Vancouver, B.C., Canada August 1-6, 2004, Paper No. 180, pp. 1-15.

21. Kramer, S. L. (1996). Geotechnical Earthquake Engineering. New Jersey: Prentice Hall. 22. Mitchell, J.K., and Wentz, F.J., Jr. (1991). Performance of Improved Ground Durirw the Loma Prieta Earthquake: Report No. UCB/ EERC-9 1/12, Earthquake Engineering Research Center, University of California at Berkeley, CA, 93 p. 23. National Center for Earthquake Engineering Research (NCEER). 1997. Proceedings of the NCEER Workshop on Evaluation of Liquefaction Resistance of Soils, Technical Report NCEER-97-0022, 276 p. 164, 2003 Commentary, Chapter 7 24. National Research Council. (1985). Liquefaction of Soils During Earthquakes, National Academy Press, 240p. 25. National Seismological Centre (2015). A report on Seismic Hazard Microzonation of NCT Delhi on 1:1000 scale, Ministry of Earth Science, Govt. of India, New Delhi, 404pp. 26. Parvez, I., Vaccari, F. Panza and G. F. (2004). Site-specific microzonationstudy in Delhi metropolitan city by 2-D modeling of SH and P-SV waves, Pure Appl. Geophys. 161, pp.1165–1185. 27. Parvez, I., Vaccari F. , and Panza G. F. (2006). Influence of source distanceon site-effect in Delhi city, Curr. Sci. Vol. 91, pp.827–835. Volume 45 Number 4 December 2015  51


28. Rajendran, C. P., Rajendran, K. Duarah, S. Baruah, B.P. and Earnest A. (2004), Interpreting the style of faulting and paleoseismicity associated with the 1897 Shillong, northeast India, earthquake: Implications for regional tectonism, Tectonics, 23, TC4009, doi:10.1029/2003TC001605 29. Rao, K. S. (2003), Seismic microzonation of Delhi region. In Proceedings of 12th Asian Regional Conference, Singapore, vol. 1, pp. 327–330. 30. Rao, K.S. and Satyam, D.N. (2007) Liquefaction studies for seismic microzonation of Delhi region, Current Science, Vol. 92, Mar 10, pp. 646-654. 31. Robertson, P.K., and Wride, C.E. 1998. “Evaluation of cyclic liquefaction potential using the cone penetration test,” Canadian Geotechnical Journal, Ottawa, 35(3), 442-459. 32. Sasaki, Y., and Taniguchi, E. (1982). “Shaking Table Tests on Gravel Drains to Prevent Liquefaction of Sand Deposits,” Soils and Foundations, Japanese Society of Soil Mechanics and Foundation Engineering, Vol. 22, No. 3, pp. 1-14. 33. Seed, H. B. (1979). "Soil liquefaction and cyclic mobility evaluation for level ground during earthquakes." J. Geotechnical Eng. Div, ASCE 105(GT2), 201–55. 134 34. Seed, H. B., and I. M. Idriss. (1971). “Simplified Procedure for Evaluating Soil Liquefaction Potential.” Journal of the ASCE Soil Mechanics and Foundations Division 97(SM9):1249-1273. 35. Seed, H. B., and Idriss, I. M. (1982). Ground Motions and Soil Liquefaction During Earthquakes, Earthquake Engineering Research Institute, Oakland, CA, 134 pp. 36. Seed, H. B., Idriss, I. M. and Arango, I. (1983). “Evaluation of Liquefaction Potential Using Field Performance Data.” ASCE Journal of Geotechnical Engineering 109(3):458-482. 37. Seed, H. B., Tokimatsu, K., Harder, L. F. and Chung, R. M. (1985). The Influence of SPT Procedures in Soil Liquefaction Resistance Evaluations, Report UBC/EERC-84/15. Berkeley, California: Earthquake Engineering 52  Volume 45

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Research Center. 38. Seed, H. B., Tokimatsu, K., Harder, L. F. Jr., and Chung, R. (1984). The influence of SPT procedures in soil liquefaction resistance evaluations. Earthquake Engineering Research Center, University of California, Berkeley, Report No. UCB/EERC-84/15, 50 pp. 39. Seed, H.B., and Idriss, I.M. (1982). “Ground Motions and Soil Liquefaction During Earthquakes,” Earthquake Engineering Research Institute (EERI) Monograph. 40. Seed, R. B., and L. F. Harder. (1990). “SPTBased Analysis of Cyclic Pore Pressure Generation and Undrained Residual Strength.” In Proceedings of the H. Bolton Seed Memorial Symposium, Vol. 2, pp. 351-376. 41. Sharma M. L. and H. R. Wason (2004), “Estimation of seismic hazard and seismic zonation at bed rock level for Delhi region, India, 13WCEE, Vancouver, Aug. 1- 6, paper 2043, pp.1-13. 42. Sharma, M. L., Wason, H. R. and Dimri, R. (2003), “Seismic Zonation of the Delhi region for Bedrock Ground Motion. J. Pure Appl. Geophys., vol. 160, pp. 2381-2398. 43. Shukla, A. K., Prakash, R., Singh, R. K., Mishra, P. S, and Bhatnagar, A. K., (2007) “Seismotectonics implications of Delhi region through fault plane solutions of some recent earthquakes”. Curr. Sci., Vol.93, pp.1848–1853. 44. Tokimatsu, K., and H. B. Seed. (1987). “Evaluation of Settlements in Sands Due to Earthquake Shaking.” Journal of Geotechnical Engineering ASCE 113(8):861-878. 45. Valdiya K.S. (1976) “Himalayan transverse faults and folds and their parallelism with subsurface structures of north Indian plains”. Tectonophysics, Vol 32, pp.353-386. 46. Youd, T. L. (1989). “Ground Failure Damage to Buildings During Earthquakes.” In Foundation Engineering—Current Principles and Practices, New York: American Society of Civil Engineers, Vol. 1, pp. 758-770. 47. Youd, T. L., and Carter, B. L. (2005). "Influence of soil softening and liquefaction on spectral The Bridge and Structural Engineer


acceleration." Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 131(7), 811-825. 48. Youd, T.L., and Noble, S.K., (1997). Magnitude scaling factors, Proc. NCEER Workshop on Evaluation of Liquefaction Resistance of Soils, National Center for Earthquake Engineering Research, State University of New York at Buffalo, pp.149-165. 49. Youd, T.L., Idriss, I.M., Andrus, R.D., Arango,

The Bridge and Structural Engineer

I., Castro, G., Christian, J.T., Dobry, R., Finn, W., D.L. Harder, L.F., Jr., Hynes, M.E., Ishihara, K., Koester, J.P., Liao, S.S.C., Marcuson, W. F., III, Martin, G.R. Mitchell, J.K., Moriwaki, Y., Power, M.S., Robertson, P.K., Seed, R.B., and Stokoe, K.H., II, (2001). Liquefaction resistance of soils: Summary Report from the 1996 NCEER and 1998 NCEER/NSF Workshops on Evaluation of Liquefaction Resistance of Soils, Journal of Geotechnical Geoenvironmental Engineering, ASCE, 127 (10), pp. 817-833.

Volume 45 Number 4 December 2015â&#x20AC;&#x192; 53


Prefabricated Vertical Drains – Recent Developments

G. Venkatappa RAO Former Professor & Dean, IIT, Delhi & Chairman, SAGES, Hyderabad, India gvrao.19@gmail.com

M.V.S. SREEDHAR Dept. of Civil Engineering, UCE(A), Osmania University, Hyderabad, India mvs_sreedhar@yahoo.com

Dr. G.Venkatappa Rao graduated in Civil Engineering from BITS, Pilani in 1965. He obtained M.E. and Ph.D. in Civil / Geotechincal Engineering from IISc.,Bangalore in 1968 and 1973 respectively. From 1972 to 1975, he worked as Lecturer in Civil Engineering in BITS, Pilani. In 1975, he joined IIT, Delhi as a Lecturer and became Professor in the year 1985 and served in various capacities including as Head, Department of Civil Engineering and Dean (Student Affairs). His teaching and research interests include Soil and Rock Mechanics, Marine Geotechnology, Ground Improvement, Transportation Engineering and Geosynthetics. He successfully guided 24 Ph.D. theses and over 100 M.Tech dissertations. He received several awards and served several professional bodies in various capacities. His was the only name from India included in IFAI’s 142 Names to know in Geo’ (Geotechnical Fabrics Report). He founded SAGES and STTRI with the objective of dissemination of knowledge for a better society.

Dr. M.V.S.Sreedhar graduated in Civil Engineering from the Jawaharlal Nehru Technological University (JNTU) in 1992. He obtained M.Tech in Civil / Geotechincal Engineering from IIT, Bombay in 1994. He carried out research on “Monotonic and Cyclic Response of Geosynthetic Reinforced Pond Ash” leading to the award of Ph.D. from Osmania University, in 2013. From 1994 to 2001, he worked in Civil Engineering design and execution in Govt. of India and the then Govt. of Andhra Pradesh. Since 2002, he has been working as Asst. Professor in Department of Civil Engineering, Osmania University, Hyderabad, where he specializes in academics, research and consultancy in Geotechnical Engineering.

Summary For improving the engineering behaviour of soft clayey soils, particularly on the coastal belt, PVD’s are found to be quite useful, as they increase the rate of consolidation and improve the subsoil when subject to pre-loading. At the moment the PVD’s in use are generally imported. The paper briefly presents an overview of design and construction with this type of ground improvement technique which is very economical, if there is time available. Developments in evaluating the performance of the drains in the 54  Volume 45

Number 4 December 2015

laboratory as well some indigenously developed drains are also brought forth.

1. Introduction 1.1 Pre-compression The design of foundations include safety against excessive settlement that may be detrimental to the structure apart from the safety against failure in shear and stability. The total settlement of a foundation consists of immediate/elastic settlement The Bridge and Structural Engineer


which is always present and the consolidation settlement will be present only when a fully saturated compressible clayey type of medium is present within the significant zone. The consolidation process is a time dependent process that may result in large magnitudes of settlement over longer periods of time during service of the structure resulting distress to the structure or may cause it un-serviceable. In view of this, in all important projects, it is preferred to enable the consolidating medium to undergo consolidation even before the intended structure is built under the application of a pre-load. This process is called precompression. 1.2 Consolidation acceleration methods In application of pre-load, the intensity of surcharge pressure due to pre-load plays an important role. If the preload exerts a pressure same as that of the intended structure, pre-compression may take place, but the time of consolidation may still by few years. It may not be practically feasible. Therefore it becomes necessary to accelerate the consolidation process which can be done by the following methods. 1.2.1 Application of additional preload In this process, the intensity of pressure due to the preload is provided more than that due to the intended structure such that, the required degree of consolidation takes place within a shorter period than what it could have been due to a preload of pressure equivalent to the intended structure. The effect of application of extra surcharge is illustrated in Fig. 1.

Fig. 1: The effect of preload (after Das, 1990)

However, fixing the magnitude of extra surcharge is governed by the ability of the consolidating medium. In soft clays with low bearing capacities, the preload and the extra preload may be applied in stages. If The Bridge and Structural Engineer

required, the bearing capacity may be improved by application of geosynthetics. 1.2.2 Application of vertical drains The application of extra surcharge may not be feasible in all the site conditions. However, the application of vertical drains may be a viable solution in all types of soils. When consolidation under the preload is aided with vertical drains, the excess pore water is provided drainage access in radial direction in addition to the already existing vertical drainage, by this, the drainage path is drastically decreased and hence the consolidation process is accelerated. The traditional vertical drains include the sand drains, which have now been replaced by the pre-fabricated vertical drains (PVDs). 1.3 Historical development of prefabricated vertical drains Prefabricated vertical drains were first used in Sweden in 1937. These drains were manufactured cardboard, the so-called cardboard wick (Kjellman, 1948). Approximately 10 years earlier sand drains were developed in California to expedite consolidation. Especially in the Netherlands sand drains were applied on a large scale since 1950. Dutch soil mainly consists of clay and peat layers which sometimes present to greater depths. A sandy surcharge was often placed on the top of compressible sub soil in those places where an industrial or residential estate or infrastructure had to be developed. Settlements in the subsoil were expedited by using sand drains. The synthetic drain was introduced in 1972 for a building pit at the Hemweg power station in Amsterdam. Its development was then accelerated. Synthetic drains are superior to sand drains because of their flexibility and better filtration, and they became a formidable competitor. Now-a-days, sand drains are hardly ever used. This chapter covers many aspects of the synthetic drain, from application to quality control, from design methodology to laboratory tests.

2. Operation 2.1 Principle Soil stabilization with vertical drains is used on compressible, saturated soils, like clays and peat. Volume 45 Number 4 December 2015â&#x20AC;&#x192; 55


These soils are characterized by a soft structure and a big pore capacity, normally filled with ware (pore water). When a heavy load, like an embankment or a dike, is placed on the top of clay or peat soils, settlements could occur due to the compressibility of the soil. These settlements could crate serious construction problems. The load created by the surcharge is initially carried by the pore water. However, when soil is not very permeable, water pressure will decrease gradually because the pore water is only able to flow away very slowly. Increased water pressure can crate instability of the subsoil, which in turn can create slip planes. This instability can decrease the rate of fill placement. A vertical drainage system enhances a quicker construction of the embankment without a risk of landslides. To increase the settlement process and the reduction of water pressure, it is necessary to decrease the flow path of the pore water. This can achieve by installing evenly spaced vertical drains. The pressure of this system enables the pressurized water to flow horizontally towards the nearest drain, and escape freely. By the vertical drains, the period of consolidation can be reduced from an average time of decades to only six months, or even a shorter period. By coupling thermal and/or vacuum systems, it is possible to accelerate even further. Soil improvement by means of vertical drains has been used in many civil engineering projects. Some of the applications are shown in Construction of embankments for roads, railways, airports and dikes

fines from adjacent soils without clogging. The drains are generally manufactured in a width of 100 mm and 2 mm to 6 mm thick -- a size accepted as standard worldwide. Details of required properties of drains are given Table 1 (after Rathmayer, and Komulanein, (1992) and typical core shapes of strip drains shown in Figs. 2 and 3. Table 1: Required properties of PVDs (after Rathmayer and Komulanein, 1992) No. Criteria

Unit

Required value

1.

Drain width

mm

Nominal ± 5%

2.

Equivalent dia.

mm

Nominal

3.

Dry mass

g/m

Nominal ± 10%

4.

Thickness at 20 kPa

mm

Nominal ± 10%

5.

Tensile strength (dry)

kN

>1

6.

Strain at break (dry)

%

> 15

7.

Discharge capacity at 125 kPa, T = 200C at 100 mm head

ml/s

> 10

at 200 mm head

ml/s

> 10

1.

Permeability at 50 mm mm/s > 50 k(soll) head, unloaded 2.

Land reclamation

EOS 090

µm

EOS 095

µm

depending on soil type3

3.

Tensile (trans) and kN shear strength of seam

> 1.5

4.

Seam strain at break

> 10

Sleeve

%

090 < d85 of soil and 090 < 0.15 mm.

a

Construction of ports

Residential and industrial areas

Preloading of storage and landfill depots

2.2 Geosynthetic drains Synthetic drains consist of a prefabricated strip which is very suitable for water transportation. The flexible core is manufactured of a high- quality polypropylene. Both sides have grooves, through which water can flow unimpeded. The core is warped in a strong and durable geotextile filter fabric with excellent filtration properties, allowing free access of pore water into drains. At the same time, this filter prevents piping of

56  Volume 45

Number 4 December 2015

Fig.2: Typical PVD (with outer filter geotextile and inner core)

The Bridge and Structural Engineer


layers, permeability of the filter needs to be at least as high as the permeability of the soil. Other important criteria a filter should meet are: Great mechanical strength

High resistance to bacteria and microorganisms

Durability against acids and solvents

Negligible loss of strength over a prolonged period

Minor creep under heavy compression.

Fig. 3: Typical core shapes of strip drains

Advantages of synthetic drains or PVDs: Easy, rapid installation is possible

Made of uniform material, easily stored and transported

Quality control of drain geometry, welding and weight of filter jacket and core are tested in laboratory.

3. Applications

Equipment needed is lighter than rings required for equivalent sand drains

Synthetic drains can be used in different applications viz.,

Consolidation

Stability problems

Least possible disturbance of soil layers

Tensile strength of the strips helps to preserve continuity

Drain installation possible to depths exceeding 40 m

Dewatering Pile foundation

Preload

o

Surcharge method

o

Vacuum consolidation

The core of the PVD guarantees a higher vertical discharge capacity than a sand drain with a equivalent diameter of a 100 mm synthetic drain amounts to 65 mm, a comparison between both drain disturbances can be made by using the Barron’s formula, theoretically, approximately twice as many drains should have to be used in order to result into a similar progress in settlement. Because of the considerably lower price of synthetic drains, this system is ultimately much more cost-effective than sand drains.

o

Deep well point system

Contrary to fabrics with a straight passage, through which soil particles can flow freely and enabling it limit the movement of soil particles and to prevent clogging. In general, the filter is required that prevents clogging by soil particles but is sufficiently permeable. When compressible subsoil contains permeable soil in horizontal layers, pressurized pore water will flow onto these layers and thus to the nearest drain. So as to make the best possible use of these permeable

The requirements of a prefab drain should meet; largely depend on the following circumstances:

Low cost (site treatment may be possible for only one-fourth the costs of traditional sand drains).

The Bridge and Structural Engineer

Environmental technique

o

Landfills

o

Sludge depots

o

Soil cleaning

o Degassing

4. Requirements

– Size of the settlement\ – Consolidation period – Drain length – Size of the embankment – Method of installation. Volume 45 Number 4 December 2015  57


The required discharge capacity for pore water must be guaranteed at all times to ensure the best possible settlement progress. In the Netherlands, the requirements for prefab drains are summarized in classification schedule drawn up by the workshop “Vertical drainage” of the C.R.O.W. (the Dutch Centre for the issuance of rules and regulations and investigation in civil engineering). Pertaining to the vertical drainage, C.R.O.W. formulated the following requirements. 4.1 Drain strength During installation high forces can occur in the strip drains. Especially when a vibrator is used to install the drains, large acceleration during free fall of the mandrel have to be transferred to the drain roll. This creates large forces that have to be absorbed by the elongation and strength of the drain. The diameter of the transportation rolls is of great importance, too. In the past, paper filters were used that teared up, while the core stayed intact. These failures could not be detected because the drain was not visible during installation. The limited elongation of the filter paper did not allow for large forces in the drain. Therefore, the following mechanical requirements were specified: Elongation of filter,

εf ≥ 2%

Strength filter Ff ≥ 0.5kN Elongation at 0.5 kN ε0.5KN ≥ 2% For the rolls in the installation rig, the following requirement was set: Diameter transportation rolls > 150 mm 4.2 Filter strength During the consolidation process, the filter fabric may be pressed into diameter channels and therefore has to retain its original strength under wet conditions. As the load in the fabric strongly depends on the configuration of the core, it is very difficult to set a uniform requirement for filter strength. Therefore, the filter strength is related to the discharge test. During this test the filter may not fail at a maximum cell pressure of 300KN/m2. Moreover, the filter should be wrapped tightly around the core to prevent penetration of the filter into the channels. To prevent damaging of the filter fabric during installation, the following requirement is proposed: 58  Volume 45

Number 4 December 2015

Tear strength Tr > 150 N 4.3 Flow capacity of prefabricated drains When calculating vertical drainage system, the resistance in the drain is taken as zero. The flow capacity is determined by the free volume of the drain. The free volume is influenced by the compression of the core and the depression of the filter in the channels as a result of the horizontal soil pressure. Depending on the length of the drain, the filling speed, the compression and the ultimate load, the discharge of prefab drain (qw) generally has to meet the following requirements: Drain length < 10 m and no stability problems: qw(straight)> 10 x 10-6m3/s= 315 m3/year qw(buckled) > 7.5 x 10-6 m3/s = 236 m3/year Drain length > 10 m and/or stability problems: qw(straight)> 50 x 10-6m3/s= 1517 m3/year qw(buckled) > 32.5 x 10-6 m3/s = 1183 m3/year The discharge capacity of prefab drains is determined in accordance with the method as described in chapter “Lab tests” of this paper. 4.4 Permeability of the filter Pore water should be enabled to enter the drain without too much resistance. Tests with filters are always carried out with clean water and a clean filter. Nowadays, permittivity is usually considered as the calculating value for permeability (ψ = kf/δf). However, as a result of the flowing out of soil particles, the filter will silt up fast, lowering permittivity by a factor 1000. Permittivity of the filter must therefore be a factor 1000 larger than most permeable soil types in which vertical drainage is applied. The filter criteria will then be:

ψ > 5 x 10-3/s

5. Determination Of Spacing Of Vertical Drains The spacing of vertical drains can be calculated by the different conditions. 5.1 Radial Consolidation The solution of radial water flow toward the central drain of a cylinder of soil undergoing one The Bridge and Structural Engineer


dimensional strain goes back to Rendulic (1935). The result is generally expressed in terms of the average consolidation ratio for radial drainage Ur Ur = 1 - exp

……… Eq.1

Where Ur = average degree of consolidation ratio for radial drainage

Tv = cvt / L2

t = time, s

cv = coefficient of vertical consolidation, m2/s

Tr = Cht / D2 = time factor for radial drainage Ch = coefficient of horizontal (or radial) drainage t

= time elapsed since application of the surcharge

D = equivalent diameter of cylinder of soil around drain = 1.06 s for triangular pattern of drains and = 1.13s for square grid pattern of drains shown in Fig. 4. s = spacing of drain n = D/d d = drain diameter (or equivalent diameter for strip drains)

L = longest drainage path in clay layer (m); equal to half of H with top and bottom drainage, and equal to H with top drainage only

step 2: Set Uvt = 0.9 (must be > Ur) step 3: Find Ur from 1- Uvr = (1-Uv)(1-Ur) step 4: Rewrite eq.1 in the form D2α = -Ch / In(1-Ur)

And solve for D by successive approximation.

step 5: Compute spacing’s’ for triangular or square pattern s = D/1.06 for triangular pattern; D/1.13 for square grid pattern

α = n2 In n/ (n2-1) – (3n2-1)/4n2

Fig. 4: Vertical drain patterns

5.2 Combined vertical and radial consolidation It would not be unusual in practice that the drain spacing giving a specified percent consolidation over a fixed time period would be based on considering radial consolidation only. However, Carillo (1942) showed how an average degree of consolidation Uvr for combined vertical and radial water flow can be calculated. Assuming all relevant soil properties as well as the drain type (and its equivalent diameter) are given, one would proceed as follow: Step 1: Calculate Tv given cv, L and t; determine Uv (from Fig. 5). Where M = (2m+1)π/2 The Bridge and Structural Engineer

Eq. 2.

Fig. 5: Vertical consolidation Vs time factor

6. Installation 6.1 General In order to prevent damage to and smearing of the drain, a rectangle steel pipe is used to install the prefab drain. the size of this mandrel is minimal, in order to prevent resistance during installation and to avoid disturbance of the subsoil. In the course of time, numerous different types of machines have been constructed to move the mandrel up and down as quickly as possible. In general, these machines, the so-called drain stitches, can be divided into groups; the static machines, pushing the mandrel into the soil, and the dynamic machines, vibrating the mandrel into the soil. Testing for the effect of the both systems has taught us that the method of insertion has no influence on the ultimate effect of the drain. The pros and cons of these systems Volume 45 Number 4 December 2015  59


are dealt with in the following chapters. Depending on the insertion system, the subsoil and the circumstances, daily production can range from 1 000 to 10 000 m. The effect of the insertion depth on production is shown in adjoining graph. as installation costs cover at least 50% of the total costs, very short or very long drains are found to be, relatively, considerably more expensive. Moreover, the price of an accessory anchor plate for short drains is a major expense factor. 6.2 Pushing There is a great variety of machines that push installation mandrels into the soil. The first used machines pulled a diamond- shaped, as regards sectional view, mandrel through transportation rolls. The rather weak insertion pipe had to be supported along the entire length. Later on, stitches were designed which pulled the mandrel into the soil using a steel cable, driven by a hydraulic winch or ram. Here, too, facilities are required to restrict buckling length. Besides these, machines were developed that pushed the mandrel downwards by direct drive. last mentioned type of machines include the roller stitcher, which uses rubber or steel wheels to move the mandrel downwards, or the rack stitcher, which has a rack welded onto the mandrel which is driven by a geared hydrometer. the latter type restricts the buckle length to 1 m so that the mandrel does not required any support and the stitcher can be manufactured as a lighter type. The installation principle is described as follows; 1.

The tip of a roll of strip drain, fitted to the side of the stitcher, is guided into the stitcher, runs over a transportation roll and is led down into the mandrel. At the bottom of the mandrel, the drains are provided with an anchor plate which is pulled against the bottom of the mandrel.

anchor plate is fitted. The stitcher is moved to the next insertion point and the cycle is to be repeated. Rolls of drainage material can be sealed together by sliding the drain ends into each other and affixing them by stapling. This makes a continuous process possible. The drain stitcher must always remain vertically. Any deviation from the vertical position could cause deviations in the drain pattern at greater depth, resulting in disturbances in the consolidation process. A view of marking out, installation and cutting of the PVDs is shown in Figs. 6 to 8.

Fig. 6: The marking out for installation of PVDs

Fig. 7: A view of installation of PVDs

2. The mandrel is pushed into the soil with a power ranging from 50 KN to 200. As soon as the required depth has been reached, it is immediately pulled back up to prevent the soft soil from being pressed into the mandrel. Because of the cohesive working of the soil, the anchor plate to which the drain is attached remains in the soil. 3.

As soon as the bottom of the mandrel is pulled up, the drain is cut off, after which another

60â&#x20AC;&#x192; Volume 45

Number 4 December 2015

Fig. 8: A view of cutting of PVDs after installation

The Bridge and Structural Engineer


6.3 Vibratory loading Vibrating facility is also available in several forms. The vibrator can be placed both centrically and eccentrically on the mandrel. Hydraulic and electric mandrels can be and high â&#x20AC;&#x201C; frequency mandrels, too, are used more and more often. Electric mandrels, however, have the disadvantages that the frequency truing on and off of the mandrel quickly wears out the brake.

of the fill. They provide a simple means of observing the lateral displacement of the failure soil during construction, they could give early warning of an impending bearing failure, even if only monitoring visually. Their rate of movement is likely to be related to a change in pore pressure under the embankment. A more sophisticated alternative to alignment makes

An advantage of the application of vibrators is that harder soil layers can be penetrated. A disadvantage, however, is that the drain can be damaged when the mandrel suddenly drops into a soft layer and the drain roll cannot follow the acceleration. The drain will then be overstressed, which may cause tearing of the filter. This often remains unnoticed as the drain is largely out of high. In addition, the installation capacity is low while insertion costs are high. All this has added to the fact vibrates are used less and less frequently. A new development is the application of a vibrating needle in the tip of a mandrel, by means of which hard layers can be penetrated. The insertion procedure for vibrators is the same as for pushing stitchers.

7. Instrumentation Monitoring

and

Performance

Monitoring of the behavior of embankments on soft ground is essential in order to prevent sudden failures, to recognize changes in the rate of consolidation (e.g., so that the construction schedule can be adjusted accordingly), and to determine or verify design parameters. Performance evaluation will also help to improve settlement predictions and construction efficiencies in future projects. Settlement readings are then taken at regular intervals; say weekly to monthly, where vertical drains are installed. For preloading without drains, settlement measurements would not have to be taken that frequently. Settlement readings are then interpreted. 7.3 Alignment Stakes Alignment stakes are set out parallel to the embankment slope at the beginning of the placement The Bridge and Structural Engineer

Fig. 9: A view of monitoring of progress of consolidation aided with PVDs

stakes would be to measure the lateral deformation in a borehole using inclinometer. A view of the monitoring of progress of consolidation aided with PVDs is shown in Fig. 9.

8.

Laboratory Tests

The properties and operation of the have been tested in independent laboratories, in which the discharge capacity of the drains was determined in both straight and buckled forms. The below tests are carried on the synthetic drains. 1. Compression tests 2. Discharge tests 3. Pore size of filter 4. Permittivity of the filter 5. Chemical resistance At Indian Institute of Technology apparatuses were developed to measure the discharge capacity of PVDs (Venkatappa Rao and Balan, 1997) and the discharge capacity under kinked conditions (Venkatappa Rao et al 2000) (Figs.10 and 11 ). Volume 45 Number 4 December 2015â&#x20AC;&#x192; 61


9.

Developments in India

A PVD with jute burlap as the filter and non-woven coir geotextile as the core was developed by (Venkatappa Rao and Balan, 1997). An equipment was developed to fabricate a braided jute – coir drain- Breco-drain by Venkatappa Rao et al.2000. Fig. 12 depicts such drain. M/s Garware Wall Ropes Ltd and subsequently M/s Tech Fab India have started manufacturing PVDs indigenously.

Fig. 10: Line diagram of discharge capacity measuring apparatus for strip drain (after Venkatappa Rao et al. 2000)

Fig. 12: Breco drain developed at IIT Delhi

10. Use In India PVDs are being extensively used in India beginning long ago at the Kakinada outer harbor project. They have been used at Visakhapatnam airport (Fig.13) and also at many port projects, for eg. at Mundra.

Fig. 13: Ground Improvement adopted at Visakhapatnam Airport (Courtesy : M.Venkataraman)

11. Conclusions Fig. 11: Triaxial cell developed for testing the discharge capacity of drains kinked condition (after Sampat Kumar et al. 2000)

62  Volume 45

Number 4 December 2015

For improving the engineering behaviour of soft clayey soils, particularly on the coastal belt, PVD’s are found to be quite useful, as they increase the The Bridge and Structural Engineer


rate of consolidation and improve the subsoil when subject to pre-loading. They are quite economical too provided time is on our side. But the main advantage is that they do not need any other materials like aggregates to improve the consolidation behavior.

12. References 1. DAS, B.M. (1990), “Principles of Foundation Engineering”, Second Edition, PWS-KENT. 2. HAUSMANN, R. M. (1990), “Engineering Principles of Ground Modifications”, McGrawHill Publishing Company, Singapore. 3. KJELLMAN, W. (1948), “Accelerating consolidation of fine grained soils by means of

The Bridge and Structural Engineer

cardboard wicks”, Proc. 2nd Intl Conf on SM & FE, Rotterdam, 2, 302-305. 4.

RATHMAYER, H and Komulanein, H. (1992). Quality Requirements of prefabricated strip drains, Finn Road Administration Report, No. 22/92, Quality Control and Methds, Helsinki, 62 p.

5.

VENKATAPPA Rao, G. and Balan, K. (1997). "Discharge capacity of natural fibre strip drain using a new drain tester,” Indian Geotechnical Journal, 27 (1), 22-38.

6. Venkatappa Rao, G., Sampat Kumar, J.P. and Banerjee, P.K. (2000), “Characterization of a braided strip drain with coir and jute yarns,: Geotextiles and Geoemembranes, 18, 367-384.

Volume 45 Number 4 December 2015  63


A Case Study on Assessment of Liquefaction Potential for Basal Reinforced Embankment on soft soil Minimol KORULLA VP-Technical Maccaferri Environmental Solutions Pvt. Ltd Gurgaon, India minikorulla@maccaferri-india.com

Minimol Korulla is currently working as the VP- Technical of Maccaferri Environmental Solutions Private Limited, India. She has obtained her bachelor degree in Civil Engineering from Mar Athanasius College, Kothamangalam and Master of Technology in Geotechnical Engineering from College of Engineering Trivandrum. She has worked in Irrigation Department and in Government Engineering College, Kerala in her initial career. In 2001 she joined with Maccaferri Environmental Solutions Private Limited. She is main author of a number of national standards and published papers and she is acknowledged as one of the main authors of IRC 34, IRC113 and IRC-HRB-SR-23.

Anusha NANDAVARAM Design Engineer Maccaferri Environmental Solutions Pvt. Ltd Pune, India anusha@maccaferri-india.com

Anusha Nandavaram graduated with a gold medal in civil engineering from Sree Vidyanikethan Engineering College Tirupati. She is currently working as Engineer at Maccaferri Environmental Solutions Pvt Ltd.

Saurabh CHAURASIA Senior Design Engineer Maccaferri Environmental Solutions Pvt. Ltd Gurgaon, India saurabh@maccaferri-india.com

Saurabh Chaurasia received his Bachelor degree in Civil Engineering from MITS Gwalior in 2010 and Masters Degree in Earthquake Engineering from IIT Roorkee in 2012. He has 2 years of experience as an Assistant Professor. Presently he is working as a Senior Engineer at Maccaferri Environmental Solution Pvt. Ltd.

Meenu P. S. Management trainee Maccaferri Environmental Solutions Pvt. Ltd Pune, India meenu@maccaferri-india.com

Meenu P S received her bachelor degree in civil engineering in 2012 from Govt. Engineering College Thrissur. She obtained Master of Technology in Geotechnical Engineering from the IIT Bombay in 2015. She achieved departmental special mention award for exemplary performance in academics for the year 2014-2015 from IIT Bombay. In July 2015, she joined Maccaferri Environmental Solutions Private Limited.

64â&#x20AC;&#x192; Volume 45

Number 4 December 2015

The Bridge and Structural Engineer


Summary Construction of embankment on fine-grained and soft soils is a challenging task as it is characterized by poor shear strength, high settlement and susceptibility to liquefaction. Unlike other countries, in India liquefaction hazard assessment and its mapping are not very much common. This paper is intended to walk through a case study on liquefaction potential assessment of basal reinforced high embankment on soft soils, in ‘Gangapath’ which is currently under execution in Patna. In addition, various ground improvement measures adopted to diminish the extent of damage of an embankment due to liquefaction, bank protection and erosion control measures adopted has been demonstrated. Furthermore, this paper attempts to capture the construction methodology recommended to reduce the effect of liquefaction and to improve the strength of foundation of high embankment.

the toe may lose shear strength and suffer lateral flow which may result in the settlement of embankment and progressive cracking of the top crest of the embankment. Keeping this in view, liquefaction potential of the deposits at numerous boreholes were determined by using simplified procedure given by Youd et al.[2] which was originally proposed by Seed and Idriss [3] while this paper presents the results of a typical borehole location (refer Table 2).

Keywords: High Embankment, Basal Reinforcement, Liquefaction Potential. (a)

1. Introduction Liquefaction is the phenomenon in which a granular material changes to a liquid state as a result of strong vibration [1]. Due to liquefaction, the structures build on these deposits gets seriously affected and it is more critical in case of water front structures. In this context, this paper aims to present a case study on liquefaction potential assessment of basal reinforced high embankment on soft soils which is currently under execution in Patna, in the state of Bihar. Initial site condition is as depicted in Fig. 1. The aim of the project was to construct a 21,5 Km long elevated road as “Ganga Path or Ganga expressway”, between Digha and Didarganj along southern bank of river Ganga in the city of Patna. In detail, the project consists of construction of a high embankment (8 m to 15m) over loosely packed sand and fine silt deposits which is susceptible to liquefaction when subjected to ground vibration. In addition, Patna lies in seismic zone IV, which is a high-risk zone and is liable to earthquakes. Furthermore, the city lies largely in the Ganga river basin, which gets affected from flood due to the spill over during monsoons. This also triggered the need to provide a high embankment which can act as flood control bund. In addition, in case of an embankment due to liquefaction, the ground beyond The Bridge and Structural Engineer

(b) Fig. 1: (a) and (b): Initial site condition

2.

Geotechnical investigation

In general, the bore log data obtained from the project site indicated a brownish grey clayey silt followed by loose to medium dense silty sand, then followed by dense to very dense silty sand till the termination depth. Also, stiff to very stiff or hard clay was observed in few borehole locations. The results of the Standard Penetration Test (SPT) and other laboratory tests conducted have shown very low shear strength for top soil deposits up to 5m and a Volume 45 Number 4 December 2015  65


medium to stiff consistency for the subsequent layers of soil. Properties of different strata observed during geotechnical investigation are enlisted in Table 1 and Table 2, which indicates a significant amount of non plastic silt and fine sands with low SPT values, along with high water table conditions. Subsequently, this increases the probability of susceptibility to liquefaction in case of a major earthquake. Table 1: Properties of different strata observed during geotechnical investigation Soil stratum Bulk Natural Consistency Particle size Density water Limits (%) characteristics (gm/ cc) content (%) Liquid Plastic (Sand) % (Silt + limit limit clay)% Top soil

1,84

27

39

24

36

64

Stratum I

1,9

30

38

24

16

84

Stratum II

1,98

23

37

20

16

78

Stratum III 1,92

25

38

20

7

76

Stratum IV 1,79

16

32

24

77

23

3. Basal reinforced embankment: Design methodology The inclusion of Geosynthetic reinforcement at the foundation level could enhance the performance of embankment as it resists the shear failure in the embankment as well as in the soft soil. Basal reinforcement stabilizes an embankment over soft soil by preventing lateral spreading of the fill, extrusion of the foundation and rotational failure. While the design of basal reinforced high embankment has been carried out in accordance with IRC 113, IRC 75 and BS 8006 by following the stability checks such as rotational stability of embankment [4], lateral sliding [5], lateral extrusion of foundation [5], bearing capacity [5], plastic squeezing [4], settlement check and assessment of liquefaction potential, the scope of the present paper is limited in step wise evaluation of liquefaction potential for a typical bore hole. 3.1. Liquefaction potential analysis As described in the previous sections, the embankment is lying in high seismic prone area with loose fine-grained subsoil. This triggered the need to check the susceptibility to liquefaction. The assessment of liquefaction potential is carried out based on “Simplified Procedure” methodology [3, 4]. In accordance with the simplified procedure, two variables are required for knowing the susceptibility 66  Volume 45

Number 4 December 2015

of soil for liquefaction, viz., Cyclic Shear Stress (CSR) and Mobilized shear resistance (CRR), where liquefaction potential or seismic resistance of the soil layers expressed in terms of CSR and liquefaction capacity or the capacity of the soil to resist liquefaction expressed in terms of CRR. If the induced Cyclic Shear Stress (CSR) is more than mobilized shear resistance (CRR), liquefaction will occur. In other words, soil is non liquefiable, if the ratio of CRR to CSR, termed as factor of safety is greater than one. 3.1.1 Calculation of Cyclic Stress Ratio (CSR) (1) Where, amax is peak horizontal acceleration at ground surface generated by earthquake; g is acceleration due to gravity; σV0 & σ ’V0 are total and effective vertical overburden stress respectively; rd is stress reduction coefficient (accounts for flexibility of soil), rd = 1.0 – 0.00765z for z < 9,15 m; rd= 1,174 – 0.0267z for 9,15 m < z < 23 m; rd = 0,744 – 0,008z for 23 m < z < 30 m; rd= 0,50 for z > 30m; where z is the depth below ground surface. 3.1.2 Calculation of Cyclic Resistance Ratio (CRR) The liquefaction resistance of soil (CRR) is determined based on Field Test Results. Tests which are used for assessment of liquefaction resistance of soil are: Standard Penetration Test (SPT); Cone Penetration Test (CPT); Shear Wave Velocity (SWVs); & Becker Penetration Test (BPT). Procedure for calculating CRR based on SPT is only discussed here, as it is the most common practice in vogue now a days a)

Calculation of (N1)60 SPT Blow count normalized to an overburden pressure of approximately 100 kPa

(N1 )60=Nm CN CE CB CR CS

(2)

Where, Nm is measured penetration resistance (five different corrections are applied for Nm and are listed below); CN is factor to normalize Nm to reference effective overburden pressure σ’V0 of approximately 100 kpa, CN = (Pa / σ’V0) 0,5 & CN ≤ 1,7; CE is correction factor for hammer energy: CE = (ER/60), CB correction factor for borehole diameter (1 for 65115 mm, 1,05 for 150 mm & 1,15 for 200 mm dia), CR correction factor for rod length, CR = 0,75 for 3,0 m to 4,0 m; CR = 0,85 for 4,0 m to 6,0 m; CR = 0,95 for 6,0 m to 10,0 m; CR = 1,0 for 10,0m to 30,0 m, CS is The Bridge and Structural Engineer


correction for sampler with or without liners (= 1 for sampler & 1,1 to 1,3 for sampler without liner)

3.2 Liquefaction potential analysis by simplified method without overburden and with overburden

b) Calculation of (N1)60cs - SPT Blow Count Normalized to an Equivalent Clean Sand Value.

Typical computation of liquefaction potential without overburden is as depicted in Table 2. From Table 2, it can be observed that the top strata up to 3m deep are liquefiable. Subsequently, ground improvement methods are adopted to mitigate liquefaction, which includes the soil replacement from 0,5m to 3m, construction of overburden at the toe on both sides of the embankment to the required heights ranging from 0,5 to 6m. Even after the replacement, if the soil is found to be liquefiable, other ground improvement methods such as dynamic compaction or Prefabricated Vertical Drains or sand compaction piles as per the soil conditions are recommended.

(N1)60CS= α + β (N1)60

(3)

Where, α = 0 & β = 1,0 for Fine Content (FC) ≤ 5%; α = exp [1,76 – (190/FC2)] & β = [0,99 + FC1,5/1000] 5% ≤ FC ≤ 35%; α = 5,0 & β = 1,2 for FC ≤ 35% c)

Calculation of Cyclic Resistance Ratio CRR7,5 (for earthquake magnitude 7,5) [2]

CRR_7,5= (/(1 ))/(34-(N1)_60 )+ (N1)_60/135+50/ (10(N1)_60+45)^2 -1/200 (4) The above equation is valid for (N1)60 < 30 for (N1)60 ≥ 30. d)

Deciding the Magnitude Scaling Factor (MSF)

The CRR evaluated in step c is applied only to magnitude 7,5 earthquakes. For magnitudes of earthquake other than M = 7,5 EQ. “Magnitude Scaling Factor” have been introduced. The values for N= 5,5; 6; 6,5; 7; 7,5; 8; 8,5 are 2,2; 1,76; 1,44; 1,19; 1,00; 0,84 & 0,72 respectively [2]. e)

Evaluating of CRR

(5)

CRR = MSF x CRR7,5

3.2.1 Parameters considered for the analysis of liquefaction potential Seismic Zone: Zone-IV The peak ground acceleration (PGA), amax /g : 0,13 Maximum earthquake intensity: 7,5 Correction for hammer energy ratio, CE = ER/60, where ER for Rope and pulley system = 60%, Hence CE=60/60 = 1 Borehole diameter = 150mm, Hence CB=1,05 Cs= Correction for standard sampler = 1,0 Magnitude Scaling Factor(MSF) = 1,19

FOS against liquefaction = CRR/CRS. If FOS > 1; soil is non liquefiable. Table 2: Typical Computation of Liquefaction Potential by simplified Method without overburden Depth below E.G.L., m

2.00

3,00

5,10

6,50

8,60

10,00

12,1

Type of strata

SM

SM

SM

SM

SM

SM

SP-SM SP-SM SP-SM

SP-SM

Observed SPT value

2

7

19

27

28

30

32

40

45

47

Saturated density(t/m3)

1,82

1,90

2,0

2,05

2,05

2,05

2,10

2,10

2,10

2,20

Submerged density (t/m3) density(t/m3)density (t/m3) density(t/m3)

0,82

0,90

1,0

1,05

1,05

1,05

1,10

1,10

1,10

1,20

Fine Content (%)

89

88,00 9,00

9,00

90,00

14,00

17,00

11,00

11,0

11,0

Stress reduction coefficient(rd) 0,99

0,98

0,96

0,95

0,93

0,91

0,85

0,81

0,76

0,72

Total overburden pressure (σV0)t/m2

5,43

10,5

13,30

17,70

20,57

24,87

27,81

32,2

35,1

2,61

The Bridge and Structural Engineer

13,50

15,6

17

Volume 45 Number 4 December 2015  67


Effective overburden pressure(σ ’V0)

1,16

2,43

5,49

6,89

9,1

10,57

12,77

14,31

16,6

18,1

Cyclic stress ratio(CSR)

0,19

0,18

0,16

0,16

0,15

0,15

0,14

0,13

0,12

0,12

CN

1,48

1,48

0,93

0,88

0,81

0,78

0,73

0,70

0,66

0,64

CE

1

1

1

1

1

1

1

1

1

1

CB

1,05

1,05

1,05

1,05

1,05

1,05

1,05

1,05

1,05

1,05

CR

0,75

0,75

0,85

0,95

0,95

0,95

1

1

1

1

CS

1,00

1,00

1,00

1,00

1,00

1,00

1,00

1,00

1,00

1,00

SPT Corrected( N1)60

2,33

6,05

15,82

23,72

22,73

23,25

24,52

29,47

31,4

31,7

α

5

5

0,56

0,56

5,00

2,2

3,01

1,21

1,21

1,21

β

1,2

1,2

1,02

1,02

1,2

1,04

1,06

1,03

1,03

1,03

( N1)60cs

7,81

12,26 16,65

24,68

32,28

26,44

29,01

31,46

33,5

33,8

CRRM=7,5

0,09

0,13

0,18

0.29

NA

0.32

0.49

NA

NA

NA

CRR

0.11

0.16

0.21

0.34

NA

0.39

0.50

NA

NA

0.49

FOS

0.6

0.86

>1

>1

>1

>1

>1

>1

>1

>1

Conclusion

L

L

NL

NL

NL

NL

NL

NL

NL

NL

NL: NON-LIQUEFIABLE SOIL; L means LIQUEFIABLE SOIL

Table 3: Typical Computation of Liquefaction Potential by simplified Method with overburden of 0.5m and 2m replacement of top soil Depth below E.G.L., m 2.00

3.00

5.10

6.50

8.60

10.00

12,1

13,50

15,6

17

Type of strata

Replaced SM Soil

SM

SM

SM

SM

SP-SM

SP-SM

SP-SM SP-SM

Observed SPT value

15

7

19

27

28

30

32

40

45

47

Saturated density(t/m3) 2,1

1,90

2,0

2,05

2,05

2,05

2,10

2,10

2,10

2,20

Submerged density (t/m3)

1,1

0,90

1,0

1,05

1,05

1,05

1,10

1,10

1,10

1,20

Fine Content (%)

8,00

88,00

9,00

9,00

90,00

14,00

17,00

11,00

11,0

11,0

Stress reduction coefficient (rd)

0,98

0,98

0,96

0,95

0,93

0,91

0,85

0,81

0,76

0,72

Total overburden pressure (σV0)t/m2

4,5

6,6

10,5

13,30

17,70

20,57

24,87

27,81

32,2

35,1

Effective overburden pressure (σ ’V0)

2,5

3,6

5,49

6,89

9,1

10,57

12,77

14,31

16,6

18,1

Cyclic stress ratio (CSR)

0,15

0,15

0,16

0,16

0,15

0,15

0,14

0,13

0,12

0,12

CN

1,41

1,15

0,93

0,88

0,81

0,78

0,73

0,70

0,66

0,64

CE

1

1

1

1

1

1

1

1

1

1

CB

1,05

1,05

1,05

1,05

1,05

1,05

1,05

1,05

1,05

1,05

CR

0,75

0,80

0,85

0,95

0,95

0,95

1

1

1

1

CS

1,00

1,00

1,00

1,00

1,00

1,00

1,00

1,00

1,00

1,00

68  Volume 45

Number 4 December 2015

The Bridge and Structural Engineer


SPT Corrected( N1)60

16,71

6,79

15,82

23,72

22,73

23,25

24,52

29,47

31,4

31,7

α

0,3

5

0,56

0,56

5,00

2,2

3,01

1,21

1,21

1,21

β

1,01

1,2

1,02

1,02

1,2

1,04

1,06

1,03

1,03

1,03

( N1)60cs

17,21

13,15

16,65

24,68

32,28

26,44

29,01

31,46

33,5

33,8

CRRM=7,5

0,18

0,14

0,18

0,29

NA

0,32

0,49

NA

NA

NA

CRR

0,22

0,17

0,21

0,34

NA

0,39

0,50

NA

NA

0,49

FOS

>1

>1

>1

>1

>1

>1

>1

>1

>1

>1

Conclusion

NL

NL

NL

NL

NL

NL

NL

NL

NL

NL

NL: NON-LIQUEFIABLE SOIL; L means LIQUEFIABLE SOIL

3.3 Geometry of the embankment The typical cross sections of the embankment are as depicted in Fig. 2 and Fig.3, which indicates at few locations a single embankment with a top width of 40,5 m as a main expressway and at other locations a main embankment having a top width of 26,5 m as main expressway along with two additional small embankments with top width 12 m as service roads is provided. Height of the embankment is varying from 8 m to 15 m with a side slope 2H: 1V with berm at a height of HFL+1,5 m and width 2,5 m, 2m on river side and city side respectively.

3.4 Ground improvement and bank protection works In order to improve the strength of foundation soil, basal reinforcement using high strength geogrids, which is termed as ‘paralink’ with strength varying from 100 to 1300KN/m is adopted. In addition, as the expressway alignment is running along the flow of the river Ganga, there are possibilities of erosion of the proposed road embankment which necessitate The Bridge and Structural Engineer

the need to provide suitable erosion control and bank protection works. Keeping this in view, for embankment protection a launching apron of length 20 m with 0,5 m thick gabion mattress and on slopes 0,23 m (till HFL) or 0,17 m (from HFL to berm level) thick renomattress was suggested. Further, to prevent erosion along the slope biodegradable mat which is termed as ‘Biomac’ was suggested above the berm. Volume 45 Number 4 December 2015  69


In addition, to prevent erosion of fines beneath the mattresses, a geotextile filter/separator, MacTex N60,1 was also recommended. Furthermore, additional ground improvement methods are suggested where the soil was susceptible to liquefaction. The extent of damage created by liquefaction can be controlled by providing berms at the toe of the embankment or increasing the berm width or by densification of the subsoil by dynamic compaction, so that N value is higher than the liquefiable limits [4]. 3.5 Construction methodology Outline of the construction methodology for various ground improvement methods such as removal and replacement of the poor strata along with basal reinforcement, construction of a berm as an additional surcharge at toe of the embankment, dynamic compaction is described in the following section. 3.5.1 Ground improvement by removal and replacement along with basal reinforcement In this method, setting out was done prior to the excavation and site clearance. After the completion of site clearance, limits of embankment or excavation were marked by fixing pegs on both sides at regular intervals. The width and levels of excavation were calculated at every 10m interval as per the cross sections. Typical cross sections of the embankment are depicted in Fig. 2 and Fig. 3. Further, the excavation was carried out to remove the existing soft soil of 2 to 3 m depth from ground level and a geotextile layer, termed as ‘MacTex N30,2’ was spread over the compacted and levelled surface which serves the function of separation. In addition, the first layer of embankment over the excavated surface was constructed in accordance with the specifications given in Clause 305,4,5 and 305,4,6 of Ministry of Road Transport & Highways (MoRTH), which describes the construction of embankment over ground incapable of supporting construction equipment and embankment construction under water and water logged areas respectively. Subsequent layers of embankment was constructed as per the specifications given in clause 305,2 with 250 mm loose thick layer and compacted in accordance with clause 305,3,5 and 305,3,6 of MoRTH, which describes specifications for spreading material in layers and bringing to appropriate moisture content and for compaction respectively. A layer of gravel/sand was 70  Volume 45

Number 4 December 2015

laid as drainage medium and paralink was laid as per approved design and drawings. Once the paralink is laid, rest of the embankment was constructed in layers as per the MoRTH specifications. 3.5.2 Improvement by construction of berm at toe of embankment to counteract liquefaction potential below the toe of the embankment In order to arrest the liquefaction effect, a berm with surcharge height (as per the liquefaction analysis) was constructed at the toe of embankment on both sides. Further, after the completion of construction of berm up to required height, Standard Penetration Test was carried out at toe portion and improved SPT was determined to check the susceptibility to liquefaction. Subsequently, if the ground layers are found nonLiquefiable, the extra berm constructed at toe on both sides was removed without disturbing the slope of the highway embankment or else the berm is retained for a month or two and again SPT was performed. Period of overburden to be maintained was theoretically determined from rate of consolidation. 3.5.3 Improvement by Dynamic Compaction Dynamic compaction is a method of ground improvement that results from the application of high levels of energy at the ground surface [4]. Existing ground levels is recorded before the start of dynamic compaction. Further, grid layout is set out on the prepared ground with a spacing of 4m between grids for a stretch length of 200 m and approximate width with average toe of 104 m. Track/Crawler mounted crane of 75/100MT fixed with a tamper of weight 25 tons was moved on the ground adjacent to the grid point to start the primary pass. For the first pass, initial tamping was started with 15 drops on each alternative drop point. After completion of primary pass, secondary pass was done in between primary pass drop points. During tamping process, once the crater depth has reached about 1m, the crater was filled with approved granular material before additional drops were performed. After completion of dynamic compaction, compacted area was further compacted by a vibratory roller. SPT is carried out to assess the susceptibility of ground layers for liquefaction at 100m interval. With the above testing, if the ground layers are found non-Liquefiable, tamping process will be discontinued and further stage of ground improvement as detailed above will be commenced The Bridge and Structural Engineer


else dynamic compaction shall be repeated for more number of passes as per the site conditions.

4. Conclusions Assessment of liquefaction potential and the necessary remedial measures adopted to diminish its effect such as removal and replacement of the soft soil, provision of berm at the toe of the embankment or increasing the berm width, dynamic compaction plays a major role, before the construction and during the service life of an embankment. In this paper, an effort was made to assess the liquefaction potential of the soil based on the data collected about the seismic history of the site and the results obtained from the geotechnical investigations conducted during the period of this study. The analysis performed is found to be fairly good enough to assess the liquefaction potential. In addition, the effectiveness of basal reinforcement with high tensile strength geogrid to stabilize an embankment over soft soil and various ground improvement techniques to eliminate liquefaction were also clearly demonstrated.

5. References 1.

IDRISS I. M., and BOULANGER R.W., (2008), “Soil liquefaction during earthquakes”, Earthquake Engineering Research Institute, p. 264.

The Bridge and Structural Engineer

2. YOUD, T.L. et al. (2001), ‘Liquefaction resistance of soils : summary report from the 1996 NCEER and 1998 NCEER/NSF Workshops on Evaluation of Liquefaction resistance of soils’, ASCE Journal of Geotechnical and Geoenvironmental Engineering., Vol.127, No. 10, pp.817-833. 3.

SEED H.B. and IDRISS, I.M. (1971), "Simplified Procedure for Evaluating Soil Liquefaction Potential", ASCE Journal of Geotechnical Engineering, Vol.97, No.9, pp. 1249-1273.

4.

IRC 75. (2015), “Guidelines for the Design of High Embankments”, Indian Road Congress, Kama Koti Marg, Sector 6, R. K. Puram, New Delhi.

5.

IRC 113. (2013), “Guidelines for the Design and Construction of Geosynthetic Reinforced Embankments on Soft Subsoils”, Indian Road Congress, Kama Koti Marg, Sector 6, R. K. Puram, New Delhi.

6.

BS 8006 Part 1. (2010), “Code of Practice for strengthened/reinforced soils and other fills”, British Standard Institution Group Headquarters, London, UK.

Volume 45 Number 4 December 2015  71


GEOCELL BASAL MATTRESS FOR HIGH EMBANKMENTS AND RETAINING WALLS

G. Venkatappa RAO (Former Professor & Head Department of Civil engineering, IIT Delhi) Chairman, Sami Master Geoenvironmental Services (P) Ltd, Hyderabad gvrao.19@gmail.com

S. Jaswant KUMAR (Former, Chief Engineer, Department of Roads and Buildings Department, Government of Andhra Pradesh, Hyderabad) Vice-President, Arwee Associates, Hyderabad

Dr. G.Venkatappa Rao graduated in Civil Engineering from BITS, Pilani in 1965. He obtained M.E. and Ph.D. in Civil / Geotechincal Engineering from IISc.,Bangalore in 1968 and 1973 respectively. From 1972 to 1975, he worked as Lecturer in Civil Engineering in BITS, Pilani. In 1975, he joined IIT, Delhi as a Lecturer and became Professor in the year 1985 and served in various capacities including as Head, Department of Civil Engineering and Dean (Student Affairs). His teaching and research interests include Soil and Rock Mechanics, Marine Geotechnology, Ground Improvement, Transportation Engineering and Geosynthetics. He successfully guided 24 Ph.D. theses and over 100 M.Tech dissertations. He received several awards and served several professional bodies in various capacities. His was the only name from India included in IFAI’s 142 Names to know in Geo’ (Geotechnical Fabrics Report). He founded SAGES and STTRI with the objective of dissemination of knowledge for a better society.

S. Jaswant Kumar graduated in Civil Engineering from College of Engineering, Osmania University, Hyderabad in 1976. He obtained M.E in Geotechnical engineering from Osmania University in 1992. He joined PWD of Government of Andhra Pradesh in 1976 and later in Roads & Buildings Department in 1982 in Design & Planning and rose to the level of Chief Engineer. He also worked as Regional Officer & Chief General Manager, Planning & Quality in NHAI for 21/2 years and associated with quality/construction of various 4laning/6laning Projects in the country. Presently associated with Highway Project Supervision as IE/Authority, contract Specialist. Associated with latest ground improvement technologies, usage of Geosynthetics in problematic soils in the State of Andhra Pradesh with technical guidance of NH Delhi.

Summary Nearly half of Indian subcontinent consists of black cotton soils occurring as residual soils in central India and transported soils in the coastal belt. Construction of embankments and retaining walls often pose problems for engineers over these weak soft soils. Over the last twenty years, geosynthetic reinforced structures have been constructed in many parts of Andhra Pradesh in south India either in the form of Geocell for foundation improvement of high embankments and Geogrid 72  Volume 45

Number 4 December 2015

reinforced soil walls. This paper delineates the design principles and details two successful case studies.

1.

Introduction

1.1 General Construction of embankments over weak, relatively thin foundation soils often poses problems for designers and contractors. Conventional methods of soil excavation and replacement or piling are not always practical or The Bridge and Structural Engineer


economic. Geocells containing a series of interlocking cells, constructed from polymer geogrid reinforcement are a convenient alternative. These contain and confine the granular layer at the base of an embankment as shown in Fig.1.

It provides a stiff platform to ensure an even distribution of load onto the foundation.

It provides a stiff platform to ensure the formation of a regular stress field within the soft foundation layer.

2.

Design

2.1 Basis for Design Fig.1: Basal mattress reinforcement

1.2 Concept Geocell is a honeycombed structure formed from a series of interlocking cells (Fig. 2). These cells are fabricated directly on the soft foundation soil from grid reinforcement and are filled with granular material. The resulting structure is usually 1 m deep. This arrangement provides a stiff platform which enables the geocell mattress to exert a degree of restraining influence on the deformation mechanism developing in the soft foundation soil. This restraining influence effectively rotates the principal stress direction from a near vertical or outward direction in the embankment fill to up to 45° inclined inward direction at the top of the foundation soil. This rotation of the stress field alters the potential slip mechanisms in the foundation soil involving slip planes throughout the width and depth of the layer and hence an enhanced bearing capacity is achieved.

Fig.2: Arrangement of geocell

1.3 Function of Geocell Mattress Providing geocell foundation mattress at the base of an embankment foundation performs the following functions: It provides a perfectly rough interface between the soft foundations.

It contains the granular fill of the mattress.

The Bridge and Structural Engineer

The stability of an embankment constructed on soft soil is governed mostly by the shearing resistance of the foundation, and the construction of an embankment on soft soil is a problem of bearing capacity. Reinforcement may be placed at foundation level to prevent shear failure both in the embankment fill and in the foundation soil, any reduction in differential settlement is of secondary importance. A geocell mattress supports the increasing overburden from the construction of an embankment while the soft soil below becomes critical and undergoes plastic deformation. The ultimate bearing capacity for this condition of soil layer was analyzed by methods of plasticity used in the pressing of metals. Johnson and Mellor developed slip line fields to analyze the compression of a block between rough, rigid plates of finite width. Bush et al (1990) describe the design and construction procedures in detail. An important consideration is that stability of an embankment on soft soil is most critical during construction. This is because the relatively low permeability of the soft foundation does not permit full consolidation in the normal time scale of construction. At the end of construction the embankment loading has been applied, but the gain in shearing resistance of the foundation due to consolidation may be insufficient for stability. Once consolidation has occurred, the resulting improvement in shearing resistance in the foundation will obviate the need for the reinforcement to improve stability. Thus during the period between the end of construction and consolidation of the foundation the fundamental strength requirement of the reinforcement is that at any instant in time the factored reinforcement design strength must be equal or exceed the design load. Geocell mattress stabilizes an embankment over soft ground by preventing lateral spreading of the fill, extrusion of the foundation and overall rotational failure. Volume 45 Number 4 December 2015  73


This stabilizing force is generated in the mattress by shear stresses transmitted from the foundation soil and fill which place the reinforcement in tension. 2.2 Design Steps Using Limit State The ultimate limit states which must be considered are as follows (Fig. 3): (BS 8006: 1995)

Foundation extrusion stability

Overall stability

The serviceability limit states which must be considered are: Excessive strain in the mattress

Settlement of foundation

The geocell mattress can be analyzed using the above procedure, but a method based on slip line fields for the analysis of foundation stability is generally used.

3.

High Embankment for Bridge Over Vasista River

3.1 Sub-soil Conditions The maximum embankment height in the bridge approach for a bridge across Vasista branch of the Godavari river is 12 m with a longitudinal gradient of 1 in 35. The embankment is to be formed on weak black cotton soil extending up to 30 m depth. The foundation soil typical profile, shown in Fig. 4, is highly plastic and swelling type with very poor bearing capacity.

Fig. 4: Sub soil properties: Vasista bridge approaches

3.2 Design Alternatives Fig. 3: Ultimate limit states for basal reinforced embankments (after BS 8006: 1995)

Local stability of the embankment fill

Rotational stability of the embankment

Lateral sliding stability of the embankment fill

74  Volume 45

Number 4 December 2015

i)

The excavation of soft material and its replacement with an imported suitable rockfill material.

ii) The partial excavation of the soft material and the displacement of the remainder by deposition of suitable backfill until finally stable. The Bridge and Structural Engineer


iii) Drainage of the soft material, together with a controlled filling of the embankment to permit dissipation of pore pressure and thus resultant gain in strength of the weak layer. iv) The construction of high strength geogrid mattress to carry the embankment over the week ground, with the granular fill of the mattress acting as a drainage blanket. Soil replacement is not feasible due to non-availability of inexpensive material as well as large thickness of soil to be replaced. The conventional pre-loading with sand drains or prefabricated vertical drains is ineffective and time consuming because of the swelling nature of the soil. After studying various alternatives it was decided to adopt a Geocell (Basal) mattress foundation' as a measure to improve the bearing capacity of the weak soil.

A bi-oriented geogrid (BG with peak tensile strength of 20 kN/m) is placed over the geotextile. A monooriented geogrid (MG with peak tensile strength of 60 kN/m) is then laid transversely across and one edge is stitched to the base using polypropylene rods (Fig. 7). A second mono-oriented grid is laid transversely at a spacing of 1m. The procedure is continued and after required transverse grids are stitched in place they are rotated by 90° about the stitched edge in to a vertical position and temporarily tensioned by timber posts. The cell structure is then formed by unrolling another roll of mono-oriented geogrids between transverse diaphragms connecting them with rods. Thus a geocell is formed. After number of cells are formed, they are gradually filled with the fill, i.e., graded metal, i.e., 40

3.3 Construction Aspects The embankment runs for about 350 m length on each side, i.e., West Godavari side and East Godavari side of the bridge across Vasista river. The typical cross-section is shown at Fig. 5. The height of embankment varies from 2 m to 12 m. The proposal consists of: Construction of geocell mattress of 1.0 m thick with end restrainment from 8 m to 12 m height of embankment.

Fig. 5: typical cross section of embankment.

Construction of geocell mattress of 0.50 m thick for 5 m to 8 m height embankment.

Raising embankment in stages.

Construction of reinforced soil retaining wall at toe (due to restriction of base width).

The final design adopted is depicted in Figs. 5 and 6. The design carried out using BS 8006 – 1995. 3.4 Basal Mattress

Fig. 6: Sectional Elevation view of Elevation of geocell mattress

The basal mattress is to be placed at 0.50 m below the ground level. Before laying, the top soil, i.e, loose and slushy material is removed. As the groundwater level is very high a non-woven needle punched geotextile (GT) with peak tensile strength of 13.5 kN/m is placed over the soil duly ensuring no folds in it. A 4 m wide end restrain is to be provided as per design to stiffen the ends of geocell mattress. The end restraint is also a basal mattress but the bottom of end restraint is 1 m below the principal mattress under embankment.

to 75 mm. The filling commenced from one boundary. The cells are filled in this manner to avoid distortion of the mattress, i.e., no cell is filled to full height before the adjacent cell is at least half filled (Fig. 8). As it may be Fig.6 Sectional Elevation view of Elevation of geocell mattress impractical to compact the fill in the cells, overfilling by 100 to 150 mm is done to cater for compaction settlements and to allow for construction plant to operate on the mattresses without damaging diaphragms.

The Bridge and Structural Engineer

Volume 45 Number 4 December 2015  75


Fig. 7a: Fabrication of basal foundation mattress

vibratory rollers. The embankment is raised by first filling in the centre of cross-section and then extending to ends. After attaining 4 to 5 m height of embankment it was allowed to settle for 15 days and then the end reinforced soil retaining walls using concrete facia and geogrid reinforcement (tensile strength 45 kN/m) were taken up. The height of wall is 3 m and after laying backfill behind wall, the full embankment is raised. The top 4 m height is raised in 4 months with observation period of 1 month after 1 m raising. The construction is done in phased manner to allow initial settlement to occur during construction period.

Fig. 7b: Typical plan view of basal Mattress

The mattress formation is commenced from 5 m height embankment and at a time about 30 to 35 m length of mattress is tackled. The base widths are varying from 21 m to 58 m. After completing the principal mattress the end restrain is taken up. In a similar manner to the principal mattress the end restrain is formed. A 4 m x 1 m deep excavation is done and a geotextile is placed on the soil. 1 m thick geocell is formed as noted above and the top of this cell is to coincide with the bottom of the principal cell. Once this additional cell is over the principal cell is extended over the ends. The top and bottom cells are jointed with Polypropylene rods. The strength of mono-oriented geogrid used for end restrain is 100 kN/m which is higher than the grid used for principal mattress, i.e., 60 kN/m. After filling the geocells with boulders another layer of non woven geotextile is placed over the fill to act as a separator between embankment fill soil and stone fill in the geocell mattress. The embankment is raised in 2 m intervals in layers with each layer of 200 to 300 mm thick. The fill used is sand and silty sand. The general construction procedure followed is as per Specifications of Roads and Bridges, MOST (1995). The compaction is achieved by using 76â&#x20AC;&#x192; Volume 45

Number 4 December 2015

Fig. 8: Technique of filling

3.5 Comments Normally in order to increase the rate of settlement of the alluvial material below the embankment and prevent the build up of excess pore pressure during its construction, the geogrid mattress option includes the provision for the installation of vertical band drains driven through the mattress into the alluvial material below.

ď Ź

The Bridge and Structural Engineer


Geocell mattress can be constructed over the existing ground without removal of top soil or vegetation and any large obstruction such as boulders and the like.

4. Geosynthetic Reinforced Soil Wall at Vijayawada

In order to maintain the rigidity of the mattress during construction and infilling, a large degree of tensioning of the transverse vertical diaphragms is necessary (as shown in Fig. 7 b)

A road portion in Vijayawada Municipal Limits of NH-9 in Andhra Pradesh collapsed due the failure of the retaining the road width in this extremely busy commercial area got reduced and called for immediate restoration.

3.6 Advantages and Disadvantages of Using a Geogrid Mattress No excavation and minimal site preparation.

No trips required for disposal of unsuitable excavated material.

No difficulty working at or below the water table.

Less prone to delays due to inclement whether.

4.1 History

4.2 Sub-soil profile Four numbers of boreholes were drilled in the canal bed up to a depth of 15 m. The geotechnical investigations revealed that the soil profile was varying. The soil in the top 2 m was clay (CH or CI type) and was very soft. Beyond this depth, the soil was generally CI and sometimes CH, but in between there were alternate layers of sand (SM). In one of the bore holes, the soil appeared to be improving with depth and the field dry unit weight was 13.73 kN/m3 and the value of undrained cohesion was 30 kPa, at a depth of around 6 m. Even below a depth of 9 m, up to 15 m depth the soil was soft clay. 4.3 Design Philosophy

Fig; 9: Sectional view of geosynthetics reinforced soil wall with basal mattress.

No difficulties due having to excavate unsuitable material adjacent to existing structures, embankments etc.

Some initial training to familiarise the workforce with construction procedures is required.

Labour intensive if high rate of progress to be maintained.

Supervision and strict pattern of working required.

The Bridge and Structural Engineer

The road carriageway level was +21.860 m. The full supply level of the Canal was +17.220 m and the canal bed level was +12.360 m, hence the required height of the retaining structure above the canal bed level was 8.5 m. With a wall height of 8.5 m above the canal bed level, and a 2 m thick soft soil near the canal bed, a gravity wall was not feasible. Pile foundations need to be at least 15 to 20 m long and hence were likely to be prohibitively expensive. Also previous construction using underreamed pile foundations was not encouraging. In addition, the road was located in a busy commercial area; there was little space available for conventional construction. Further the major work of foundation and as well as wall up to full supply level was necessarily to be constructed during the 30 days closure of the canal. Keeping the above in view, the best solution thought was to be geosynthetics reinforced soil wall with stone filled basal mattress. 4.4 Design The detailed design has been carried out using BS 8006 (1995) and the general principles of geotechnical engineering. A sectional view of geosynthetic reinforced Volume 45 Number 4 December 2015  77


soil wall with basal mattress is shown in Fig. 9. It consisted of 1 m thick basal mattress formed 1.4 m below the bed level of the canal. Over this the first tier of the wall was 6.5 m high the top of which was 1 m above the FSL of the canal. The second tier was then constructed with an offset of 1.5 m and was 4.5 m high. The properties of the geosynthetics used are given in Table- 1.The overall length of the wall constructed was 400 m. Table 1: Strength Properties of Geosynthetics used Sl. No.

Nomenclature

1 2 3 4 5 6 7

GG1 GG2 GG3 GG4 GG5 BG1 GT

Peak Tensile Strength (kN/m) 45 60 90 120 160 20/20 800 gsm

Yield point Elongation (%) 13 13 13 13 13 13/10 800 gsm

Indian Institute of Technology, Delhi (while the senior author was serving) as consultancy assignments by the Department of Roads and Buildings, Government of Andhra Pradesh. Grateful thanks are due to Mr. P. V. Rama Raju, the then Engineer-in-Chief, for the confidence reposed in undertaking such a work for the first time in India. Dr. Rakesh Kumar Datta, then Research Scholar, Department of Civil Engineering, IIT Delhi, currently, Professor, Department of Civil Engineering, NIT, Hamirpur, has carried out the design detailing. Construction has been ably carried out by M/s P. V. Raj & Co. and M/s Sai Constructions.

7. References 1.

BS 8006 : 1995. British Standard Code of Practice for Strengthened/Reinforced Soils and other fills.

2.

Bush, D.I. Jenner, C.J. and Bassett, R.H. (1990). “The design and construction of geocell foundation mattress supporting embankments over soft ground,” Geotextiles and Geomembranes, Vol.9, No.1, pp.83-98.

3.

Bush, D.I., Jenner, C.J. and Bassett, R.H. (1990). "The design and construction of geocell foundation mattresses supporting embankments over soft ground", Geotextiles and Geomembranes, Vol. 9, No. 1, pp. 83-98.

5. Conclusions 5.1 The 700 m long embankment of 12 m maximum height using geocell mattress was constructed in one year time on a weak soil whose bearing capacity is hardly 50 kN/sqm. The basal mattress is very effective in mobilizing the maximum shear strength and in increasing the bearing capacity of soft foundation. The construction was completed in 2001 and the bridge was opened to traffic in 2003. The settlement that occurred was nearly 90% of the estimated value of 1.4 m. Tsolution adopted was about 2.5 times cheaper then the only feasible alternative of pile foundation. 5.2 Based on the above design all the activities are preplanned such that work of excavation and laying the mattress can start immediately after closer of the canal. The construction of a 200 m stretch was successfully completed during 2002, despite difficulties experienced in making a deep cut next to a busy road, and the water level in the canal remaining at least 1 33m above the bed level. The remaining 200 m were subsequently completed next year and the road opened to traffic. Comparing the costs with the gabion faced walls Geosynthetics retaining system was found to be more economical.

6. Acknowledgements The work described in this paper was awarded to 78  Volume 45

Number 4 December 2015

4. Edgar, S. (1984). The use of a high tensile polymer grid mattress on the Musselburgh and Portobello bypass", Proc. Conf. on Polymer Grid Reinforcement, London, , pp. 103-111. 5.

IIT, Delhi (1999). "Construction of geocell (basal) mattress foundation and geosynthetic reinforced soil wall for high embankment approaches of Bridge across Vasista branch of river Godavari near Chinchinada, West Godavari district, Andhra Pradesh" - Consultancy Report.

7.

IIT, Delhi (1999). "Instrumentation for deformation measurements in Approaches to Bridge at Chinchinada and Palacole ROB" - Consultancy Report.

8.

MOST (1995). Specifications for Roads and Bridges Works, Indian Roads Congress, New Delhi. The Bridge and Structural Engineer


IMPROVEMENT OF SOFT CLAY FOUNDATION BED FOR EMBANKMENTS USING GEOCELL - AN EXPERIMENTAL STUDY

Sefali BISWAS Ph.D. Student in Civil Engineering, Jadavpur University, Kolkata, INDIA sefali.india@gmail.com

R. B. SAHU Professor & Head of Civil Engineering, Jadavpur University, Kolkata, INDIA rbsahu_1963@yahoo.co.in

Sefali, born in 1985, received her civil engineering degree from Jadavpur University, Kolkata, India.

Sahu, born in 1963, received his civil engineering degree from Jadavpur University, Kolkata, India.

Satyendra MITTAL Associate Professor of Civil Engineering, IIT Roorkee, INDIA satyendramittal@gmail.com

G. BHANDARI Associate Professor of Civil Engineering, Jadavpur University, Kolkata, INDIA bhandari.ju@gmail.com

Mittal, born in 1959, received his civil engineering degree from University of Allahabad, India.

Bhandari, born in 1968, received his civil engineering degree from BE College (Presently IIEST), Shibpore, Howrah, India.

Abstract Construction of embankment over soft clay bed causes problem due to bearing capacity failures and higher compressibility. Soil reinforcement using geocells has been utilized in many areas of geotechnical engineering, for example, highway/ railway embankment, pavement, slope protection etc. through lateral confinement of soil. In the present study the use of geocell reinforcement on the performance of embankments over very soft soil foundation clay The Bridge and Structural Engineer

bed has been evaluated through a series of laboratory model tests. Model embankment 300 mm high was constructed over a layer of geocell, formed using geogrids and pocket size 50 x 50 mm, 75 x 75 mm and 100 x 100 mm, on top of a 550mm thick very soft foundation bed in a steel test tank. Surcharge pressure was applied in increments on the top of the embankment and pressure-settlement behavior of the embankment was monitored continuously until the failure was reached. It was observed that improvement Volume 45 Number 4 December 2015â&#x20AC;&#x192; 79


factor decreases with the incerase in pocket size of geocell and also with the increase in deformation. Further, a semi-emperical logarithm model has been proposed for prediction of deformation characteristics of embankment reinforced with geocell. Keywords: Embankment; Geocell; Reinforcement; Bearing Capacity; Improvement Factor

Soft Soil; Settlement;

1. Introduction A technique of ground improvement using geosynthetics was initiated in early eighties of twentieth Century. Since then polymeric fabrics ( i.e., geotextiles) and geogrids are being extensively used all over the world in foundations, walls, roads and embankments in order to improve their performance through interface friction and membrane effect. Subsequently during early nineties advancement in this field was started through use of geocell by providing three dimensional confinements to the soil. Geocell is a series of interconnected cells constructed from geogrid of polymeric strip which contains and confines the soil effectively. Bathurst and Jarret [1] first indicated a guideline for use of geocell reinforcement in construction of pavements over soft peat soils on the basis large scale model tests over peat subgrade. Dash et al. [3, 4 and 5], Sitharam et al. [19], Mittal [17], Yoon et al. [22], Zhou and Wen [23], Sitharam and Hegde [21] conducted model footing tests on geocell reinforced soil and found that interconnected geocells increases load carrying capacity and reduces settlement significantly through increased rigidity of geocell layer by confinement of foundation soil. A number of case studies on use of geocell reinforced soil have also been reported by various researchers (Cowland and Wong [2]; Hendricker et al. [8]; Dash et al. [6] ). Effect of various parameters, like, height, width, pocket size, pattern of formation, length and position have been studied extensively by various investigator (Shimizu and Inui [20]; Mhaiskar and Mandal [15]; Mandal and Gupta [14]; Krishnaswamy et al. [10]; Dash et al. [5]; Madhavi Latha et al. [12]; Madhavi Latha G [13]; Madhavi Latha et al. [11]; Pokharel et al. [18] ). Construction of highway/railway embankments, containment dykes, flood protection levees etc. over soft clay are challenging for engineers as they cause very often bearing capacity failure and excessive 80  Volume 45

Number 4 December 2015

settlement. Madhavi Latha et al. [12], Sitharam et al. [21] have reported that geocell reinforcement helps in improving its overall performance by reducing deformations of the earth embankment over soft clay foundation. With the help of a series of model tests Madhavi Latha et al. [12] have shown that geocell reinforcement is advantageous in improving bearing capacity and reducing settlement of the embankment on the basis of a series of model tests with embankment height 400 mm and geocell of height 250 mm, pocket size 400 mm and 200 mm. This gives a qualitative picture of the improvement of geocell reinforced ground, but hardly gives any quantitative picture of the improvement. With this background, the present paper highlights the findings of a series of model tests on geocell reinforced embankment prepared in the laboratory with a scale factor of 10. The results of the model tests would give to some extent an idea of quantitive improvement for the prototype cases when transformed by scale factor amount.

2.

Model test - Scale effect

The geometry of the soft foundation clay bed – geocell – embankment system were simulated using dimensionless parameters (Görtler [8] ) as given in Table 1. Table 1: Dimensionless parameters

Dimensionless parameter* Geometry λL =Lp / LM Stress λσ =σp / σM Displacement λS =Sp / SM

* P = Prototype and M = Model The scale factor assumed in the present analysis is 10 ( λ = 10) due to restriction in the availability of the experimental facilities, i.e., size of the test tank, in the laboratory. Accordingly the sizes of soft foundation clay bed–geocell-embankment system considered in the analysis are given in Table 2. Table 2: Dimension of different components Model

Prototype

Height of Thickness of Height of Thickness of embankment foundation embankment foundation bed (Hm) bed (Fm) (Hp) (Fp) 300 mm

550 mm

Height of geocell = 25 mm

3.0 m

5.50 m 250 mm

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3.

Parameters Studied

Tests were conducted on model embankment resting over soft clay bed without and with geocell at the interface of embankment and soft clay bed. Diamond pattern geocell layer, height 25 mm and truncated at toe was filled with embankment soil.

used for forming the joints to make geocells as shown in Fig.1b.

 Height of the embankments - 300 mm  Pocket opening size of geocells - (50x50) mm, (75x75) mm and (100x100) mm  Aspect ratio of geocells - h/D = 0.25, 0.33 and 0.50

4. Materials used Fig.1a: Geogrid used in study

Embankment and foundation soil used in the present investigation were collected locally. The soil used in the embankment was brownish grey silty clay / clayey silt collected from the top layer of normal Kolkata deposit while the soil used in the foundation was dark grey silty clay with decomposed wood / organic matter collected from layer II of normal Kolkata deposit. The various properties of soil used are presented in Table 3. In accordance with the Unified soil classification system, the embankment soil can be classified as inorganic clay of low to medium plasticity with symbol CL and the foundation soil can be classified as organic clays with high plasticity with symbol CH (Mittal [16] ).

Fig.1b: Geocell used in study

Table 4: Properties of Geogrid

Table 3: Properties of the embankment soil Property Liquid limit (%) Plastic limit (%) Optimum Moisture content (%) Max Dry Density(kN/m3) Mean Specific Gravity(G) Classification

Embankment soil 44 25 17.2

Foundation soil

15.4

10.13

2.65

1.56

CL

CH

Ultimate tensile strength Failure strain Initial modulus Secant modulus (at 5% strain) Secant modulus (at 10% strain) Aperture size

56 26 24.91

4.1 Reinforcement used in experimental study The geocell were made by using planar geogrid by cutting from full rolls to the required length and breadth (Fig.1a).The properties of planar geogrid are shown in Table 5. After placing them in transverse and diagonal directions, thin mild steel wire were The Bridge and Structural Engineer

5.

20 kN/m 25 % 32 kN/m 28 22 25 mm x 25 mm

Test set-up

The test set-up was prepared in a transparent (Perspex-fitted) steel tank. Geocell layer was placed on the soft foundation clay bed within the tank and the embankment was constructed above the geocell mattress. Fig. 2 represents the sectional elevation and test set-up of the model embankment including foundation clay bed used in the present investigation. Heights of model embankment over foundation clay bed was 300 mm and thickness of foundation clay bed Volume 45 Number 4 December 2015  81


was 550 mm. Side slope of embankment was 1H:1V, with a moisture content of 12.5%.

5.3 Laying of geocell mattress and preparation of embankment After leveling the soft soil bed, a layer of geocells/ geogrid was laid at the interface of the soft foundation and embankment (Fig. 3). Geocell pockets were filled with embankment soil and were compacted uniformly using a tamping rod of 20 mm diameter.

Fig. 3: Laying of geocell mattress Fig. 2: Sectional elevation and test set-up of model embankment

5.1 Equipments used in the test set-up A steel tank of dimensions 1150 mm Ń&#x2026; 550 mm Ń&#x2026; 950 mm was specially fabricated in the Soil Mechanics and Foundation Engineering Laboratory of Jadavpur University, Kolkata, for conducting model tests on embankment. The tank was fitted with Perspex from two sides so that the failure of the embankment could be observed. Load was applied over the embankment using a 20 MT capacity hydraulic jack, equipped with a calibrated load cell. Vertical and horizontal deformations were measured using dial gauges of sensitivity 0.01 mm and maximum deflection of 50 mm.

Fig. 4: Preparation of embankment

5.2 Preparation of soft foundation bed

Then the model embankment was prepared by compacting it in four equal layers with a slope of 1:1 (Fig.4). The height of fall of the wooden block, used during preparation of formation soil, to achieve the desired density was determined a- priori by performing a series of trials.

The foundation bed was prepared by compacting the soil using a flat wooden block size 100 mm x 200 mm weighing about 2 kg in six equal layers till the desired height was reached. Through a series of trials the amount of soil, water content of soil, height of fall and number of blows of the wooden block required to achieve the desired density for each lift were determined before hand. By carefully controlling the water content and compaction, a fairly uniform test condition was achieved throughout the test program. Undisturbed samples were collected to check the uniformity of the test bed during each test for determination of the in-situ unit weight and moisture content of foundation soil.

The moisture content maintained during compaction and corresponding densities achieved were determined by collecting samples in small aluminum cans of known volume placed at different locations in the test tank. The placement moisture content and density for both foundation and embankment soil are given in Table 5. Samples were collected from both the embankment soil and foundation clay bed for determination of shear strength parameters. For embankment soil unconsolidated undrained triaxial tests and for foundation clay vane shear tests were conducted. The shear strength parameters are given in Table 5.

82â&#x20AC;&#x192; Volume 45

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Table 5: Placement density and moisture content Property Placement moisture content (%) Placement density(kN/m3) Cohesion (kN/ m2) Angle of internal friction Ă&#x2DC;

5.4

Foundation soil Embankment soil 40 25

16.7

19

5.5

22

00

230

Loading frame and test procedure

Loading frame consisted of a vertical frame with channel base shown in Fig.5. The footing used in the present investigation was 200 mm x 400 mm x 6 mm rigid Mild steel plate. Footing load was applied over the embankment using a 20 MT capacity hydraulic jack, equipped with a calibrated 10 MT capacity load cell. Load was applied gradually in increments as per the provisions of Indian Standard code IS 1888-1982 [10]. Dial gauges were placed at different locations in the embankment to measure vertical, horizontal and heaving of the soft clay bed.

6.

Results and Discussions

Pressure intensity versus deformation curve for model embankments 300mm heigh, unreinforced, reinforced with geogrid and geocell of three different aspect ratio 0.25, 0.33 and 0.50 are shown in Fig.6. From this figure it is seen that pressure at a particular deformation increases in general with the increase in aspect ratio of the geocell. The ultimate bearing capacity and corresponding settlement are determined by double tangent method and are shown in Table 6, alongwith corresponding settlement. From this Table 6, it can be observed that geocell-supported embankments have exhibited higher capacity and lower deformations compared to unreinforced embankment. Further, improvement in bearing capacity and reduction in settlement are given in the same table as a ratio of bearing capacity / settlenmeet for reinforced bed and that of unreinforced one in the brackets.

Fig. 6: Pressure intensity versus deformation response of model embankment supported on geocell layers of different aperture size

Table 6: Results from model embankment tests Fig. 5: Loading assembly and measurement of deformation during the test

During each test, the dial gauge readings were taken after each load increment. This process is continued till the test bed reaches failure. After every test, the foundation and embankment soil were completely scooped out and freshly prepared for the next experiment maintaining the same moisture content and density. Some tests were repeated to ensure the uniformity of test conditions. The Bridge and Structural Engineer

Type of geogrid used for making geocell Unreinforced soil Geogrid Geocell

Aspect ratio (h/D) -

Ultimate Settlement bearing capacity (mm) qult (kPa) 8.75 12.0

0.25 0.33 0.50

14.0[1.60] 14.75[1.68] 16.25[1.86] 18.50[2.11]

11.5[0.96] 5.0[0.42] 3.5[0.30] 2.0[0.17]

Note: h= height of geocell, D= aperture size of geocell Volume 45 Number 4 December 2015â&#x20AC;&#x192; 83


7. Improvement with settlement A non-dimensional factor, called the bearing capacity improvement factor (I.F) has been introduced to compare results from different tests. This improvement factor was defined as the ratio of average pressure over embankment with the geogrid/geocell reinforcement at a given settlement to the average pressure on unreinforced soil at the same settlement. Improvement factor for model embankment with different settlement are shown in Table 7. From this table, it was noted that improvement factor increases with the inclusion of geocelll reinforcement. Further it is observed that for geogrid/geocell improvement factor becomes more or less constant beyond 10 mm (δ/B=5%) settlement. Table 7: Improvement factor (I.F) for different settlement (δ/B) % 2.5 5.0 7.5

Geogrid 1.54 1.47 1.46

h/D=0.25 2.44 2.17 2.14

h/D=0.33 2.91 2.46 2.44

h/D=0.50 3.33 2.72 2.70

Based on these graphs, an embankment for a particular soft foundation bed can be easily designed. For a particular soft foundation bed, the value of acceptable settlement of the overlying embankment may be either assumed or known. From this settlement the acceptable load for that particular embankment can be obtained based on the graph for a particular aspect ratio (height of geocell/ aperture size of geocell). In this way the value of acceptable superimposed load can be predicted based on the limiting deformation of the embankment. To assess the validity of proposed equation, the percentage difference between the predicted and experimental values have been calculated and are tabulated in Table 8. From this Table, it is clear that the difference between the experimental value and the value obtained from proposed equation is within 15% only. Hence it can be said that the proposed equation (Eqn.1) would calcullate the ultimate bearing capacity very efficiently.

8. Non - dimensional analysis for different laboratory tests An attempt has been made to develop a universal design on the basis of the above described experimental work. Hence, load-deformation behaviour has been plotted normalizing both load and deformation axis. The deformation axis is normalized as (deformation/ top width of embankment) while applied load is normalized as (pressure/γd B) where γd is dry unit weight of soil and B is the top width of model embankment. Typical normalized graphs for the model embankment with different aspect ratio of geocell are shown in Fig.7. The generalized logarithmic equation as obtained for different cases are also given in the same figure. The general form of the equation is given in equation (1). P = a ln(δ/B)+b (1) Where P= δ = deformation, a and b are the coefficients, ln means natural log Corresponding values of R2 as given in figure 7 are quite close to 1.0. 84  Volume 45

Number 4 December 2015

Fig. 7: Non dimensional Pressure-deformation response of model embankment supported on geocell layers of different aperture size

Table 8: Percentage (%) difference from experiments and proposed equation for different aperture size of geocell δ/B

Unreinforced % % diff 2.5 6.88 5.0 9.89 7.5 11.13 10.0 11.05 12.5 8.58 15.0 4.74

Geogrid % diff 2.09 12.72 4.53 1.69 3.04 8.1

h/ D=0.25 % diff 4.79 6.36 3.5 3.49 2.71 2.14

h/ D=0.33 % diff 0.19 4.06 2.72 2.11 1.54 0.72

h/ D=0.50 % diff 5.13 6.1 5.37 4.56 3.64 2.59

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Conclusions

supported on geocell reinforced sand underlain by soft clay”, Geotextiles and Geomembranes, 219(4), 197-219.

On the basis of present investigation following conclusions may be drawn : 1. Ultimate bearing capacity of reinforced embankment increases with the decrease in pocket size or increase in aspect ratio. Corresponding failure settlement also reduces.

6. DASH, S.K., RAJAGOPAL, K., and KRISHNASWAMY, N.R. (2004). “Performance of different geosynthetic materials in sand foundations”, Geosynth. Int., 11(1), 35 - 42.

2. Percentage improvement in load carrying capacity decreases with the increase in settlement of the embankment.

7.

3.

The semi-emperical log function shows that the difference between predicted and experimental values is within 15%.

Acknowledgement Authors acknowledge the support provided by the Department of Civil Engineering, Jadavpur University, Kolkata, for conducting laboratory tests in the present study. The first author acknowledge the moral support of the Engineers of Public Works Department, Government of West Bengal, during the study.

References 1.

BATHURST, R.J., and JARRETT, P.M. (1988). “Large-scale tests of geo-composite mattresses over peat subgrades”, Transportation Research Record 1188, Transportation Research Board, Washington, DC, pp.28-36.

2. COWLAND, J.W., and WONG, S.C.K., 1993. “Performance of a road embankment on soft clay supported on a geocell mattress foundation”, Geotextiles and Geomembranes 12(8), 687–705. 3. DASH, S.K., KRISHNASWAMY, N.R., and RAJAGOPAL, K., (2001a). “Bearing capacity of strip footings supported on geocell-reinforced sand”, Geotextiles and Geomembranes 19(4), 235–256. 4. DASH, S.K., RAJAGOPAL, K., and KRISHNASWAMY, N.R., (2001b). “Strip footing on geocell reinforced sand bed with additional planar reinforcement”, Geotextiles and Geomembranes 19(8), 529–538. 5. DASH, S.K., SIREESH, S., and SITHARAM, T.G. (2003). “Model studies on circular footing The Bridge and Structural Engineer

GÖRTLER, H. (1975). “Dimensions analyze.” Berlin Heidelberg New York, Springer.

8. HENDRICKER, A.T., FREDRANELLI, K.H., KAVAZANJIAN JR, E., and MCKELVY III, J. A., (1998).“Reinforcement requirements at hazardous waste site”, Proc. of 6th International Conference on Geosynthetics, Atlanta, 1, pp. 465- 468. 9.

IS: 1888-1982 (reaffirmed). “Method of load test on soils”, Bureau of Indian Standards.

10. KRISHNASWAMY, N.R, RAJAGOPAL, K., and MADHAVI LATHA, G., (2000). “Model studies on geocell supported embankments constructed over soft clay foundation”, Geotechnical Testing Journal, ASTM, 23, pp. 45-54. 11. MADHAVI LATHA, G., DASH, S. K., and RAJAGOPAL, K. (2009). “Numerical simulation of the behavior of geocell reinforced sand in foundations”, Int. J. Geomech., 9(4), 143–152. 12. MADHAVI LATHA G., RAJAGOPAL K., and KRISHNASWAMY, N.R., (2006). “Experimental and Theoretical Investigations on Geocell-Supported Embankments”, International Journal of Geomechanics, ASCE, Volume 6, No.1, January 1, 2006, pp. 30-35. 13. MADHAVI LATHA, G., (2008). “Design of Geocell Reinforcement for Supporting Embankments on Soft Ground”, The 12th International Conference of International Association for Computer Methods and Advances in Geomechanics (IACMAG), October 1-6, Goa, India. 14. MANDAL, J. N., and GUPTA, P., (1994). “Stability of geocell-reinforced soil”, Construction and building materials, 8 (1), pp. 55-62. Volume 45 Number 4 December 2015  85


15. MHAISKAR, S.Y., and MANDAL, J.N., (1994). “Soft clay sub grade stabilization using geocells”, In J N Mandal (Ed) Geosynthetic World, Wiley Eastern Ltd. 1994, pp.139-148. 16. MITTAL, SATYENDRA (2013), "An Introduction to Ground Improvement Engineering", SIPL Publishers, New Delhi. 17. MITTAL, S., SHUKLA, J.P. (2010). “Soil Testing for Engineers”, Khanna Publishers, Delhi., India. 18. POKHAREL, S.K., HAN, J., PARSONS, R.L., QIAN, Y., LESHCHINSKY, D., and HALAHMI, I., (2009). “Experimental study on bearing capacity of geocell-reinforced bases”, 8th International Conference on Bearing Capacity of Roads, Railways and Airfields, June 29 - July 2, 2009, Champaign, Illinois. 19. SITHARAM, T.G, SIREESH, S., and DASH, S.K. (2005). “Model studies of a circular footing

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supported on geocell-reinforced clay”, Canadian Geotechnical Journal, Vol. 42, (2), 693-703. 20. SHIMIZU, M and INUI, T., (1990). “Increase in the bearing capacity of ground with geotextile wall frame”, Geotextiles, Geomembranes and Related Products, Den Hoedt (ed.), Balkema, Rotterdam, 254. 21. SITHARAM, T.G and A. HEGDE (2013). “Design and construction of geocell foundation to support the embankment on settled red mud”, Geotextiles and Geomembranes, 41 pp. 55–63. 22. YOON, Y. W., HEO, S. B., and KIM, K. S. (2008). “Geotechnical performance of waste tires for soil reinforcement from chamber tests”, Geotextiles and Geomembranes, 26(1), 100– 107. 23. ZHOU, H., and WEN, X. (2008). “Model studies on geogrid or geocell reinforced sand cushion on soft soil”, Geotextiles and Geomembranes., 26(3), 231–238.

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GEOTECHNICAL INVESTIGATIONS IN GRAVEL-BOULDER DEPOSITS

Ravi SUNDARAM Director Cengrs Geotechnica Pvt. Ltd. Noida, UP, India ravi@cengrs.com

Sanjay GUPTA Managing Director Cengrs Geotechnica Pvt. Ltd. Noida, UP, India sanjay@cengrs.com

Ravi Sundaram completed his M.Tech. from IIT Delhi in 1980 and has about 36 years experience in geotechnical investigations in India and abroad. His interests include foundations in difficult ground conditions, ground improvement and load tests.

Sanjay Gupta, completed his M.Tech from IIT Delhi in 1974 and has over 42 years experience in geotechnical engineering. His professional interests include geotechnical investigations, installation and testing of pile foundations, artesian conditions and ground improvement.

Summary

1. Introduction

Geotechnical investigation through gravel-boulder formations poses difficulties due to the problems of drilling in such deposits. Various drilling methods such as percussion, rotary and DTH have been used on projects with different degree of success; a combination of two methods is usually effective. In addition to drilling, geophysical tests and load tests help in better assessing the stratigraphy and selecting suitable values of safe bearing capacities.

Gravel / boulder formations are a driller’s nightmare – advancing boreholes through these deposits is a slow and difficult process. Particularly where gravel size exceeds 50-100 mm and the percentage of boulders/ gravel in the formation exceeds 30-50 percent, drilling 100-150 mm diameter boreholes by standard methods such as shell and auger (percussion) or rotary techniques using diamond bits as specified in IS: 1892-1979 is extremely slow and often fails.

Two case studies are presented here to demonstrate the geotechnical investigation methodologies. The first one is for a bridge on the Jammu-Pathankot highway where percussion drilling combined with odex (DTH) drilling was used to drill the boreholes. This was supplemented by electrical resistivity tests and footing load tests. The second case study is for a bridge on the Kohalpur-Mahakali highway in the Terai region. Here, percussion drilling and resistivity tests were performed.

Therefore, geotechnical investigations in boulder deposits pose a veritable challenge. Engineers have tried different methods with varying degrees of success. Boreholes by appropriate methods combined with geophysical methods and load tests, a reasonable assessment can be made to evaluate the stratigraphy and the safe bearing capacity for foundation design.

Keywords: gravel-boulder deposits; boreholes; percussion drilling; odex drilling; resistivity tests; footing load tests. The Bridge and Structural Engineer

Used in conjunction with boreholes, geophysical tests such as electrical resistivity tests and seismic refraction tests can confirm continuity of various strata and the depth of layers. In strata containing boulders and rock in which drilling is time consuming and expensive, substantial savings in cost and time Volume 45 Number 4 December 2015  87


can be achieved by judicious inclusion of geophysical tests in the geotechnical investigation program.

2. Behaviour of Gravel-Boulder Deposits under Load IS: 10042-1981 [1] states that â&#x20AC;&#x153;the performance of bouldary deposits under load is a matter of intelligent guess. The behaviour of boulder deposits under high loads depends upon the size and quantity of gravelboulder and also the nature and amount of the fillerâ&#x20AC;?. Where the boulder/gravel is the predominant material and filler material (sand) exists only in the interstices of the boulder, its behaviour will depend upon the state of packing of the boulders, nature and the size of the boulder and the gradation. If the quantum of filler material is less, the load carrying capacity is high and the compressibility is low. If there is substantial filler material in the interstices of the boulders/gravel, there is an initial compression stage followed by low compression stage when the load carrying capacity is high. The boulder-soil matrix, unlike ordinary soil, shows certain peculiar characteristics when the boulder proportion is large (>30 percent); the deposit shows an initial rapid compression followed by a stage where the compression decreases considerably as the boulders take over the load-carrying function.

3.

Geotechnical Investigation Methods

3.1 Boreholes 3.1.1 Percussion Drilling Advancing borehole by percussion method (mechanized shell and auger) is the most commonly employed method to drill boreholes for geotechnical investigation in gravel / boulder strata. Chiselling may be done using heavy sinker bars pulverize the boulders, push the gravel into the surrounding strata, thus helping the borehole to advance. But, due to the large size of the boulders, driving / pushing casing may damage its shoe. The bore tends to collapse while withdrawing drilling/sampling tools.

blocks the progress of the hole. Grouting the boulders using cement slurry is a reasonably effective solution to keep the hole open; however the scheme is slow since after each run of 1-1.5 m, time has to be given for the grout to set. 3.1.3 Large Diameter Boreholes Large diameter holes (400-500 mm diameter) using piling rigs have been tried at some project sites, but with limited success. Bailer boring or using DMC methods have been successful in some cases. Cleaning the hole and conducting SPT on the natural strata can be difficult. Also, quality of the samples collected may be questionable. However, the cost of the borehole may be quite high. 3.1.4 DTH Odex Drilling Down-the-hole vibratory hammer with air flushing and simultaneous lowering of casing (DTH odex drilling) is an effective way of drilling through strata with large sized boulders / pebbles. The vibratory hammer pulverizes the boulders and the air flushing removes the cuttings. The method is normally not used for geotechnical investigation but may be used in conjunction with percussion or rotary drilling. Sampling should be done carefully to obtain the representative samples of the natural strata. 3.2 Geophysical Tests The commonly used geophysical tests used for assessing the stratigraphy are electrical resistivity tests and seismic refraction tests. In this paper, use of resistivity tests is illustrated as a geotechnical investigation tool.

3.1.2 Rotary Drilling Another method is to use rotary drilling. Usually, holes of NX or HX diameter are used. But in many cases, the gravel tends to slip under the drill bit and 88â&#x20AC;&#x192; Volume 45

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Fig.1: Schematic of Electrical Resistivity Test

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Resistivity is governed by Ohm’s Law, which states that the change in potential across a resistor is proportional to both the current and the resistance. It is a fundamental property of the material and is independent of geometry. Electrical resistivity tests can be effectively used to assess the continuity of the strata and to identify layering. For geotechnical engineering purposes, resistivity testing can provide information regarding lithology that can be correlated with borehole information [2]. The test may be performed using Wenner’s fourelectrode configuration. In this method, four electrodes are spaced at equal distance along a line as illustrated on Fig. 1. The test procedure is in accordance with IS: 3043:1987 [3].

The effective depth of current penetration, and hence that of the investigation, increases with increase in electrode spacing. Based on an analysis of layered formations and empirical studies, Sanker Narayan & Ramanujachary [4] proposed a graphical procedure for computing the true resistivity of various layers. The analysis is called the Inverse Slope Method as illustrated in Fig. 2. Since the resistivity values vary with moisture content, salinity, level of groundwater, degree of compactness, mineralogy and other factors, caution should be exercised in selecting the appropriate resistivity ranges corresponding to the different soil ranges. 3.3 Load Tests

Plate load tests / footing load tests may be effectively used to assess the in-situ bearing capacity of the gravel-boulder deposit. To obtain meaningful results, the size of the test plate / footing should be at least ρ= 2π a R equal to 10 times the size of the largest particle. Thus, where: for gravel, pebbles, boulders, etc., the minimum required plate size shall be rather high, ranging from ρ = apparent resistivity 100 to 300 cm or even more. For such large size, a = electrode spacing either a suitably stiffened plate or RCC footing shall be required. Also, the load required being large, the R = resistance cost of the test is very high. ρ represents the true resistivity of the material if the formation is homogeneous and isotropic in nature. 4. Case study-1: Bein Bridge on JammuHowever, it represents only the apparent resistivity Pathankot Highway (NH-1A) ρa if the formation consists of two or more layers of different resistivities. The apparent resistivity, 4.1 Bridge Details depending on the geology, may be a crude weighted The site for the Bein Bridge in the Jammu-Pathankot average of the true resistivity of the different layers. section of NH-1A is located in the foothills zone of the Himalayas. The salient details of the bridge are as follows: The following equation is used to compute the apparent resistivity for the Wenner configuration:

Structure No.

: 56/1 on NH-1A

Bridge Length

: 259 m

No. of Spans

: 10

No. of Abutments

: 2

No. of Piers

: 9

Span length

: 25.9 m

Average bed level of river

: RL 368.5 m

Maximum scour level at Piers : RL 366.1 m Fig. 2: Inverse slope method

The Bridge and Structural Engineer

Proposed Founding level

: RL 364.69 m (approx. 3.8 m below EGL)

Volume 45 Number 4 December 2015  89


Bed protection was planned to limit possibility of scour.

places. The pebbles / boulders are rounded to subrounded due to the river action.

4.2 Regional Geology

Sand is present as a filler material in the interstices of the boulders and the proportion of sand in the deposit may be between 15 to 30 percent. The voids are only partially filled with sand. The packing of the boulders is generally medium to compact. The photo on Fig. 3 illustrates the nature of the gravel-boulder deposition.

The project area is covered by alluvial deposits of the River Jhelum and its tributaries. It is in the northern rim of the Indo-Gangetic Basin where it joins the foothills zone of the mountains. It is one of considerable faulting and structural strain. The alluvium in this zone consists primarily of coarse gravel/boulders/ cobbles, grits sands and clays. It conceals two or three transverse ridges and pre-Tertiary basins due to crumpling and dislocation of the basement floor during the Himalayan Orogeny. Overall, favourable conditions prevailed for the quick accumulation of sediments in the zone of lodgment at the foot of the mountains. The alluvium is underlain by the Muree and the Siwalik deposits [5]. The deposits in the area consist primarily of large size pebbles, sub-rounded boulders and gravel. The sand infill in between the voids is less than 20-30 percent. 4.3 Generalized Stratigraphy of the Area

4.4 Geotechnical Investigation The investigation methodology used a combination of three methods â&#x20AC;&#x201C; 1. Boreholes through gravel/boulders by odex (DTH) method using compressed air in conjunction with boring by bailer/shell in sand and clay layers. 2.

Geophysical testing (electrical resistivity tests), and

3.

In-situ load tests on a model footing.

4.5 Boreholes through Gravel-Boulder Formation

The alignment of NH-1A in the section between Jammu and Pathankot is in the foothills zone of the Himalayas. In this sector, boulders and pebbles/ shingles are the principal material encountered at shallow depths. As per local information, boulders are likely to be met to at least 70-80 m depth. At several locations, some intermediate sand and clay layers are met within the bouldary formation.

Fig.4: Odex & Percussion Boring

Fig. 3: Boulders, pebbles on river bed

In the project area, the pebbles, boulders and gravel range in size from 20 mm to more than 200 mm with isolated boulders of even 500 mm size being met at 90â&#x20AC;&#x192; Volume 45

Number 4 December 2015

Initially, percussion boring was attempted using heavy sinker bars and chisel. However, where large pebbles / boulders were present, progressing the borehole was found to be slow and difficult. In such cases, the boreholes were advanced by DTH method. The vibratory hammer of the DTH machine pulverizes the boulders, which are brought out by using an air compressor. After drilling every 1.5 m using DTH, the percussion rig was set in place. The borehole was cleaned out using a bailer to remove all disturbed material. Standard penetration test was conducted in accordance with IS: 2131-1981 [6] after the borehole was cleaned. Fig 4 is a photo of odex The Bridge and Structural Engineer


drilling in progress in conjunction with the percussion boring. It is usually difficult to even obtain representative disturbed samples of the bouldary strata since the boulders are broken during chiselling / drilling. Undisturbed samples cannot be collected in such strata. Fig. 5 illustrates the alternate use of DTH and percussion for drilling and sampling.

The boulders / cobbles size range from 100 mm to more than 300 mm. About 30-40 percent of the material is gravel size (4.75-100 mm size) and about 10-20 percent of the material may be more than 300 mm in size (boulders). In general, the packing of boulders is medium to compact. The borehole profiles are illustrated on Fig 6. 4.7 Electrical Resistivity Tests Electrical resistivity test has been used to assess the continuity of the strata and to identify layering. For geotechnical engineering purposes, resistivity testing can provide information regarding lithology and can be correlated with borehole information (Ravi Sundaram & Sanjay Gupta, 2001[7]).

Fig. 5: Borehole drilling in progress. Left: Drill bit for odex drilling. Right: Bailer being used to clean borehole prior to conducting SPT

4.6 Stratigraphy at Bridge Location

The interpretation is based on true resistivity values interpreted from the inverse slope method in conjunction with geology of the area and borehole data. After evaluation of available borehole data, geology of the area, range of measured resistivity values and comparison with published resistivity values, an assessment was made of the probable range of resistivity in each of the different layers. The range of resistivity values and the interpreted stratigraphy as assessed by the authors after review of borehole data and comparison with published values are summarized on Table 1. Table 1: Interpretation of Resistivity Values â&#x20AC;&#x201C; Bein Bridge True Resistivity, ohm-m 30-75 10-50 > 100 50-100 > 200

Fig. 6: Typical Borehole Data â&#x20AC;&#x201C; Bein Bridge

The deposits met at the site classify as boulders / gravel with sand infilling from the ground surface to about 5.0 m depth. Below this, clayey silt is met to about 7.0 to 8.0 m depth. This is underlain by boulders/gravel with sand to final explored depth of 15.0 m. The Bridge and Structural Engineer

Interpreted Stratigraphy Sand and silty sand Stiff clay and sandy silt Hard clay Sand with gravel / pebbles Pebbles and boulders intermixed with sand

As can be seen some of the ranges overlap. Interpretation has been done by comparing the true resistivity values with the nearby borehole data to develop a profile consistent with the expected stratigraphy. Typical result from a test conducted at P-9 is presented on Fig. No.7. The resistivity tests indicated that the boulder formation is continuous to over 50 m depth. Volume 45 Number 4 December 2015â&#x20AC;&#x192; 91


reference to a stable reference bar. After reaching the final load, the load was removed in stages to measure elastic rebound. Fig 8 shows the footing load test in progress. The bearing pressure versus settlement plot is presented on Fig, 9.

Fig.9: Footing Load test – Load-Settlement Curve

Fig.7: Litholog interpreted from Electrical Resistivity

4.8 Footing Load Tests

The tests results are extrapolated for 8 to 10 m wide square footing using the following equation as given in IS 1888-1982:

(1) where Sp = Settlement of test plate/footing of size Bp Sf = Settlement of test plate/footing of size Bf Fig.8: Footing Load test in progress

The footing load test is a model foundation test and can help in making a better assessment of the behaviour of boulder strata under the load. Load test data on suitable size model footing can be used effectively to assess safe bearing capacity of bouldary deposits. Two footing load test were performed at the site of the Bein Bridge at the locations of Pier P-1 and P-6 on a RCC footing of size 2.5 m x 2.5 m at the proposed foundation level. The footing was cured for over 28 days prior to conducting the test. The test procedure was in general accordance with IS: 1888-1982 [8]. Dead load was used through kentledge to provide the reaction. The footing was loaded by pushing up against the dead load using two hydraulic jacks of 300 T capacities each. Four dial gauges were used to measure the footing settlement with 92  Volume 45

Number 4 December 2015

Based on the test results on a footing of 2.5 m size at P-1, the settlement of the footing is 1.7 mm corresponding to a load of 25 T/m2. The extrapolated settlement for 8 to 10 m wide footing shall be about 10 to 15 mm. Some of the salient points interpreted from the test are listed below: 1. The ultimate bearing capacity of the boulder strata exceeds the maximum applied pressure of 6.5 kg/cm2. Applying a safety factor of 2.5, the safe bearing capacity exceeds 25 T/m2. 2.

The back-calculated value of ϕ exceeds 42°.

3.

The settlement at each applied load is essentially immediate.

4.

The modulus of subgrade reaction (k) calculated as per IS 9214:1979 [9] works out as 13 kg/cm3.

5.

Using the correlation,

(2)

The Bridge and Structural Engineer


designed. the modulus of elasticity (E) of the boulder deposit was calculated as 8875 T/m2. 4.9 Safe Bearing Capacity for Design Considering the potential for saturation and possibility of local variations, the maximum settlement of the foundation at P-1 under an applied bearing pressure of 25 T/m2 may be on the order of 10 to 15 mm. Some consolidation settlement will also occur in clay layer met between 5 to 8 m depth (RL 363.5 ~ 360.5 m). The geotechnical investigation carried out confirmed the continuity of the strata and provided a basis for selection of net bearing pressure for design. The footing load test was used as the basis for evaluating the immediate settlement of the boulder deposit. For bearing capacity analysis, the soil parameters used were: c = 0

ϕ = 42 degrees

Table 2: The Soil Parameters used for Settlement Analysis. Depth, m

Soil Classification γ, T/m3 E, T/m2 pc, T/m2

e0

cc1

cc2

-

-

From To

5. Case study-2: Charaila Bridge on Kohalpur-Mahakali Highway (Nepal) 5.1 Project Details The Kohalpur-Mahakali Highway (also known as the East-West Highway) runs almost parallel to the Indo-Nepal border, about 15 to 30 km inside Nepalese territory. The sector investigated is in District Kailali of south-west Nepal. Several bridges are being constructed across minor and major rivers along the road alignment. Fig. 10 presents a vicinity map showing the route investigated and the important rivers of the area. 5.2 Bridge Details The site for the Charaila Bridge is located in the Terai region of the Himalayas. The salient details of the bridge are as follows: Bridge Length

: 259 m

No. of Spans

: 10

No. of Abutments

: 2

No. of Piers

: 9

Span

: 25.9 m

Average bed level of river

: RL368.5 m

0

5.0

Boulders / gravel 1.90 with sand

5000

-

-

5.0

8.0

Clayey silt

2.00

1200

30

0.68 0.015 0.20

Maximum scour level at Piers : RL 366.1 m

8.0

15.0

Boulders / gravel 2.20 with sand

9000

-

-

Proposed Founding level (approx. 3.8 m below EGL)

-

-

where γ = bulk density

: RL

364.69

m

Bed protection was planned to limit possibility of scour

E = modulus of elasticity pc = pre-consolidation pressure cc1 = compression index cc2 = compression index for for stress level ≤ pc stress level > pc

The E value of the top layer to 5 m depth was conservatively selected as 5000 T/m2 to account for any local variations. The soil parameters for the clayey silt layer between 5 and 8 m are based on the SPT values and laboratory test results. Using the above design parameters, the factor of safety against shear failure exceeds 3. The settlement of 8-10 m size foundations of the bridge foundations was calculated as 24.3 mm for a design net bearing pressure of 25 T/m2. The testing generated the necessary confidence that the structure is safe as The Bridge and Structural Engineer

Fig.10: Vicinity Map – River Charaila

Volume 45 Number 4 December 2015  93


5.3 Regional Geology The region between the Rivers Mahakali, Seti and Karnali expose the Siwaliks and the Lesser Himalayas to the north of the Main Boundary Thrust (Bashyal, 1982 [10]). The area is bordered by Kumaon Himalayas in the north and west and Indo-Gangetic alluvial plains in the south. The Siwalik foothills occupy a wide belt and reach a width of 52 km in southwestern Nepal. The Nepalese sub-Himalayas belt has been classified into the Lower, Middle and Upper Siwaliks (Gansser, 1964 [11]). Most of the visible sediments belong to the Middle Siwaliks and are of Middle to Late Miocene & Pliocene to Pleistocene Age.

Rotary drilling was attempted at some selected locations using heavy-duty rotary drill machine. A 32 carat diamond impregnated bit was used. However, the pebbles were slipping under the bit and also causing collapse of the hole, thus making the progress of the hole very slow. Some holes were cement grouted in an attempt to advance the hole. By repeated grouting, the hole could be advanced; but the process was slow and time consuming since one has to allow the cement to set for at least 24 hours before attempting to re-drill. Table 3: Available Borehole Data Borehole

Depth of borehole

5.4 Generalized Stratigraphy of the Area

APP – 1 (approach on right bank)

15.0 m

The Recent deposit of alluvial fan comprises pebbles and boulders set in a sandy/silty matrix. The foothills along the Indo-Nepal border are called the Terai region and are very fertile and well drained. Terrace alluvium is restricted to the paleo-banks of the rivers. It is made up of coarse clusters (pebbles and boulders with sand) in the upstream reaches and grades into finer sediments downstream.

A – 1 (right bank abutment)

40.1 m

P – 1 (Pier No. 1)

16.0 m

P – 2 (Pier No. 2)

13.0 m

A – 2 (left bank abutment)

40.0 m

APP – 2 (approach on left bank)

20.0 m

5.5 Geotechnical Investigation The investigation methodology used a combination of two methods – Boreholes using mechanized bailer/shell with chiselling through the boulder / pebbles strata, and

Geophysical testing (electrical resistivity tests).

The boreholes were progressed by mechanized heavyduty percussion type rigs using shell. Heavy sinker bars and chisel were used to achieve penetration through the bouldary strata and hard clays. Flush threaded casing pipes of 150 mm ID (172 mm OD) were lowered by rotation and hammering. The casing pipe was kept above the boring level so as to avoid disturbance during sample collection and conducting standard penetration tests. However, where large sized pebbles/boulders (100200 mm size) were encountered, the casing could not be penetrated through these boulders. Due to borehole collapse and difficulties in advancing casing through this strata, the progress was very slow. 94  Volume 45

Number 4 December 2015

To evaluate the stratigraphy to the required depth of 30 m at pier locations and 40 m at abutment locations (as per the project specifications), it was decided to conduct electrical resistivity tests. The tests were conducted in accordance with IS 3043-1987 [3] at several locations along the bridge alignment as well as at upstream and downstream locations. Table 3 presents the depth to which borehole data was available. To assess the stratigraphy to 30 to 40 m depth at the pier and abutment locations, resistivity tests were performed along the centre line of the bridge alignment as well as at upstream and downstream locations. The locations were suitably selected so as to get a clear stretch of 80 to 120 m on level dry ground. Sufficient tests were done so as to calibrate the borehole data against the resistivity values. The tests were conducted using the Wenner Array at different electrode spacings ranging from 1 to 40 m. Table 4 presents the details of the resistivity tests conducted. The Bridge and Structural Engineer


Table 4 : Resistivity Tests conducted Location

No. of Resistivity Tests Conducted

Up-stream of Bridge Alignment.

3

Along Centre Line

1

Downstream of Bridge Alignment

7

Total no. of resistivity tests conducted

11

The resistivity data was analyzed in conjunction with the borehole data assess the probable stratigraphy at the required pier / abutment location. Using the resistivity data from the various locations at which the tests were conducted, a three dimensional picture of the stratigraphy was visualized so as to interpolate the soil profile at the required locations. Based on this analysis, a geo-electric litholog that matches with the anticipated stratigraphy is generated. 5.6 Interpretation of Stratigraphy from Resistivity Values Table 5 presents the ranges used by the authors for interpreting the soil layers. These ranges are sitespecific and have been developed after careful review of data and comparison with borehole information. Fig.10 presents one typical geo-electric litholog based on the resistivity tests conducted. Table 5: Soil Strata correlated to Resistivity values True Resistivity, ohm-m

Interpreted Stratigraphy

30 – 75

Sand and silty sand

10 – 50

Stiff clay and sandy silt

> 100

Hard clay

50 – 100

Sand with gravel / pebbles

> 200

Pebbles and boulders intermixed with sand

A generalized surface profile along the centre line of the bridge is presented on Fig. 11. The stratigraphy beyond the depth investigated by the boreholes is The Bridge and Structural Engineer

Fig.11: Interpreted subsurface profile along the centre line of the bridge

obtained by interpolating and projecting the geoelectric profiles along the centre line of the bridge alignment. Based on this evaluation, the stratigraphy at the required pier/abutment location is generated. The geotechnical parameters required for the analysis is interpreted by comparison with the borehole data and the engineer’s assessment of the trend of values.

6.

Concluding Remarks

In strata containing boulders / gravel, drilling is expensive and time consuming. Judicious inclusion of resistivity tests and load tests in the geotechnical investigation programme can save substantially on both time and money and also ensure the quality and reliability of the data for foundation design. It cuts down the time required for a geotechnical investigation programme substantially. In the current scenario of fast track projects in the highways and infrastructure sector, it can be used to reduce the number of boreholes required to be drilled. However, the inherent limitations of geophysical tests and model tests on prototype footings should be recognized. The tests should not be considered as an alternative to borehole drilling. It should be used in conjunction with sufficient borehole data for realistic interpretations of stratigraphy and anticipated loadsettlement behaviour that match well with actual ground condition. Thorough knowledge of local conditions is essential so as to correlate the results with strata conditions. Prior to conducting geophysical tests, information on geology, geomorphology and anticipated stratigraphy of the project area should be collected. The interpretations should be done by experienced Volume 45 Number 4 December 2015  95


personnel with geophysics.

a

thorough

understanding

of

Extrapolation of load test data should include consolidation of any cohesive strata in between the boulders / granular strata for a realistic assessment of foundation behaviour.

7. References 1.

IS: 10042-1981 Reaffirmed 2002, “Code of Practice for Site Investigation for Foundation in Gravel-Boulder Deposit”, Bureau of Indian Standards, New Delhi

2. WOODS R.D., “Geophysical Characterization of Sites”, Volume prepared by International Society of Soil Mechanics and Foundation Engineering, ISSMFE Technical No.10 for 13th ICSMFE, Oxford & IBH Publishing Co. Eds., 1994.

5. RAVINDER KUMAR, “Fundamentals of Historical Geology and Stratigraphy of India”, Wiley Eastern Ltd., New Delhi, 1985. 6.

IS: 2131-1981 Reaffirmed 2002, “Method for Standard Penetration Tests for Soils”, Bureau of Indian Standards, New Delhi.

7. RAVI SUNDARAM and SANJAY GUPTA, “Use of Electrical Resistivity Tests for Geotechnical Investigations for Bridges”, Journal of Indian Roads Congress, 2001, pp. 115-131. 8.

IS: 1888-1982 Reaffirmed 2002, “Method of Load Test on Soils”, Bureau of Indian Standards, New Delhi

9.

IS 9214:1979 Reaffirmed 2007, “Method of Determination of Modulus of Subgrade Reaction (K-Value) of Soils in Field”, Bureau of Indian Standards, New Delhi.

3.

IS: 3043:1987 Reaffirmed 2006, “Code of Practice for Earthing”, Bureau of Indian Standards, New Delhi

10. BASHYAL R.P. “Geological Framework of Far-Western Nepal”, Himalayan Geology, XII, 1982, pp. 40-50.

4.

SANKER NARAYAN P.V. and RAMANUJACHARY K.R., “An Inverse Slope Method of Determining Absolute Resistivity” Short note, Geophysics, XXXII (6), 1967, pp. 1036 - 1040.

11. GANSSER A., “Geology of the Himalayas”, Interscience Publishers, London, 1964.

96  Volume 45

Number 4 December 2015

The Bridge and Structural Engineer


Rectification scheme of collapsed sheet pile adjacent to existing railway line â&#x20AC;&#x201C; A case study

D. N. NARESH General Manager NTPC Ltd. India dnnaresh@ntpc.co.in

Jitendra KUMAR Dy. General Manager NTPC Ltd. India jitendrakumar07@ntpc.co.in

Dr. D. N. Naresh is PhD in civil engineering from NIT Warangal. He is currently holding a position of General Manager in NTPC Ltd. He has worked in various areas of civil engineering and has got expertise in soil and foundations engineering.

Jitendra Kumar received his bachelor degree in civil engineering in 1997 from IIT Roorkee & master degree in soil mechanics & foundation engineering from IIT Delhi in year 1998. He joined Bharat Heavy Electrical Limited in year 1999 & subsequently joined NTPC Ltd in year 2007.

Abstract This paper presents a case study of a deep excavation restoration scheme at a power project site located in central India. The project site was having an operating power plant and the construction of additional track hopper and wagon tippler was undertaken for the capacity addition. The coal is unloaded inside power plant area for which structures like track hopper, wagon tippler are required which are underground structures necessitating deep excavation. This case study presented here is for the supporting scheme of deep excavation done for the construction of new track hopper adjacent to the existing track hopper. Since, the project was an expansion project & the area available for construction of new units was very limited, the possibility of excavation with mild slope was not possible & the excavation was required to be done nearly vertical. The deep excavation required for track hopper and wagon tippler construction was required to be supported. The main challenge at the The Bridge and Structural Engineer

Mohit JHALANI Manager NTPC Ltd. India mohitjhalani@ntpc.co.in Mohit Jhalani received his bachelor degree in civil engineering in 2008 from MNIT Jaipur. He joined NTPC Ltd in year 2008. Afterwards, he received his master degree in Geotechnical & Geoenvironmental Engineering from IIT Delhi in year 2015. He rejoined NTPC Ltd. in year 2015.

site was that near to the excavation site, track hopper of existing plant was present and the railway wagon movement was there at just 13.0m from the excavation site that did not permit slopes in excavation. Keeping all the site conditions in mind, sheet piling scheme was designed. The sheet piling was to be done in two tiers and to be driven upto the designed embedment depth. During construction, when a railway wagon came in the existing railway track the sheet pile supporting system collapsed and huge distress in the earth was observed. In this stretch the sheet pile also got distressed and lead to collapse of soil. The distress portion was temporarily supported by the wooden planks/logs and the wagon movement had to be stopped till restoration scheme was done. To sustain coal supplies for the generation of existing power plant, immediate remedial measures were required at site to enable to support excavation and the movement over nearby railway track. A site visit was conducted and the restoration scheme was planned as Volume 45 Number 4 December 2015â&#x20AC;&#x192; 97


per site condition. The scheme involved temporary supporting the soil cuts, providing lateral support in the undisturbed portion and horizontal beam with bracing in the disturbed portion. The scheme was validated on staad-pro and plaxis softwareâ&#x20AC;&#x2122;s and the same was implemented at site. The new track hopper has been constructed in time and is in operation now. Initial sheet pile scheme and detailed restoration scheme followed is presented here in this paper.

investigation in the areas pertaining to this project. Based upon the borelog data the top layer of 6 m thickness consists of stiff silty clay with N values varying from 10 to 40, followed by layer of silty sand upto the depth of 21 m with N values higher than 50. Below this sand stone is encountered. A typical borelog of this area is at fig no.1. Ground water level is at a depth of about 1.5 m below natural ground level at the time of investigation.

1. Introduction Coal is one of the lifelines of a Thermal Power Plant. In order to cater the requirement, coal is transported through railway wagons and is unloaded inside power plant area for which structures like track hopper, wagon tippler are required which are underground structures & need deep excavation. The paucity of land has forced the power generators to use the available land within the existing power plants and make use of some common facilities. This not only reduces the total area requirement it also gives the possibility to achieve economies of scale. The construction activity near to the existing plant facilities pose a challenge to both the design and construction engineers wherein they needs to ensure the safety of existing structure and carry out construction activity in the available limited area. The case presented here is of construction of a track hopper adjacent to the existing track hopper. Track hopper is an underground structures which needs deep excavation of the order of 12 m from the ground level under high water table. The sheet pile supporting scheme for deep excavation is included in this paper. The deep excavation was also modelled on Plaxis software (finite element analysis based) and the design was validated.

Fig. 1: Typical borelog of CHP area

This case study is for the construction of track hopper-IV which is adjacent to the existing track hopper-II. The centre to centre distance between the two track hoppers is about 31 m and the excavation depth required is about 12 m. The clear distance between the two track hoppers is 11.0 m. The same is indicated in the layout shown in fig.2. Photograph showing the vicinity of existing and new track hopper is placed at Fig. 3.

While during construction due to sudden rise in pore water pressure and wagon movement, the sheet pile system got collapsed. An immediate remedial measure was suggested to correct the collapsed sheet pile system and resume the existing plant operation which got affected due to collapse. The rectification system is also presented in the paper.

2. Layout and Geotechnical Parameters of site Employer has carried out detailed geotechnical 98â&#x20AC;&#x192; Volume 45

Number 4 December 2015

Fig. 2: Plan layout of track hopper area

The Bridge and Structural Engineer


Following excavation sequence was proposed

Fig. 3: Photograph showing the vicinity of existing and new track hopper

3.

Proposed sheet pile scheme

Since the construction of new track hopper is very near to the existing track hopper, possibility of mild slope excavation is ruled out. The excavation of about 12 m is to be done within the available area i.e. nearly vertical. In case of nearly vertical cut, based upon the soil conditions the excavation is to be supported by sheet piles. As the water table is high, continuous dewatering is also required during excavation. For 12 m deep excavation, sheet pile scheme is designed & validated on plaxis software and it was proposed that the earth adjacent to existing track hopper be retained by driving two rows of steel sheet piles of 12 m long having sectional modulus of 1940 cm3/m.

Existing Track Hopper

-

Make vertical cut of 3.0 m deep

-

Keep a horizontal 3.5 m wide berm

-

Drive 12 m long sheet pile to a depth of 3 m below existing foundation level (RL(+) 275.0 m).

-

Excavate 4 m depth of soil adjacent to sheet pile.

-

Keep a horizontal 2 m wide berm

- Drive second layer of 12 m long sheet pile to a depth of 7 m below existing foundation level (RL(+) 275.0 m). - Consider earth pressure on combined sheet pile wall i.e. further excavation upto foundation level shall be done in 1H:1V slope. The detailed scheme drawing indicating the excavation sequence is shown in the Fig. 4.

4.

Collapse of excavation pit

The excavation was started at site as per the designed scheme. The sheet piles were driven in two tiers and upto designed embedment depth. Fig-5 indicates the sheet piles as per design at site. In some areas the second tier of sheet piles could not be driven upto the required embedment depth. Further the passive earth pressure which was to be maintained by excavating the last 5 m in 1H:1V slope could not be done, instead a vertical cut was done. As a result

New Track Hopper Fig. 4: Sectional view of existing and new track hopper

The Bridge and Structural Engineer

Volume 45 Number 4 December 2015â&#x20AC;&#x192; 99


of this, after some days when the wagon movement started on the adjacent existing railway track, a stretch of about 10-15 m of sheet piles in front of

5.

Rectification scheme

After the collapse of sheet pile near MH-1 area, the railway movement on the nearby track hopper-II was stopped to avoid any further distress in the slope and existing track hopper structure. The distressed sheet pile was temporarily supported by wooden planks. The rectification measures were immediately required at site so that the existing track hopper-II can be put back in operation as soon as possible as it was affecting the existing plant operation. Following measures were proposed to restore the sheet piles and strengthening of earth in the distress portion of MH-1, south side of existing track hopperII area. First of all the undisturbed portion of sheet piles at both side of the deflected portion of sheet pile in MH-I area was strengthened by providing lateral support in form of props. The arrangement of strutting/prop was designed considering surcharge of wagon movement.

Fig. 5: Photograph showing the sheet pile scheme as per design

MH-I is deflected. The vicinity of operational track hopper-II at just 11 m away from the excavation site lead the trigger of sheet pile wall collapse. The initial photographs of the sheet pile deflection and soil collapse is placed at Fig. 6 & 7.

For the disturbed portion, horizontal beams with bracing was provided supporting on the portion of sheet piles at both sides of deflected area.

Vertical cut of soil in the disturbed portion adjacent to railway track was strengthened by providing sheet and strutting to the horizontal beam and sand bags were provided between sheet and horizontal beam.

Fig. 6: Collapse of sheet pile

Fig. 8: Plan of restoration scheme Fig. 7: Collapse of sheet pile

100  Volume 45

Number 4 December 2015

Based upon the scheme envisaged the supporting arrangement was designed and it was analyzed on The Bridge and Structural Engineer


plaxis software for stability. The plan view of the detailed scheme is as shown in the Fig. 8. The sectional view of the detailed scheme is as shown in the Fig. 9. It is to be noted that sand bags were placed to fill the space created by the distress between the two tiers of sheet piles. Plum concreting was done behind the tier-1 sheet pile to avoid any soil movement.

Fig-9: Sectional elevation of restoration scheme

The sections for the supporting scheme were proposed after analyzing it on staad-pro. The staad-pro model is also placed at fig 10.

scenario, the challenge was to do construction in limited area without affecting the safety of existing structures especially underground utilities. The case study presented here suggests that in order to construct underground structure/ deep excavation, suitable supporting structure such as sheet piles, diaphragm walls, contiguous piles etc along with proper dewatering system should be designed adequately considering the field conditions. The more important aspect of deep excavation is during execution which is to be implemented as per the construction sequence. The sequence of activities plays a dominant role in ensuring the safety of structure. After the collapse, the restoration scheme was developed based upon the prevailing site conditions. The restoration scheme was executed successfully and finally the track hopperIV structure was constructed without affecting the operation of existing plant. This paper may be useful for the sheet piling activity during bridge construction and any other construction activity which requires deep excavation.

7. Acknowledgement Authors would like to thanks M/s. NTPC Ltd. management for giving necessary support and permission to present the case study on restoration scheme of deep excavation.

8. References 1.

Fig. 10: Basic Staad model of restoration scheme

Deep well dewatering scheme were used to lower the groundwater levels to provide stable working conditions in deep excavations. A deep well system comprises an array of bored wells, each fitted with electrical submersible borehole pump.

6.

Concluding Remarks

Deep excavation near the existing structure(s) is a challenge. Construction near to the existing structure(s) and also in the areas where land is scarcely available, remains the need of the hour. In the present

The Bridge and Structural Engineer

IS: 800 General construction in steel - Code of practice

2. IS:9527(3) - 1983 Code of practice for construction of port and harbor structures – Sheet pile walls 3. IS:2314 -1986 Code of practice for steel sheet pile sections 4.

Piling handbook – Arcelar mittal, 8th edition, reprint in 2008

5.

California trenching and shoring manual, reprint in 1995.

6. United states steel – Steel sheet piling design manual, updated by FHWA, 1984. 7.

Technical specification of coal handling plant of super thermal power plant, stage-IV, NTPC Ltd.

Volume 45 Number 4 December 2015  101


Engineering Aspects of Storage Caverns

Altaf USMANI Manager Engineers India Ltd. New Delhi, India altaf_usmani@rediffmail.com Altaf Usmani, born 1977, received his civil engineering degree from AMU Aligarh, and Ph.D in Geotechnical Engineering from IIT Delhi.

Chandan Kumar, received his civil degree from MIT Masters in Rock from IIT Delhi.

Summary Long-term stability of geotechnical structures is critical to any engineering design. Design of large underground structures such as storage cavern is initially planned using empirical approaches based on rock massâ&#x20AC;&#x201C;classification system. This design is mainly based on limited data obtained from boreholes and surface geological mapping. This design is continuously updated based on the information gathered as the excavation work progresses in the field. In underground crude oil storage facilities, concrete plugs are required to contain crude oil inside caverns and are designed as gas-tight to prevent movement of any oil or vapour outside the cavern. Construction of plugs involves mass concreting under the typically hot and humid conditions found in underground construction. Therefore, the construction of plugs requires an engineering design that takes into account not only design loads and reinforcement, but also the systematic cooling arrangement and efficient grouting mechanism so that the basic purpose of the plug design is not defeated. This paper discusses various design and engineering aspects involved in the construction of large underground crude oil storage projects with specific emphasis on different kind of concrete plugs. 102â&#x20AC;&#x192; Volume 45

Atul NANDA Head, Technology Engineers India Ltd. New Delhi, India ananda@eil.co.in

Chandan KUMAR Manager Engineers India Ltd. New Delhi, India c.kumar@eil.co.in

Number 4 December 2015

born 1976, engineering Bihar, and Mechanics

Atul Nanda, born 1958, received his civil engineering degree from IIT Bombay and Ph.D in Geotechnical Engineering from Virginia Tech, USA.

Keywords: Caverns, rock, plugs, tunnel, shaft, cooling, grouting

1. Introduction In the present geo-political scenario of world, energy security of a nation has gained paramount importance. In order to ensure energy security, various countries have stored their federally owned oil stocks in underground structures by adopting various underground technology i.e. unlined rock caverns; solution mined salt caverns or underground concrete tanks. The established technology of unlined rock cavern which have been successfully adopted in various Scandinavian and Asian countries, has been implemented in India as part of the strategic petroleum reserves (SPRs) entailing crude oil reserves of 5.33 MMT under Phase I storage program at three locations viz. Visakhapatnam, Mangalore and Padur (Udupi) with storage capacity of 1.33 MMT, 1.5 MMT and 2.5 MMT respectively.

2.

Unlined Rock Caverns

The basic principle of storage of crude oil in underground The Bridge and Structural Engineer


unlined rock cavern is the hydro-geologic containment. Thus the rock caverns are planned at a depth such that adequate hydrostatic pressure is maintained to counter the vapour pressure of the stored product. In order to prevent bubbles of gas moving upwards through any of the fractures, it is established that fracture shall be filled by water which inflows in the downward direction with such a velocity that there is a hydraulic gradient in the rock, i.e., the value of which exceeds 1.0 when measured in the vertical direction. This requirement is an important feature to be considered in the design, especially in a condition when a rock is crossed by fractures in more than one direction. For to secure the flow vector of water from the rock mass towards the cavern, a water curtain system is provided consisting of galleries located above the crown of the cavern. Boreholes are drilled from water curtain tunnel to intersect the predominant joint sets of rock mass. Saturated rock mass coupled with ground water flowing into caverns, ensures proper sealing of the stored product and prevents leakage, since de-saturation of rock mass is an irreversible process. Therefore, a mandatory requirement is specified to have a charged water curtain system and resultant saturation of rock mass 50m in advance of the underlying excavation face of storage cavern. Figure 1 illustrates the layout of storage facilities for a typical unlined rock cavern. The project consists of parallel caverns generally of U-shaped in plan and D shape in cross section. The underground facilities essentially consist of access tunnel, main cavern (around 30 m height, 20m width and about 300-1000 m length) and water curtain tunnels running parallel to and 20 m above each U shaped storage caverns with a series of water curtain boreholes drilled perpendicular to it. The pillar width between the two caverns is kept

Fig. 1: Typical Project Layout

at 30 m to maintain the stability of the cavern and to avoid short circuiting of the crude flow. Also vertical shafts for various pump and instrument installation The Bridge and Structural Engineer

are excavated from the surface which connects the water curtain tunnel & cavern. The underground caverns, access tunnels and shafts are sealed with large concrete plugs in a suitable place to ensure gas tightness. Plugs are constructed in tunnels connecting the caverns and in the shafts. Tunnel plugs have a size of 8mx8m with a thickness of 3-5 m and the shaft plugs have a size varying between 4m to 12 m and thickness between 2.5 to 5 m. The plugs are reinforced concrete structures with M40 concrete. Tightness of these plugs is ensured by cooling and grouting arrangements. The space above the plug in the shaft is either filled with water or concrete.

3.

Site investigation

Though the unlined storage cavern technologies are well proven and implemented elsewhere in the world, adoption to the specific site conditions remains a significant challenge to overcome because of the associated underground uncertainties. Extensive and planned investigations are necessary to minimize geological surprises during construction. From project conception through feasibility studies to the basic design stage and throughout the construction phase, geotechnical investigations are designed to provide the level of information appropriate to the particular project development stage. The investigation program is made to collate, generate and analyse engineering geological and rock mechanics data for site characterization in different stages of investigation and also to identify potential geohazards that may exist at project sites. Exploratory investigations like hydrofracturing, water pressure test, vertical and inclined boreholes are carried out during feasibility studies to decide on selection of the most favourable site areas within the regional physical setting, determine the general layout of structures best suited to the site conditions and evaluate the influence of hydrogeology on site design and construction. Various investigations performed during detail design stage are toward site-specific studies which provide the detail and depth of information. Geotechnical design parameters are established, the basic hydrogeological model is updated with respect to the geological hotspots and the seepage assessments Volume 45 Number 4 December 2015â&#x20AC;&#x192; 103


are reviewed and the grouting requirements are ascertained. During construction, investigations are performed to confirm certain geological hotspots and other anomalous geological structures. Also additional field investigations and laboratory investigations or any other types of measures that deems necessary for fulfilling the design and construction requirements is also be performed during construction.

4.

Engineering & Design

A specific methodology for the design of such storage system is employed which is a wise combination of computation, comparison and monitoring. This approach of design of hydrocarbon storage is more adaptive, in order to better cope with the associated uncertainties. Specific unlined storage cavern requirement need to be considered during design stage which includes minimum and maximum pressure variations during empty and fill conditions, water inflow that needs to be minimized while avoiding any kind of de-saturation and long term stability which is ensured by a careful geological setting, product pressure and cross-section selection. The layout, cross-section and elevations are finalized considering the product storage and operational requirements as well as the geological and geotechnical conditions at site. The layout and cross-section is selected so as to achieve a favorable stress situation in the rock mass and also take into account any major geological structures. The excavation sequence and methodology is also considered while finalizing the cavern layout and sizing of tunnels. Before excavation, detailed design is performed to ensure the quality, lifetime, future maintenance and overall safety of structures. Detailed design is performed on state-of-the-art practice. Validity of material models used in the design is verified by monitoring and other observation during construction phase. Areas where the rock stress situation needs to be analyzed in more detail is identified and also areas where the stress situation is favorable. The critical areas identified from the analytical calculations are analyzed by numerical methods. 4.1 Geological model The Geological model cover the regional geological settings so as to further focus on the project geological setting. The project geological setting encompasses 104â&#x20AC;&#x192; Volume 45

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the key aspects such as Topography, Geomorphology, Rock Type, lineaments and Major Discontinuities as shown in Fig. 2. Geological Model describes the thickness of soil cover and extent of weathered rock. The intrusive features like dyke and hydrothermal zone are delineated from the bed rock. The anticipated region of weakness zones intersecting the layout are marked on model as geological hot spots. They are classified as risky zones and are termed â&#x20AC;&#x153;Geological Hot Spotsâ&#x20AC;? during construction.

Fig. 2: A typical Geological Model

4.2 Hydrogeological model Owing to the containment principle of storage, the hydrogeological regime of the project forms an important aspect, which encompasses aspects such as topography, annual rainfall, recharge, geological setting and hydrogeological properties. Hydrogeological model studies are carried out to check the flow pattern around the caverns so as to confirm hydraulic containment and estimate seepage rates based on the data collected during the investigations. The considered design parameters include provision of water curtain above the cavern with boreholes charged to a head equivalent to required pressure, maximum operating gas pressure at the cavern crown and the vertical distance between water curtain gallery and cavern to satisfy the requirement of hydraulic gradient greater than 1.0 at cavern roof level. Finite element studies are carried out to estimate hydraulic gradient, seepage in the caverns during construction and water requirement during construction. The analyses are carried out for several conditions of operation, i.e. completely empty at atmospheric pressure, maximum normal vapour pressure, cavern units under different pressures, etc. Based on the hydro-geological studies, criteria for grouting are established including requirements of probing, grouting scheme and likely grouting zones. The Bridge and Structural Engineer


4.3 Geotechnical model & Rock support Developing of geotechnical model of the project essentially consist of a comprehensive evaluation of geological & geotechnical data, review of additional site and laboratory investigations, site characterization, identification of main geological features such as dykes, faults, folds, shear zones etc., rock mass classification, engineering properties of soil & rock mass required for numerical analysis, description of in-situ stress condition, ground water condition, blast damage factor, excavation sequence etc, numerical analysis for structurally induced instability & due to overstress, support and stabilization measures including calculations to verify the long-term stability of design and verification and modification through in situ monitoring and back analysis. Initially, rock support is designed using the typical rock support chart proposed by Barton et al. (1974). The rock caverns are then numerically analyzed by this support system to check the stability of the cavern in terms of wedge stability, displacement, and stresses. In numerical studies, two- dimensional (2D) Elasto plastic analysis are performed using a finite- element code as shown in Fig. 3 to study the rock mass behavior as a result of staged excavation in the cavern based on the generalized Hoek-Brown criterion (Hoek & Brown, 1980). Intersections such as shaft with caverns, cavern junctions etc are more vulnerable to the problem of instability because of an increased span and chance of a simultaneous blast. Therefore, these intersections in the cavern are analyzed using a linear 3D boundary element code, to study the effect of a larger span of the excavation in all directions.

Fig. 3. : Numerical Analysis of Caverns with Rock Support Installed

4.4 Design Updation The basic design needs to be checked, validated, updated and modified based on the actual geological, The Bridge and Structural Engineer

geotechnical and hydro-geological conditions encountered during excavation. Before actual cavern excavation, water curtain tunnels are constructed and water curtain system above the cavern is put in place by drilling numerous systematic water curtain boreholes. This water curtain tunnel also acts as a pilot tunnel before cavern excavation revealing the geological setting of caverns below. The predicted lineaments and the major discontinuities (geological hotspots) in the geological model are probed/ cored in advance during excavation of this tunnel. This information along with coring/BHTV studies in water curtain bore holes further increases the knowledge of geological conditions and the geological model is refined through this active design process with main focus on the geological hot spots. Thus the geological, geotechnical and hydro-geological conditions are regularly updated after every round of excavation (Nanda, 2012). After completion of cavern heading, a detail back analysis of the rock mass condition is made to evaluate the rock support installed. The support evaluation of heading is made keeping in mind the rock mass which may be encountered in the subsequent benching with the help of geological L-sections. The geotechnical monitoring results measured using optical targets and extensometers are analyzed with respect to excavation sequence and the assessed rock class type. The adequacy of rock support installed for roof stability and for full cavern section is also checked.

5. Construction Construction of storage cavern involves excavation by drill & blast method with fully grouted untensioned rock bolts and fibre reinforced shotcrete as the principal means of support, along with concrete barriers, pavements and water curtain systems. Steel sets and ground anchors are also adopted in exceptionally poor conditions. A typical construction cycle involves surveying, probing, drill & blast, mucking, scaling, geological mapping, rock support (rock bolt & shotcrete) and grouting as and when required. The water curtain system is the most critical part of the store. Water curtain boreholes are charged at least 40-50m ahead of the cavern excavation to ensure saturation of the joints present in the rock mass The water curtain tunnel along with the boreholes also Volume 45 Number 4 December 2015â&#x20AC;&#x192; 105


serve to update the geology and hydrogeology and decide on requirements of pre-grouting and any other specific requirements for excavation of the caverns. In compliance to the laid out guidelines, each storage units are tested independently for air tightness. Tightness testing is carried out after completion of all concrete barriers in tunnels and shafts, installation of casings/pipes, water filling in water curtain tunnel & access tunnel to design level and backfilling of shaft with mass concrete/water. The tightness testing of each storage unit is performed by applying a moderate over-pressure of compressed air inside the caverns. Before application of air pressure, all casings/ pipes is sealed with temporary covers/flanges. Air pressure is applied by deployment of compressors through one of the casing pipes, which is followed by recording pressure, atmospheric pressure, seepage and temperature changes over time

6.

Concrete Plugs

Concrete plugs in crude storage facilities are a critical component used to contain the crude oil inside the rock cavern. These plugs are designed as gas-tight to prevent any escape of oil or gas and to withstand differential pressure occurring on account of different fluid pressures stored across the plugs. There are generally two types of plugs required in underground storage; tunnel plugs (Fig. 4) and shaft plugs (Fig. 5). The tunnel plug is vertically located at a point

Fig. 4 : Plan and sectional view of tunnel plug along with manhole

Fig. 5 : Plan and sectional view of shaft plug along with casings

to isolate crude oil, forming a separation between oil and water. On the other hand, the shaft plug is 106â&#x20AC;&#x192; Volume 45

Number 4 December 2015

horizontal in the section and is covered by a long column of water or concrete, under which oil and vapor is retained. Plug construction involves a substantial volume of concrete, which results in the development of a great amount of heat- hydration. Therefore, the probability of cracks occurring due to heat of hydration of the cement needs to be avoided. Provision is made in the design stage as well as during the construction process to prevent the development of large temperature gradients caused by hydration of cement using concrete cooling arrangements. In order to ensure gas tightness of the plug for prevention of any gas leakage, contact grouting of the plug is carried out at the concrete and rock interface until a proper seal is established. 6.1 Design of Plugs There are three stages of the design of the concrete plug: (1) structural analysis and concrete design for worst loading conditions to transfer the load safely to the surrounding rock, (2) thermal analysis to study heat generation and cooling-system design, and (3) design of a grouting system that will provide desired longterm sealing for the plug. The concrete plug thickness and reinforcement are designed to safely transfer the load coming from the plug into the surrounding rock. This should be well supported by the cooling system, which is designed to reduce the risk of thermal cracks that can develop during the strength gain in the concrete followed by grouting arrangement to ensure gas-tightness of the plug (after shrinking of the concrete). Cooling systems are provided first to limit the risk of thermal cracking during hardening of the concrete and later to produce an injectable gap after shrinking through which the grouting material can percolate; thus, grouting is carried out to make the plug densest and ensure tightness of the plug. The total pressure acting on the plug is transferred to the surrounding rock by means of a key that is cut into the rock surface all along the periphery of the tunnel walls (Fig. 4). The plug is retained both because of the cut in the rock (key-in) and because of the forces of friction and adhesion occurring along the contact surface of concrete with the rock. The plug design is based on long-term operational requirements as well as construction methodology adopted specific to each plug, either in the vertical or horizontal plane. Thus, the design philosophy is different in the case The Bridge and Structural Engineer


of a shaft plug versus a tunnel plug. Access tunnel plugs are designed for single-stage plug construction to avoid creation of any joint of weakness; hence, the construction sequence has no impact on its design philosophy. However, the shaft plug is constructed in two stages or more to avoid heavy shuttering required to cater for several hundred tons of reinforcement, piping, cooling, and grouting-pipe loads. The plugs are constructed in specially excavated plug key-ins designed for the purpose. In order to construct stable plugs, plug key-in locations are chosen in rock masses that are massive, less jointed, and have high rock quality and Q-values. The selected location should have minimal-to-no seepage. It should also not be very close to any tunnel/cavern junction areas. In case of any explosion inside the caverns, plugs can be affected by the resulting pressure wave; hence, an accidental load of 1.0 MPa is also considered during the design stage. Thus, the plug is designed to withstand maximum loads considering the preceding requirements. The plug design is checked for minimum crack width requirements prescribed for liquid retaining structures, which is limited to 0.2 mm as per Cl. 4.4.1.2, IS 3370: Part 2 (2009). A manhole in the center of the tunnel plug is also provided to access the other side of the plug to facilitate removal of workers and machinery after completion of the plug. The design of the tunnel and shaft plugs was carried out using an M40 concrete mix with sulfate-resisting cement owing to its contact with the oil’s surface. Shaft plugs are located on an interface at which crude oil is located in the lower zone below the plug, and the upper zone is filled either with concrete or a water column. Shafts have sizes varying from 4 to 12 m. This plug is designed to withstand differential pressures on either side and also to support the load of casing pipes that are connected to different seepage and crude oil pumps. These casings loads are quite large and vary from 200 to 300 tons, which is an operational requirement of storage structures. In this project, shaft plugs are designed to resist the worst combination of pressure coming from the water/ concrete column, self-weight, and weight of the casings and carrier pipes. Plugs consist of a trapezoidalshaped bottom layer and top layer, as shown in Fig. 5, and are sealed in an enlarged part of the shaft section known as the plug key. This widening provides a good seat for the plug and distributes the unbalanced load into the surrounding rock. The Bridge and Structural Engineer

6.2 Cooling Design It is well known that heat is produced during hydration of concrete. The temperature continues to rise within the structure as long as the rate of heat development is higher than the rate of dissipation to its surroundings. The structure will expand in volume during this initial phase, and later it will contract as it cools. It is observed that volumetric changes within such a structure do not take place freely owing to the presence of different restraints that would also result in thermal stresses. In order to ensure plug tightness, the risk of thermal cracking is reduced by using cooling arrangements in the plug concrete during and after casting of concrete until the concrete temperature reduces to ambient levels. Plug tightness is also achieved through a combination of cooling arrangements used to shrink the concrete plug (as part of secondary cooling) followed by grouting to fill any gap created while the concrete cools. In order to study the behavior of concrete, a set of semi adiabatic temperature tests is carried out on 1 × 1 × 1_m3 samples on site in confined conditions, with the concrete filling in a box closed on all four sides. Cube samples were made of M40 mix (IS 456 2000), and a water–cement ratio of 0.32 was considered per the actual concrete mix design recommended for plug construction. Four temperature sensors were placed inside the cube, with one at the centre and three on the side-faces of the cube. One sensor was placed outside the cube to measure ambient temperature, which maintained a constant 30–31°C. The temperature of concrete in the adiabatic box was monitored on an hourly basis continuously for seven days, as shown in Fig. 6, for four sensors.

Fig. 6: Temperature variation of sensors during adiabatic test

6.2.1 Different stages of cooling Arrangement As tunnel and shaft plugs in an underground storage facility constitute the most vital part, no risk to the Volume 45 Number 4 December 2015  107


life of these concrete structures is allowed. Hence, the design of a compatible cooling system is seen as the only practical solution in such types of plugs. The advantage of using these cooling pipes is two fold: first, to limit the risk of thermal cracking occurred during hardening of the concrete, and later, to produce an injectable gap for the grouting process to ensure plug tightness. The design of the cooling system is carried out according to the results of the numerical analysis along with specific project requirements; however, the final system of pipe arrangement is governed by availability of space and construction sequence. The standards ACI 207.4R-93 (ACI 1993) and ACI 207.2R-07 (ACI 2007) both suggest a basic methodology for calculation of flow and cooling-pipe arrangements for mass concrete structures, which is based on an old but good detailed investigation conducted during the construction of dams and other mass concreting works. In case of tunnel plug, the geometry of the plugs and the properties of the concrete allow for a maximum spacing of 800 mm as shown in Fig. 7. The required distribution of the cooling system into individual cooling coils calls for a maximum length of 60–80 m of cooling pipes. The required flow in a cooling coil is set between 20 and 27 (× 10−3) m3 /min to limit the increase of water temperature in the coil.

Fig. 7 : Plan and sectional view of cooling arrangement with cooling pipes

A typical rise of temperature inside a full 8-m plug, along with time at four different sensor locations, is shown in Figure 8.

Fig. 8 : Temperature measurement inside a tunnel plug during concrete pouring

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Number 4 December 2015

All of the sensors were placed at four locations, as shown in Fig. 8. It is observed from the figure that owing to the arrangement of the cooling pipes, for the same concrete mix, the maximum temperature in the concrete has only reached 54°C, which is less than the limits prescribed by different standards for DEF, which can result in the cracking of concrete. Soon after the concrete in the tunnel plugs is mature (28 days of hardening), the surfaces of the plugs are isolated and the cooling temperature is slowly reduced in the tunnel plugs. The reduction in temperature will result in thermal contraction of the tunnel plug and will widen the gap between rock and concrete at the top and sides of the plug. This temperature reduction has to be done slowly to avoid cracking in this phase. When the complete plug has been cooled down to the desired temperature, grouting needs to be performed. Once the grout has hardened, the plug will be restored to ambient temperature, which will result in compressed plugs. 6.3 Grouting Once the plug key is excavated per the design requirements, rock-mass grouting is performed to grout all the blast-induced cracks near the key-in surface in the excavated damaged zone. The rock mass surrounding a minimum 5 m of the plug area on all sides is grouted with higher pressure of up to 20 bars. OPC cement is used for this purpose; depending on the rock type, micro-cement may also be used. Rock-mass grout holes are spaced 0.5 m on all sides. Two consecutive stages of contact grouting are performed. Grout is injected in the gaps between the concrete plug and rock mass during cooling and once set, the cooling stops and the plug attains a normal ambient temperature. At this stage, the plug expands to its normal size and the grouted contact between the plug and rock mass are further sealed, resulting in tight plug. The first stage is carried out with ordinary Portland cement and later with microcement. The first-stage contact grouting acts more as backfill concreting and fills any large voids, such as those created in the roof and upper wall area of the plugs. These are implemented by embedded GI pipes inserted approximately 50 cm inside the rock mass. These pipes are placed at 0.5-m spacing all along the sides, similar to rockmass grout holes. Second-stage contact grouting The Bridge and Structural Engineer


starts once the first-stage grout is completed and has gained strength under secondary cooling (normally after two days). This is performed with micro cement, which penetrates the leftover small spaces of the first-round contact grout with OPC. In this, regroutable perforated injection tubes are embedded inside the plug. They are placed in three peripheral rows in contact with the rock at 1 m interspacing distance. These pipes are laid not more than 5 m in length. They are overlapped to cover the entire perimeter in each row. These pipes are made of PVC, which can withstand the high pressure created by the hardened concrete. They have an outer protective layer that prevents cement slurry entering the grout pipes during casting of the plug. Before and during second-stage contact grouting with microcement, water is injected through perforated grout-injection tubes to check the injection tube function and quantify the volume of leak and void space in the contact. This is called as tightness test and is carried out through the central tube at 3 bar pressure, as shown in Figure 9 using water. Grouting and testing is repeated as required till tightness is attained.

Fig. 9 : Sectional view of contact grouting pipe arrangement around a tunnel plug

7.

In view of uncertainty involved in the design of such large structures, a systematic and detailed approach for investigations along with the active design approach linked to construction progress is followed. Concrete plugs which are required to separate crude oil and water are designed as gas tight to prevent flow of any oil vapor outside the cavern. Because mass concreting is involved in the construction of these plugs, suitable arrangements for the cooling of concrete are carried out to prevent development of high temperatures inside the plugs along with grouting to ensure tightness.

8. References 1. ACI (American Concrete Institute). (1993). “Cooling and insulating systems for mass concrete.” ACI 207.4R-93, Farmington Hills, MI. 2. ACI (American Concrete Institute). (2007). “Report on thermal and volume change effects on cracking of mass concrete.” ACI 207.2R-07, Farmington Hills, MI. 3. Barton. N., Lien. R, Lunde, J. (1974). “Engineering classification of rock masses for the design of tunnel support, Rock Mechanics” Vol.6, N 4, pp 189-236. 4.

Indian Code IS 456. (2000). “Code of practice for plain and reinforced concrete.”

5.

Hoek, E., and Brown, E. T. (1980). “Empirical strength criterion for rock masses.” J. Geotech. Engrg. Div., 106(GT9), 1013–1035.

6.

IS 3370: Part 2. (2009). “Code of practice for design of concrete structures for storage of liquids.”

7.

Nanda A. (2012). “Design and construction of storage caverns” Keynote Lecture, Proceedings of Indian Geotechnical conference, New Delhi, India.

Discussion and Conclusions

This paper presents a study on different aspects involved in engineering and design of large underground storage caverns for storage of crude oil with specific emphasis on design of concrete plugs.

The Bridge and Structural Engineer

Volume 45 Number 4 December 2015  109


Deformability of Rock Mass for Dam Foundation

Dr. Rajbal Singh, born 1955, received M.Tech. and Ph.D. degrees in Rock Mechanics from IIT Delhi, India. He worked in CSMRS, New Delhi for 30 years. His main area of research is related to rock mechanics, insitu testing, quality control for hydropower development in India and Bhutan, numerical modelling and landslide hazards mitigation. He has published 210 papers. Rajbal Singh Former Head, Rock Mechanics Central Soil and Materials Research Station (CSMRS), New Delhi rajbal.singh@nic.in

Summary This paper deals with the deformability of rock mass by conducting in-situ rock mechanics tests with different methods. In-situ tests conducted by using plate jacking test, plate loading test, flat jack test, Goodman jack test and laboratory test have been compared with reference to large size plate jacking tests. The deformations from two dimensional analyses of a concrete gravity dam have been compared at dam top, heel and toe by varying modulus of deformation in complex geological condition. This analysis has shown the significance of modulus of deformation in numerical modelling. Due to scale effects on deformability of rock mass, large size plate jacking test must be utilized to evaluate the modulus of deformation of rock mass particularly due to jointing in rock mass. The modulus of deformation of rock mass obtained by plate loading test and Goodman jack test may be multiplied by a factor of 2.5 to arrive at a reasonably representative value. Keywords: deformability of rock mass; rock mechanics; in-situ tests; plate jacking test, plate loading; Goodman jack; laboratory.

1. Introduction Dams are constructed on complex weak rock mass foundations in the Himalayan region. It is, therefore, necessary to investigate the properties of dam foundation properly with due care. The modulus of 110â&#x20AC;&#x192; Volume 45

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deformation of rock mass is an important engineering parameter required for the stability analysis and design of dam foundation on rock mass. The deformability of rock mass is determined by conducting deformability tests on dam foundation by making drift on both abutments. Different equipment and techniques are used to arrive at the design modulus value. Since any error in the estimation of modulus values results in multiplication of its effects in the analysis, it is necessary to know about the reliability of testing equipment and procedures. The different procedures used for direct measurement provide values that often differ from one another by as much as 2 to 3 times depending mostly on the loading area used for a particular test and deformation measured at the surface or inside drill holes as discussed by Palmstrom and Singh [1]. This is inevitable, not least due to the fact that the rock mass volume concerned differs from one test to another. Also, the parameter is notably sensitive to a scale effect because of the discontinuities. Part of this may arise from the deformation characteristics of discontinuities, which are difficult to analyse. Deformability of rock mass is characterized by a modulus describing the relationship between the applied load and the resulting deformation. The fact that jointed rock masses do not behave elastically has promoted the usage of the term modulus of deformation rather than modulus of elasticity or Youngâ&#x20AC;&#x2122;s modulus. Since a rock mass contains weakness besides the intact rock material, the modulus values of the latter The Bridge and Structural Engineer


is in the order of three to twenty times higher than in situ values as discussed by Singh [2]. The difference between laboratory testing on intact rock and in situ testing of rock mass depends mainly on the joint system and in-filled material in the rock joints. The modulus of deformation of rock mass can be determined by conducting plate jacking test, plate loading test, Goodman jack test, flat jack test, cable jacking test, radial jack test, and dilatometer test. Out of these, first four methods are being used extensively in India. The results of plate loading tests, plate jacking tests and Goodman jack tests shall be discussed in this paper. Using different equipment and suggested procedures, a number of tests were conducted by the Central Soil and Materials Research Station (CSMRS), New Delhi in the same rock mass for the determination of the modulus of deformation as discussed by Singh [3, 4], Singh et al. [5], Singh and Bhasin [6], Singh and Rajvanshi [7], Singh and Dhawan [8], Sharma and Singh [9, 10] and Sharma et al. [11, 12]. This paper deals with importance and evaluation of modulus of deformation from different methods of measurement in the field by using plate jacking, plate loading, Goodman jack and flat jack tests. A comparison has also been discussed with intact rock samples tested in laboratory. Importance of determining the modulus of deformation has been shown by an example of analyses of concrete gravity dam on foundation in complex geological media.

2.

Due to scale and jointing effects, the modulus values determined in laboratory on intact rock is in the order of five to twenty times higher than in situ values of rock mass as discussed by Beiniawski [13, 14].

3.

Deformability of Rock Mass

Modulus of deformation and elasticity are required for the design of foundation of dams, tunnels, other underground structures (underground power house and nuclear waste deposits, underground oil storage) where the ground response to the normal load is desirable. Typical stress versus deformation curve recorded in a deformability test of a rock mass is shown in Fig. 1. Definitions of modulus of deformation and modulus of elasticity of rock mass has been defined by the Commission of Terminology of the International Society for Rock Mechanics (ISRM) in 1975 [15, 16] as follows (Fig. 1): Modulus of deformation of rock mass (Ed): The ratio of stress (P) to corresponding strain during loading of a rock mass, including elastic and inelastic behaviour (Wd in Fig. 1). Modulus of elasticity of rock mass (Ee): The ratio of stress (P) to corresponding strain during loading of a rock mass, including only the elastic behaviour (We in Fig. 1).

Laboratory versus In-Situ Tests

Testing in laboratory is conducted under controlled environment on intact rock sample. Small fissures are also not present in rock specimen prepared for laboratory testing after drilling, grinding and polishing. Hence, the rock specimen tested in the laboratory may be considered as the strongest part of the rock mass. However, testing in laboratory forms the basis for all engineering classification and design input parameter for numerical modelling. Field or in-situ tests are very expensive and time consuming due to site preparation by excavating test drifts or trenches in the open. There are further difficulties to conduct in-situ tests. The data needs to be analysed by experience hands. However, the results are true representation of rock mass. The Bridge and Structural Engineer

Fig. 1: Typical stress versus deformation curve recorded in a deformability test of a rock mass

4.

Methods of Measurements

The deformability of rock mass can be characterised by using following tests based on the methodology of measurements: Volume 45 Number 4 December 2015â&#x20AC;&#x192; 111


conducting tests in cycles of loading and unloading. Hence, the behaviour of loading and unloading curves in last 2 cycles is almost similar.

Surface measurement

 Plate loading test (PLT), and

 Flat jack test (FJT), Large scale testing

 Plate jacking test (PJT), and

 Radial jack test (RJT), Drill hole testing

 Goodman jack test (GJT), and

 Dilatometer test (DT). Indirect methods

 Rock Mass Quality (Q) by Barton et al. [17, 18]

 Rock Mass Rating (RMR) by Beiniawski [14]

 Geological strength index (GSI) by Hoek and Brown [19]

 Rock mass index (RMi) by by Palmstrom [20]

Based on the site conditions and availability of resources at project for test site preparation, the different methods are utilised to evaluate the modulus of deformation of rock mass. It is always suggested to utilise at least two methods in the field in addition to indirect methods. There are following reasons to conduct the deformability tests in five cycles:

Construction activities in rock mechanics for dam and underground structures are done in stages/cycles. Dam loading is applied on rock mass in stages and it takes some time due to construction to apply full loading on rock mass. Similarly, large underground structures are also excavated in stages.

Rock mass in drift is disturbed due to blasting.

Joints get closed by doing testing in cycles (opening of joint due to stress release during excavation of drift by drill and blast method).

Equipment gets adjusted during loading and unloading in first cycle. It is, therefore, suggested to repeat the first cycle second time to calculate modulus value at low stress or first cycle may be avoided for calculating modulus of deformation. In-situ rock mass conditions are created by

112  Volume 45

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Final value of modulus of deformation is obtained for in-situ rock mass from last loading/ unloading cycle. Hence, it is very important to decide the final loading in all deformability tests.

In view of above justifications, the deformability tests are conducted by using 5 cycles of loading and unloading to determine the deformability characteristics of rock mass. The final stresses expected to be transferred by structures to the rock mass must be made available much before the start of in-situ testing. The structures may be dam to be founded on rock mass and stresses to be transferred to the rock mass foundation are due to self weight and filling of reservoir including dynamic loading during earthquake. It should also include the effects of loading and unloading during operation of reservoir. The stresses may be due to unloading during excavation of tunnel and its loading during filling of hydraulic tunnels. The maximum stress may be applied equal to 1.5 to 2 times the stress due to structure. It may further be divided into five cycles of loading and unloading with application of maximum stress in fifth cycle as per ISRM [16]. The behaviour of rock mass is almost similar during fourth and fifth cycle of loading and unloading (Fig. 1). The modulus of deformation is determined based on loading in fifth cycle. The variation in modulus of deformation with applied stress is given in Table 1. In the case of Trap rock mass, the modulus of deformation increases from 2.49 GPa to 6.75 GPa as the applied stress increases from 10 MPa to 60 MPa. It is confirmed from Table 1 that the modulus value increases with the increase in applied stress during testing. Hence, it is very important to select applied stress properly during testing. Table 1: Applied stress and modulus of deformation Rock type

Applied stress MPa

Trap

10 20 40 60

Modulus of deformation GPa 2.49 3.69 5.47 6.75

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5.

Dam Foundation

As discussed earlier, most of the dam foundations are placed at complex geological strata with lot of variations in rock mass properties. It is essential to evaluate the rock mass properties with utmost care due to variations in jointing and discontinuities under the dam foundation.

gravity dam are given in Table 3 for variations of modulus of deformation of rock mass with 3 sets of data from Table 2.

One such example of complex dam foundation is Lakhwar dam with foundation geology as shown in Fig. 2. The layout plan of Lakhwar dam project is shown in Fig. 3. The right bank of the dam is quartzite slate, the centre of dam is basic rock (trap) and left bank geology is slate with interface of shear zone as gousy material. There was no problem on right bank. However, there was dam design problem due to variations in modulus values in basic rock and very low values of slate and shear zone which is an interface between trap and quartzite slate.

Where: 1. Intake, 2. Pressure shaft, 3. Drainage gallery, 4. Eraction bay, 5. Transfer hall, 6. Machine hall, 7. Control room, 8. Adit to eraction bay, 9. Adit to control room, 10. Cable tunnel, 11. Expansion chamber, 12. Tailrace tunnel, 13. Dam monolith, 14. Plunge pool, 15. Exploration drift. Fig. 3: Layout plan of Lakhwar dam project

Table 2: Material parameters for a two dimensional FEM analysis of concrete gravity dam Fig. 2: Spillway section of Lakhwar dam with foundation geology

Dam concrete Modulus of Deformation

- 20 GPa

An extensive testing was done to determine the modulus of deformation of rock mass with all available testing methods by different testing agencies. The insitu and laboratory testing was done for all rock types including shear zone material. The testing was also done under the Yamuna river bed level by excavating drifts along and across the flow of river through 65m deep shaft at dam foundation level. The position of the shaft has been marked in Fig. 3 at serial no. 15 as exploration shaft on the right bank of the river.

Poisson’s ratio

- 0.20

Unit weight

- 2.4 t/m3

The properties of rock mass for stability analysis of dam foundation are given in Table 2. The deformations from two dimensional FEM analyses of concrete

The vertical deformation at dam top is 6.73 mm in Table 3 for homogenous rock mass with modulus of 5 GPa. The vertical deformation increases to 8.89 mm

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Modulus of deformation rock mass Rock Type

Modulus of Deformation GPa Set I Set II Set III Trap 5.6 8.0 5.0 Quartzite 4.3 0.8 5.0 Slates 2.0 0.42 5.0 Gougy 0.2 0.05 5.0 Material

Poisson’s Ratio 0.235 0.235 0.235 0.235

Volume 45 Number 4 December 2015  113


for set I and 17.20 mm for set II due to large variation in modulus of deformation of different rock mass and shear zone material. The same variations were also noticed at dam heel and toe also.

values determined from Goodman jack are in horizontal direction and the moduls from PJT is in vertical direction. Some difference may be due to anisotropy in rock mass.

In view of above, it becomes very essential to determine the modulus of deformation by conducting in-situ tests for all rock mass types of geological media in dam foundation.

A comparison of large size plate jacking test with plate loading test, Goodman jack test, flat jack test (FJT) and laboratory test conducted on trap rock mass has been given in Table 4. The modulus of deformation by PJT varied from 4.11 GPa to 7.04 GPa with an average value of 5.57 GPa. The modulus of deformation by PLT varied from 0.92 GPa to 5.66 GPa with an average value of 2.90 GPa. The modulus of deformation by GJT varied from 1.46 GPa to 5.29 GPa with an average value of 2.71 GPa. The modulus of deformation by FJT varied from 1.35 GPa to 4.85 GPa with an average value of 3.17 GPa. However, the modulus of elasticity by laboratory test (LT) on intact rock specimen varied from 38.60 GPa to 153.30 GPa with an average value of 85.70 GPa. The results of PJT are 1.9, 2.1 and 1.8 times higher than PLT, GJT and FJT, respectively. However, modulus value of intact rock is 15.4 times higher than PJT value.

Table 3: Deformations from two dimensional FEM analysis of concrete gravity dam Set No.

Modulus of deformation GPa

Deformation in mm Direction

Dam Top

Dam Heel

Dam Toe

I

Trap Quartzite Slates Gougy Material

= 5.6 = 4.3 = 2.0 = 0.2

Horizontal Vertical Resultant

19.6 8.89 21.52

6.89 2.77 7.42

7.29 11.40 13.53

II

Ed for Trap Quartzite Slates Gougy Material

= 8.0 = 0.8 = 0.42 = 0.05

Horizontal Vertical Resultant

49.72 17.2 52.60

18.12 4.48 18.66

19.72 30.21 36.07

III

Ed for Trap Quartzite Slates Gougy Material

= 5.0 = 5.0 = 5.0 = 5.0

Horizontal Vertical Resultant

8.53 6.73 10.86

3.95 3.97 5.6

3.99 5.55 6.72

7. Comparison among Different Testing Methods A comparison of PJT, PLT and GJT is shown in Fig. 4 with results in Tables 4 and 5 and is also described below: In plate jacking test (PJT) with borehole extensometer measurement: the deformations are measured inside the drill hole away from the damaged zone towards the undisturbed rock mass.

Table 4: Comparison of PJT with PLT, GJT, FJT and Laboratory Test Sl. Test No. of No. type tests

Variations in modulus values (Ed), GPa

Ratio of PJT/ (PLT, GJT, FJT, Minimum Maximum Average LT)

1

PJT 4

4.11

7.04

5.57

5.57/5.57 = 1.0

2

PLT 20

0.92

5.66

2.90

5.57/2.90 = 1.9

3

GJT 11

1.46

5.29

2.71

5.57/2.71 = 2.1

4

FJT 20

1.35

4.85

3.17

5.57/3.17 = 1.8

5

LT

38.60

153.30

85.70

10

5.57/85.7 = 0.06*

*LT/PJT=85.7/5.57 = 15.4

In plate loading test (PLT) with surface measurement: larger deformations measured at the rock surface in these tests include the top damaged zone due to blasting during drift excavation.

Goodman jack test (GJT) performed inside the drill hole: gave lower values of the modulus of deformation because, in hard rock, the loading platens deform. Thus, the displacement devices record the increase in borehole diameter plus deformation of the loading plates. Further, the stress is applied on a very small area as compared to large size plate jacking test. The modulus

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Number 4 December 2015

Fig. 4: Scale effects of three main methods for in situ deformation measurement

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8. Correlations between Various Types of In-Situ Measurements The values of the modulus of deformation and modulus of elasticity from different projects are given in Table 5 by using different test procedures based on total deformation and elastic deformation, respectively. It is seen from Table 5 that the modulus values based on in situ tests vary from different project sites. It is, therefore, necessary to conduct in situ tests for determination of modulus of deformation at a particular site. The experience is that different procedures used for in situ measurements provide values that often differ from one another by as much as 100%. This is inevitable, due to the fact that the volume of rock mass structure differs from one test to another particularly in terms of degree of jointing. As the modulus is notably sensitive to the presence of joints, the rock mass conditions at each test site should be carefully described as part of the test procedure. By comparing the variations in rock mass quality, some of the difference in test results may be explained. The CSMRS has performed in situ deformation tests with the plate jacking test, plate loading tests, flat jack test and Goodman jack test during the last three decades at most of the important river valley projects in India, Nepal and Bhutan. The procedures and suggested method of the International Society of Rock Mechanics (ISRM [16]) have been closely followed for conducting all the tests. From the test results compiled from Palmstrom and Singh [1], Singh [3, 4], Singh et al. [5], Singh and Dhawan [8], Sharma et al. [11], Bieniawski [21], CSMRS [22], it has been possible to compare and correlate the in situ measurement as given in Table 5. Table 5: Ratio between plate jacking test (PJT) and other types of field deformation measurements Ratio

Measurements in hydropower projects

Suggested ratio between in situ Lakh- Jamrani Tala Bieniawski CSMRS measurements war Sand- Augen Trap stone Gneiss

PJT/PLT 1.9

Experience by

4.0

PJT/FJT 1.8 PJT/GJT 2.1

2.6

2.4

2

2-3

2.5

2-3

2.5

2-3

2.5

PJT= Plate jacking test, PLT= Plate loading test; GJT= Goodman jack test; FJT= Flat jack test

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As earlier pointed out by several researchers (Bieniawski [21]; Heuze and Amadei [23]), the value obtained by the various in situ deformation tests will not give the same deformation modulus. From the measurements carried out, the ratio between these types of deformation measurements is given in Table 5 where in the results of Bieniawski [21] is also given. Singh [3] and Palmstrom and Singh [1] concluded that the modulus of deformation determined by PLT, FJT and GJT may be multiplied by a factor of 2.5 to arrive at a reasonably good representative value. This factor may be derived exactly for a particular site by conducting in situ tests. Bieniawski [21] has stated that the flat jack test is the least reliable due to difficulties with interpretation of the results as well as the small volume of rock tested near to the rock surface. Singh [3] and Benson et al. [24] suggested that the modulus values must be obtained from large size plate jacking tests. The PJT is less sensitive to the variations in the pressure distribution and displacements are directly measured under the loaded area. The measurements of deformation in drill holes at various depths provide a check against any gross errors (blunders) of the measurements. The PJT also allows a better assessment of the properties at depth as the displacements outside the loaded area are influenced to a much greater extent by the behaviour of rock mass. The low modulus value by GJT is due to the fact that loaded area in GJT is much smaller than PJT as also concluded by Singh et al. [5] and Sharma et al. [11]. Heuze and Amadei [23] have suggested by trial and error method for improving the moduli values obtained by borehole jack method. They tried to increase the value of constant K factor suggested by Goodman et al. [25], which was also discussed by Singh et al. [5]. Beiniawski [21] tried to compare the rock deformability by GJT with Petite seismique and flat jack methods. It is, therefore, suggested that the results obtained by plate loading, flat jack and Goodman jack tests must be multiply by a factor of 2.5 to arrive at a reasonably good value of modulus of deformation of rock mass. This factor can be derived for a particular site by conducting plate jacking and Goodman jack tests simultaneously. Volume 45 Number 4 December 2015â&#x20AC;&#x192; 115


9. Conclusions

It is recommended to utilize large size plate jacking test with borehole deformation measurements to arrive at a final design value of any project. However, the modulus of deformation of rock masses obtained by plate loading tests and Goodman jack tests may have to be multiplied by a factor of 2.5 to arrive at a reasonably good representative value. This factor may be derived exactly for a particular site by conducting in situ tests.

Based on this study, the following conclusions are drawn: The modulus of deformation of rock mass is calculated by taking total deformation of loading cycle at a particular applied stress level. The modulus of elasticity of rock mass is calculated by considering deformation of unloading cycle at a particular applied stress level.

The modulus of elasticity of rock cores is 3 to 20 times higher than the modulus of deformation of rock mass. The order of magnitude of the ratio of the modulus of deformation (Ed) and modulus of elasticity of intact rock depends upon the extent of discontinuities present in the rock mass.

Modulus value increases with the increase in applied load during testing. It is, therefore, very essential to known the magnitude of loading due to structure to be constructed on or inside the rock mass. The stress due to structure may be multiplied by a factor of 1.5 to 2 times to determine application of maximum stress during test for evaluation of the deformability of rock mass.

Scale effects play an important role to determine the modulus of deformation of rock mass. Due to scale and jointing effects, the modulus values determined in laboratory on intact rock is in the order of three to twenty times higher than in situ values of rock mass.

The deformability tests are conducted in five cycles of loading and unloading to simulate the actual rock mass and maximum stress is applied in the fifth cycle.

Ratio of Ee/Ed decreases with the increase in stress level in all the methods. The decrease in moduli ratio shows the closing of joints at high stress level to create in-situ rock mass conditions. In good rock mass condition, moduli ratio becomes almost one in fifth cycle of loading and unloading.

There are variations in the modulus values determined by different methods. Sometimes these variations are due to the change in the rock mass properties also. The results of deformability measurements must be analysed by experience hands working in the field.

References 1.

Palmstrom Arild and Singh Rajbal, “The Determination of Modulus of Rock Mass: Comparison between In Situ and Indirect Estimate”, Journal of Tunnelling and Underground Space Technology, 16, 2001, pp. 115-131.

2.

Singh Rajbal, “Arma 14-7731: Scale Effects on Deformability of Rock Mass by Different Methods in Himalayas”, 48th US Symposium on Rock Mechanics, Minneapolis, 1-4 June 2014.

3.

Singh Rajbal, “Deformability of Rock Mass by Different Methods inside the Underground Desilting Chamber”, Journal of Rock Mechanics and Tunnelling Technology, 2009, Vol. 15, No. 1, pp. 37-54.

4.

Singh Rajbal, “Deformability of Rock Mass and a Comparison by Plate Jacking and Goodman Jack Tests”, International Journal of Rock Mechanics & Mining Sciences, 2011, 48, 12081214.

5.

Singh R.B., Hari Dev, Dhawan A.K. and Sharma V.M., “Deformability of Rock Mass by Plate Jacking and Goodman Jack Test”, Proc. Indian Geotechnical Conf. (IGC), 1994, pp. 385-388.

6.

Singh Rajbal and Bhasin R., “Q-system and Deformability of Rock Mass”, Proc. of Conf. on Recent Advances in Tunnelling Technology, New Delhi, 1996, pp. 57-67.

The experience obtained at one project site with the same rock type cannot be utilized at another project site with the same type of rock mass. Therefore, the deformability of rock mass must be determined by any available method.

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7.

8.

9.

Singh Rajbal and Rajvansi U.S., “Effect of Excavation on Modulus of Deformation”, Proc. of Conf. on Recent Advances in Tunnelling Technology, New Delhi, 1996, pp. 68-77. Singh Rajbal and Dhawan A.K., “Experience of deformability measurement using Goodman jack”, Proc. of Int. Conf. on Rock Engg. Techniques for Site Characterization, Bangalore, India, 1999, pp. 29-36. Sharma V.M. and SINGH R.B., “Deformability of Rock Mass”, Proc. Conf. on Application of Rock Mechanics in River Valley Projects, Roorkee, 1989, pp. II 7-12.

10. Sharma V.M. and Singh R.B., “Deformability of Rock Mass using Goodman Jack”, 56th Research & Development Session, Central Board of Irrigation and Power, 1990, pp. 99-101. 11. Sharma V.M., Singh R.B. and Chaudhary R.K., “Comparison of Different Techniques and Interpretation of the Deformation Modulus of Rock Masses”, Indian Geotechnical Conference (IGC-89), Vishakhapatnam, 1989, Vol. 1: 439443. 12. Sharma V.M., Singh R.B. and Chaudhary R.K., “Rock Mechanics Investigation of a Water Resources Project”, Proc. Trend in Geotechnical Investigations in Last Twenty Five Years, Indian Society of Engineering Geology, Calcutta, 1990, pp. 177-186. 13. Bieniawski Z.T., “Engineering Classification of Jointed Rock Masses”, Trans. S. Afr. Instn. Civil Engrs., 15, 1973, 335-342. 14. Bieniawski Z.T., “Determining Rock Mass Deformability Experience from Case Histories”, Int. J. Rock Mech. Mineral Science & Geomechanics Abst., Vol. 15, 1978, 237-247. 15. ISRM: Commission on terminology, symbols and graphic representation, International Society for Rock Mechanics (ISRM), 1975. 16. ISRM, “Suggested Methods for Determining In Situ Deformability of Rock”, Int. J. Rock Mech.

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Min. Sci. & Geomech. Abstracts, Vol. 16, No 3, 1979, pp 195 214. 17. Barton, N., R. Lien and J. Lunde, “Engineering Classification of Rock Masses for the Design of Tunnel Support”, Rock Mechanics, 6, 1974, 189-236. 18. Barton, N., “Application of Q-system and Index Tests to Estimate Shear Strength and Deformability of Rock Masses”, Proc. Int. Symp. Eng. Geol. Underground Constr., A.A. Balkema, Boston, 1983, 51-70. 19. Hoek, E. and Brown E.T., “Practical Estimates of Rock Mass Strength”, International Journal of Rock Mechanics & Mining Sciences, 34 (8), 1997, 1165-1186. 20. Palmstrom A., “RMi - a Rock Mass Characterization System for Rock Engineering Purposes”, Ph.D. Thesis Univ. of Oslo, 1995, 400 p. 21. Bieniawski Z.T. (1979) A comparison of rock deformability measurement by Petite Seismic, the Goodman Jack and Flat Jacks, Rapid Excavation and Tunnelling Conference, Atlanta. 22. CSMRS (1999) Report on deformability of rock mass in desilting chamber in drift DR-2 at Tala H. E. Project, Bhutan, Central Soil and Materials Research Station (CSMRS), New Delhi. 23. Heuze F.E. and Amadei B. “NX Borehole Jack: A Lesson in Trial and Errors”, Int. J. Rock Mech. Min. Sci. & Geomech. Abstracts, Vol 22, No 2, 1985. 24. Benson R.P., Murphy D.K. and Mccreath D.R., “Modulus Testing of Rock at the Churchill Falls Underground Powerhouse, Labrador”, Determination of the In situ Modulus Deformation of Rock, ASTM STP 477, 1970, 89116. 25. Goodman R.E., VAN T.K. and Heuze F.E., “Measurement of Rock Deformability in Boreholes”, Proc. 10th U.S. Symp. on Rock Mechanics, AIME, New York, 1972, pp. 523-555.

Volume 45 Number 4 December 2015  117


Disaster Mitigation and Role of Civil Engineers Dr. R. KUBERAN Senior Editor Civil Engineering & Construction Review New Delhi, INDIA rkuberan@gmail.com

Dr. R. Kuberan received his PhD in civil engineering from the Indian Institute of Technology, New Delhi, India. He worked for the Government of India for twenty years carrying out geotechnical investigations for water resources projects before availing voluntary retirement. He worked for UNDP in Vietnam on flood control structures and disaster management.

Summary

2.

Basic concepts of disaster management have been briefly explained in this article. Some of the hazards that affect civil engineering structures have been covered. Specific information on the recent disaster in Uttarakhand has been covered. Disaster management activities performed during the floods in Uttarakhand is narrated and the concept of disaster management cycle has been explained. The importance of disaster mitigation and the role of civil engineers are explained for various hazards.

In June 2013, a multi-day cloudburst centred on the North Indian state of Uttarakhand caused devastating floods and landslides in the country's worst natural disaster since the 2004 tsunami. Though parts of Himachal Pradesh, Haryana, Delhi and Uttar Pradesh in India, some regions of Western Nepal, and some parts of Western Tibet also experienced heavy rainfall, over 95% of the casualties occurred in Uttarakhand. As of 16 July 2013, according to figures provided by the Uttarakhand government, more than 5,700 people were "presumed dead." This total included 934 local residents.

Keywords: Disaster management, disaster mitigation, hazard, earthquake, flood, cyclone, landslide.

Tragedy in Uttarakhand

1. Introduction Disasters are as old as human history but the dramatic increase and the damage caused by them in the recent past have become a cause of national and international concern. Over the past decade, the number of natural and manmade disasters has climbed inexorably. From 1994 to 1998, reported disasters average was 428 per year but from 1999 to 2003, this figure went up to an average of 707 disaster events per year showing an increase of about 60 per cent over the previous years. The biggest rise was in countries of low human development, which suffered an increase of 142 per cent. The scenario in India is no different from the global context. The super cyclone of Orissa (1999), the Gujarat earthquake (2001) and the Tsunami (2004) affected millions across the country leaving behind a trail of heavy loss of life, property and livelihood. Major disasters cause colossal impact on the community. 118â&#x20AC;&#x192; Volume 45

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Destruction of bridges and roads left about 100,000 pilgrims and tourists trapped in the valleys leading to three of the four Hindu Chota Char Dham pilgrimage sites. The Indian Air Force, the Army and paramilitary troops evacuated more than 110,000 people from the flood ravaged area. The Bridge and Structural Engineer


2.1 Rainfall From 14 to 17 June 2013, the Indian state of Uttarakhand and adjoining area received heavy rainfall, which was about 375 per cent more than the benchmark rainfall during a normal monsoon. This caused the melting of Chorabari Glacier at the height of 3800 metres, and eruption of the Mandakini River which led to heavy floods near Gobindghat, Kedar Dome, Rudraprayag district, Uttarakhand, Himachal Pradesh and Western Nepal, and acute rainfall in other nearby regions of Delhi, Haryana, Uttar Pradesh and some parts of Tibet.

The upper Himalayan territories of Himachal Pradesh and Uttarakhand are full of forests and snow-covered mountains and thus remain relatively inaccessible. They are home to several major and historic Hindu and Sikh pilgrimage sites besides several tourist spots and trekking trails. Heavy rainfall for four consecutive days as well as melting snow aggravated the floods. Warnings by the India Meteorological Department predicting heavy rains were not given wide publicity beforehand, causing thousands of people to be caught unawares, resulting in huge loss of life and property. In the city of Dehra Dun, capital of Uttarakhand, this was the wettest June day for over five decades. 2.2 Damages Landslides, due to the floods, damaged several houses and structures, killing those who were trapped. The heavy rains resulted in large flash floods and massive landslides. Entire villages and settlements such as Gaurikund and the market town of Ram Bada, a transition point to Kedarnath, have been obliterated, while the market town of Sonprayag suffered heavy damage and loss of lives. Pilgrimage centres in the The Bridge and Structural Engineer

region, including Gangotri, Yamunotri, Kedarnath and Badrinath, the hallowed Hindu Chardham pilgrimage centres, are visited by thousands of devotees, especially after the month of May onwards. Over 70,000 people were stuck in various regions because of damaged or blocked roads. People in other important locations like the Valley of flowers, Roopkund and the Sikh pilgrimage centre Hemkund were stranded for more than three days. National Highway 58, an important artery connecting the region was also washed away near Jyotirmath and in many other places. Because summers have more number of tourists, the number of people impacted is substantial. For more than three days, stranded pilgrims and tourists were without rations or water survived on little food. The roads were seriously damaged at more than 450 places, resulting in huge traffic jams, and the floods caused many cars and other vehicles to be washed away. On June 18, more than 12,000 pilgrims were stranded at Badrinath, the popular pilgrimage centre located on the banks of the Alaknanda River. Rescuers at the Hindu pilgrimage town of Haridwar on the river Ganga recovered bodies of 40 victims washed down by the flooded rivers as of June 21 2013. Bodies of people washed away in Uttarakhand were found in distant places like Bijnor, Allahabad and Bulandshahr in Uttar Pradesh. 2.3 Damage at Kedarnath town Situated in the Himalayan ranges, the town that grew around the Kedarnath Temple one of the famous temples of Shiva in India and part of the Char Dham yatra, was damaged by the floods and landslides caused by heavy rain. Although the Kedarnath Temple itself had not been damaged, its base was inundated with water, mud and boulders from the landslide, damaging its perimeter. Many hotels, rest houses and shops around the temple in Kedarnath township were destroyed, resulting in several casualties. Most of the destruction at Kedarnath was caused by a sudden rapid melting of ice and snow on the Kedarnath Mountain, 6 km from the temple, which flooded the Charbari lake (upstream) and then Kedarnath. Temple was flooded with water resulting in several deaths due to drowning and panicdriven stampede. The Uttarakhand Government announced that due to the extensive damage to the infrastructure, the temple will be temporarily closed to regular pilgrims and tourists for a year or two, Volume 45 Number 4 December 2015â&#x20AC;&#x192; 119


but the temple rituals will still be maintained by priests. Even after a week, dead bodies had not been removed from Kedarnath town, resulting in water contamination in the Kedarnath valley and villagers who depend on spring water suffered various types of health problems like fever, diarrhoea. When the flood receded, satellite images showed one new stream at Kedarnath town.

3.

Rescue & Relief

The Army, Air Force, Navy, Indo-Tibetan Border Police (ITBP), Border Security Force, National Disaster Response Force (NDRF), Public Works Department and local administrations worked together for quick rescue operations. Several thousand soldiers were deployed for the rescue missions. Activists of political and social organizations are also involved in the rescue and management of relief centres. The national highway and other important roads were closed to regular traffic. Helicopters were used to rescue people, but due to the rough terrain, heavy fog and rainfall, manoeuvring them was a challenge. By 21 June 2013, the Army had deployed 10,000 soldiers and 11 helicopters, the Navy had sent 45 naval divers, and the Air force had deployed 43 aircraft including 36 helicopters. From 17 June to 30 June 2013, the IAF airlifted a total of 18,424 people - flying a total of 2,137 sorties and dropping/ landing a total of 3,36,930 kg of relief material and equipment. On June 25, one of 3 IAF Mil Mi-17 rescue helicopters returning from Kedarnath, carrying 5 Air Force Officers, 9 of the NDRF, and 6 of the ITBP crashed on a mountainous slope near Gauri Kund, killing all on board. The deceased soldiers were given a ceremonial Guard of honour by Home minister Sushilkumar Shinde at a function organised by the Uttarakhand State Government. Prime Minister of India undertook an aerial survey of the affected areas and announced ` 1,000 crore aid package for disaster relief efforts in the state. Several state governments announced financial assistance, with Uttar Pradesh Government pledging ` 25 crore, the governments of Haryana, Maharashtra and Delhi ` 10 crore each, the governments of Tamil Nadu, Odisha, Gujarat, Madhya Pradesh and Chhattisgarh ` 5 crore each. The US Ambassador to India extended a financial help of USD $150,000 through the United 120â&#x20AC;&#x192; Volume 45

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States Agency for International Development (USAID) to the NGOs working in the area and announced that the US will provide further financial aid of USD $75,000. The Government of India also cancelled 9 batches, or half the annual batches of the Kailash-Mansarovar Yatra, a Hindu pilgrimage. The Chardham Yatra pilgrimage, covering Gangotri, Yamunotri, Kedarnath and Badrinath was cancelled for 2 years to repair damaged roads and infrastructure, according to the Uttarakhand Government. Government agencies and priests of Kedarnath temple were planning mass cremation of the hundreds of victims, after one week of tragedy. Local youths from several affected villages near Gangotri helped stranded tourists and pilgrims, by sending messages to their places and by providing food. Rescuers also retrieved approximately ` 1 crore and other jewellery from local persons, including some sadhu babas, who reportedly collected it from a destroyed building of a Bank and damaged shops.

4. Disaster Mitigation & Management in Uttarakhand Disaster Mitigation & Management Centre is working as autonomous institute under aegis of Department of Disaster Management Government of Uttarakhand and Disaster Mitigation and Management Centre (DMMC) is the apex centre in the field of Disaster Mitigation & Management in Uttarakhand, for the protection of the community and the environment from the overwhelming obliteration caused by disasters. DMMC, located in the Uttarakhand secretariat compound, is to generate the sense of worth amongst common people and the government authorities in formulating appropriate policies and strengthening their capabilities to cope up with all aspects of disaster management. In addition to offering an extensive range of training programs, gearing-up to providing advance information about likely disaster through latest technologies available for the purpose, maintaining a network of experienced experts working in the field and institutions of excellence, DMMC will also provide consultancy services to all levels of government, international agencies and nongovernment organizations. The centre has also The Bridge and Structural Engineer


undertaken the responsibility of training communities and community based organizations and through them develop a strong regional knowledge base towards disaster policy, prevention mechanisms, mitigation measures, preparedness, and response plans. Perhaps the most important need at the State level is to strengthen and develop capacity to undertake disaster mitigation strategies. Disaster vulnerability assessment is to be incorporated in the state development process so that projects and future investments reduce, rather than increase vulnerability. In order to overcome resource constraints and to be effective and sustainable, the action plan for disaster reduction is to be incorporated in the overall economic and social development plans. No matter what lossreduction strategy is used, major reductions in losses of life and property come only when the emphasis shifts from reaction to anticipation. That is, the emphasis must be on proactive pre-disaster measures rather than post-disaster response. Disaster Mitigation and Management Centre (DMMC) expected to function as a think-tank for the Ministry/Department, will look into, and incorporate prevention, preparedness and mitigation aspects for all projects.

5.

Disaster Management Cycle

Disaster Risk Management includes sum total of all activities, programmes and measures which can be taken up before, during and after a disaster with the purpose to avoid a disaster, reduce its impact or recover from its losses. The three key stages of activities that are taken up within disaster risk management are: –

Before a disaster (pre-disaster). Activities taken to reduce human and property losses caused by

The Bridge and Structural Engineer

a potential hazard. For example carrying out awareness campaigns, strengthening the existing weak structures, preparation of the disaster management plans at household and community level etc. Such risk reduction measures taken under this stage are termed as mitigation and preparedness activities. – During a disaster (disaster occurrence). Initiatives taken to ensure that the needs and provisions of victims are met and suffering is minimized. Activities taken under this stage are called emergency response activities. – After a disaster (post-disaster). Initiatives taken in response to a disaster with a purpose to achieve early recovery and rehabilitation of affected communities, immediately after a disaster strikes. These are called as response and recovery activities. Disaster Risk Reduction can take place in the following ways: Preparedness: This protective process embraces measures which enable governments, communities and individuals to respond rapidly to disaster situations to cope with them effectively. Preparedness includes the formulation of viable emergency plans, the development of warning systems, the maintenance of inventories and the training of personnel. It may also embrace search and rescue measures as well as evacuation plans for areas that may be at risk from a recurring disaster. Preparedness therefore encompasses those measures taken before a disaster event which are aimed at minimizing loss of life, disruption of critical services, and damage when the disaster occurs. Mitigation: Mitigation embraces measures taken to reduce both the effect of the hazard and the vulnerable conditions to it in order to reduce the scale of a future disaster. Therefore mitigation activities can be focused on the hazard itself or the elements exposed to the threat. Examples of mitigation measures which are hazard specific include water management in drought prone areas, relocating people away from the hazard prone areas and by strengthening structures to reduce damage when a hazard occurs. In addition to these physical measures, mitigation should also aim at reducing the economic and social vulnerabilities of potential disasters. Volume 45 Number 4 December 2015  121


6.

Disaster Mitigation

It has been made clear that disaster management includes mitigation, which covers the measures taken to reduce the effect of hazards as well as the vulnerable conditions. That means effective mitigation efforts may be useful to avoid a hazardous event to become a disaster. Mitigation refers to all actions taken before a disaster to reduce its impacts, including preparedness and long-term risk reduction measures. Mitigation activities fall broadly into two categories: – –

Structural mitigation – construction projects which reduce economic and social impacts. Non-structural activities – policies and practices which raise awareness of hazards or encourage developments to reduce the impact of disasters.

Mitigation includes reviewing building codes; vulnerability analysis updates; zoning and land-use management and planning; reviewing of building use regulations and safety codes; and implementing preventative health measures. Mitigation can also involve educating businesses and the public on simple measures they can take to reduce loss or injury, for instance fastening bookshelves, water heaters, and filing cabinets to walls to keep them from falling during earthquakes. Ideally, these preventative measures and public education programmes will occur before the disaster. The primary focus of disaster management is to prevent disasters wherever possible or to mitigate those which are inevitable. Four sets of tools that could be used to prevent or mitigate disasters include: –

Hazard management and vulnerability reduction

Economic diversification

Political intervention and commitment

Public awareness

The first two apply exclusively to disasters caused by natural phenomena while the latter are used to mitigate any other hazards.

7.

Role of Civil Engineers

In order to understand the role of civil engineers, we should first understand the effects of various hazards on different civil engineering structures and civil engineering activities. Then the role of civil engineers will be in taking all necessary steps in mitigation. Similarly, the role of geotechnical engineers will be in taking all necessary measures in mitigating the effect of a hazard. 7.1 Earthquake India falls quite prominently on the 'AlpineHimalayan Belt'. This belt is the line along which the Indian plate meets the Eurasian plate. This being a convergent plate, the Indian plate is thrusting underneath the Eurasian plate at a speed of 5 cm per year. The movement gives rise to tremendous stress which keeps accumulating in the rocks and is released from time to time in the form of earthquakes. Due to ground shaking, damage occurs to human settlement, buildings, structures and infrastructure, especially bridges, elevated roads, railways, water 122  Volume 45

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towers, pipelines, electrical generating facilities. Aftershocks of an earthquake can cause much greater damage to already weakened structures. Secondary effects include fires, dam failure and landslides which may block water ways and also cause flooding. Damage may occur to facilities using or manufacturing dangerous materials resulting in possible chemical spills. There may also be a breakdown of communication facilities. The effect of an earthquake is diverse. There are large number of casualties because of the poor engineering

design of the buildings and close proximity of the people. About 95 per cent of the people who are killed or who are affected by the earthquake is because of the building collapse. There is also a huge loss to the public health system, transport and communication and water supply in the affected areas. In earthquake-prone areas, buildings need to be designed and constructed as per the building by laws to withstand ground shaking. Architectural and engineering inputs need to be put together to improve building design and construction practices. The soil type needs to be analysed before construction. Building structures on soft soil should be avoided. Buildings on soft soil are more likely to get damaged even if the magnitude of the earthquake is not strong. Similar problems persist in the buildings constructed on the river banks which have alluvial soil. India has been classified into four zones of earthquake hazard risk. Earlier, five zones were considered, and as earthquakes occurred even in areas that were considered to be safe from earthquake hazard, the revision was made some years ago. A location that was considered to be under ‘Zone-3’ in the old code is now coming under ‘Zone-4’. That means higher earthquake forces will have to be considered for designing a structure in this area. Also, all important large structures existing in this area should be analysed and checked whether they will be able to withstand the earthquake forces due to the new guidelines. Failure in doing so may result in damage to buildings as witnessed in Ahmedabad in 2001 during the Gujarat earthquake on January 26 by failure of concrete columns supporting apartment buildings. 7.2 Tsunami Tsunami wave train may be caused by an undersea earthquake, landslide or volcanic eruption. Whatever the cause may be sea water is displaced with a violent motion and swells up, ultimately surging over land with great destructive power. While it is of course not possible to prevent a tsunami, in certain tsunami prone countries some measures have been taken to reduce the damage caused on shore. Japan has implemented an extensive programme of building tsunami walls of up to 4.5m high in front of populated coastal areas. Other localities have built flood gates and channels to redirect the water from incoming tsunamis.

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However, their effectiveness has been questioned, as tsunamis are often higher than the barriers. For instance, the tsunami which hit the island of Hokkaido on July 12, 1993 created waves as much as 30m tall - as high as a 10-story building. The port town of Aonae on Hokkaido was completely surrounded by a tsunami wall, but the waves washed right over the wall and destroyed all the wood-framed structures in the area. The wall may have succeeded in slowing down and moderating the height of the tsunami but it did not prevent major destruction and loss of life. Most of the habitation of the fishing community is seen in the coastal areas. The houses constructed by them are mainly of lightweight materials without any engineering inputs. Therefore there is an urgent need to educate the community about the good construction practices that they should adopt such as: –

Site selection – Avoid building or living in buildings within several hundred feet of the coastline as these areas are more likely to experience damage from tsunamis.

Construct the structure on a higher ground level with respect to mean sea level.

Elevate coastal homes: Most tsunami waves are less than 3 metres in height. Elevating house will help reduce damage to property from most tsunamis.

Construction of water breakers to reduce the velocity of waves.

Use of water and corrosion resistant materials for construction.

– Construction of community halls at higher locations, which can act as shelters at the time of a disaster. 7.3 Cyclone Cyclone is a region of low atmospheric pressure surrounded by high atmospheric pressure resulting in swirling atmospheric disturbance accompanied by powerful winds blowing in anticlockwise direction in the Northern Hemisphere and in the clockwise direction in the Southern Hemisphere. They occur mainly in the tropical and temperate regions of the world. On 29th October 1999, Super-cyclone with wind speed of 260-300 km/hour hit the 140 kilometre coast 124  Volume 45

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of Orissa with a storm surge created in the Bay-ofBengal with water level 9 metres higher than normal. The super storm travelled more than 250 km inland and within a period of 36 hrs ravaged more than 200 lakh hectares of land, devouring trees and vegetation, leaving behind a huge trail of destruction. The violent cyclone was merciless and broke the backbone of Orissa’s economy and killed thousands and devastated millions. In cyclone-prone areas, structures need to be built to withstand wind forces. Good site selection is also important. Majority of the buildings in coastal areas are built with locally available materials and have no engineering inputs. Good construction practice should be adopted such as: –

Cyclonic wind storms inundate the coastal areas. It is advised to construct on stilts or on earth mound.

– Houses can be strengthened to resist wind and flood damage. All elements holding the structures need to be properly anchored to resist the uplift or flying off of the objects. For example, avoid large overhangs of roofs, and the projections should be tied down. –

A row of planted trees will act as a shield. It reduces the energy.

Buildings should be wind and water resistant.

– Buildings storing food supplies must be protected against the winds and water. –

Protect river embankments.

– Communication lines should be installed underground. –

Provide strong halls for community shelter in vulnerable locations.

7.4 Flood Flood is a state of high water level along a river channel or on the coast that leads to inundation of land, which is not usually submerged. Floods may happen gradually and also may take hours or even happen suddenly without any warning due to breach in the embankment, spill over, heavy rains etc. The most important consequence of floods is the loss of life and property. Structures like houses, bridges; roads etc. get damaged by the gushing water, landslides The Bridge and Structural Engineer


triggered on account of water getting saturated, boats and fishing nets get damaged. There is huge loss to life and livestock caused by drowning. Lack of proper drinking water facilities, contamination of water (well, ground water, piped water supply) leads to outbreak of epidemics, diarrhoea, viral infection, malaria and many other infectious diseases.

to the population in the area at risk. In areas where people already have built their settlements, measures should be taken to relocate to better sites so as to reduce vulnerability. No major development should be permitted in the areas which are subjected to high flooding. Important facilities like hospitals, schools should be built in safe areas. In urban areas, water holding areas can be created like ponds, lakes or lowlying areas. Construction of engineered structures in the flood plains and strengthening of structures to withstand flood forces and seepage: The buildings should be constructed on an elevated area. If necessary build on stilts or platform.

Flooding also leads to a large area of agricultural land getting inundated as a result there is a huge crop loss. This results in shortage of food, and animal fodder. Floods may also affect the soil characteristics. The land may be rendered infertile due to erosion of top layer or may turn saline if sea water floods the area. Mapping of the flood prone areas is a primary step involved in reducing the risk of the region. Historical records give the indication of the flood inundation areas and the period of occurrence and the extent of the coverage. Warning can be issued looking into the earlier marked heights of the water levels in case of potential threat. In the coastal areas the tide levels and the land characteristics will determine the submergence areas. Flood hazard mapping will give the proper indication of water flow during floods.

Land use control will reduce danger of life and property when waters inundate the floodplains and the coastal areas. The number of casualties is related The Bridge and Structural Engineer

Flood Control aims to reduce flood damage. This can be done by decreasing the amount of runoff with the help of reforestation (to increase absorption could be a mitigation strategy in certain areas), protection of vegetation, clearing of debris from streams and other water holding areas, conservation of ponds and lakes etc. Flood Diversion includes levees, embankments, dams and channel improvement. Dams can store water and can release water at a manageable rate. But failure of dams in earthquakes and operation of releasing the water can cause floods in the lower areas. Flood Proofing reduces the risk of damage. Measures include use of sand bags to keep flood water away, blocking or sealing of doors and windows of houses etc. Houses may be elevated by building on raised land. Buildings should be constructed away from water bodies. 7.5 Landslide The term ‘landslide’ includes all varieties of mass movements of hill slopes and can be defined as the downward and outward movement of slope forming materials composed of rocks, soils, artificial fills or combination of all these materials along surfaces of separation by falling, sliding and flowing, either slowly or quickly from one place to another. Although the landslides are primarily associated with mountainous terrains, these can also occur in areas where an activity such as surface excavations for highways, buildings and open pit mines takes place. They often take place in conjunction with earthquakes, floods and volcanoes. At times, prolonged rainfall causing landslide may block the flow of river for quite some time. The formation Volume 45 Number 4 December 2015  125


Table 1: Types of Landslides

of river blocks can cause havoc to the settlements downstream on its bursting. Landslides constitute a major natural hazard in our country, which accounts for considerable loss of life and damage to communication routes, human settlements, agricultural fields and forest lands. The Indian subcontinent, with diverse physiographic, seismic, tectonic and climatological conditions is subjected to varying degree of landslide hazards; the Himalayas including North-eastern mountains ranges being the worst affected, followed by a section of Western Ghats and the Vindhyas. Removal of vegetation and toe erosion has also triggered slides. Torrential rainfall on the deforested slopes is the main factor in the Peninsular India namely in Western Ghat and Nilgiris. Human intervention by way of slope modification has added to this effect. Possible risk reduction measures include the following: Hazard mapping locates areas prone to slope failures. This will help to avoid building settlements in such areas. These maps will also serve as a tool for mitigation planning. Land use practices such as: –

Areas covered by degraded natural vegetation in upper slopes are to be afforested with suitable species.

Existing patches of natural vegetation (forest and natural grass land) in good condition should be preserved.

Any developmental activity initiated in the area should be taken up only after a detailed study of the region has been carried out.

In construction of roads, irrigation canals etc. proper care is to be taken to avoid blockage of natural drainage.

Total avoidance of settlement in the risk zone should be made mandatory.

Relocate settlements and infrastructure that fall in the possible path of the landslide.

No construction of buildings in areas beyond a certain degree of slope.

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Type of material Bedrock

Type of movement

Falls Slides

Engineering soils Predominantly Predominantly fine coarse

Rockfall

Earth fall

Debris fall

Topples

Rock topple Earth topple

Debris topple

Rotational

Rock slump Earth slump

Debris slump

Translational Few units

Rock block Earth block slide slide

Debris block slide

Rock slide

Debris slide

Many units Lateral spreads

Earth slide

Rock spread Earth spread

Debris spread

Rock flow

Debris flow

Earth flow

Rock avalanche

Flows

(Deep creep) Complex and compound

Debris avalanche (Soil creep)

Combination in time and/or space of two or more principal types of movement

Slope stabilisation methods in rock or in earth, can be collocated into three types of measure: –

Geometric methods, in which the geometry of the hillside is changed (in general the slope);

Hydrogeological methods, in which an attempt is made to lower the groundwater level or to reduce the water content of the material;

Chemical and mechanical methods, in which attempts are made to increase the shear strength of the unstable mass or to introduce active external forces (e.g. anchors, rock or ground nailing) or passive (e.g. structural wells, piles or reinforced ground) to contrast the destabilising forces.

Retaining Walls can be built to stop landfrom slipping. These are constructed to prevent smaller sized and secondary landslides that often occur along the toe portion of the larger landslides. The surface drainage control works are implemented to control the movement of landslides accompanied by infiltration of rain water and spring flows. Engineered structures with strong foundations can withstand or take the ground movement forces. Underground installations (pipes, cables, etc.) should be made flexible to move in order to withstand forces caused by the landslide. Increasing vegetation cover is the cheapest and most effective way of arresting landslides. This helps to bind the top layer of the soil with layers below, while The Bridge and Structural Engineer


preventing excessive run-off and soil erosion.

10. Conclusions

Insurance will assist individuals whose homes are likely to be damaged by landslides or by any other natural hazards.

Civil Engineers should play major role in disaster mitigation by investigating, designing and constructing structures taking into account the potential of hazards that may take place and take appropriate measures to mitigate the same. Though the engineering codes and standards provide sufficient guidelines for this, conscious efforts should be made while dealing with structures in disaster-prone areas. Awareness and training should be imparted to all stakeholders. Experience sharing should take place through workshops and conferences.

8.

What can technical institutions do?

Technical Universities and Engineering Colleges can contribute in the following manner for disaster management. –

Provide disaster awareness to staff

– Conduct hazard-vulnerability analysis of the territory –

Prepare disaster management plans for university & college campuses

Set up & train task forces in campuses

Take up community projects

Provide training to prepare disaster management plans for state, districts, towns, and villages

Organize seminars, workshops, conferences for knowledge dissemination

Create a centre/laboratory & pursue research & development in specific areas

Take lab research results to field & bring field experience back to lab

Liaise with NGOs to develop field-based trainings

9.

What can civil engineers do?

Contribution of Civil Engineers can be in the following ways. – Thoroughly investigate and safely design foundations & structures –

Use quality materials & safe practices

Ensure disaster preparedness

Develop disaster management plan for large premises

Update knowledge on disaster safe construction

– Share knowledge & experience with fellow engineers –

Train contractors & workers on disaster safe construction

– Contribute under CSR for disaster resilient society The Bridge and Structural Engineer

The recent Uttarakhand disaster as well as similar events that happen from time to time is eye opener for the governments that mitigation activities should be systematically planned and implemented. Mitigation activities need to be planned and implemented when there is no disaster in that area. In the case of Uttarakhand, making the river banks stable in critical locations where villages are situated will be one such mitigation activity. In order to contain large floods, reservoirs of adequate capacity should be planned. In earthquake prone zones, all major structures should be checked for their earthquake stability and corrective measures like retrofitting should be taken to make them earthquake safe. It is time all governments of the states as well as the central government to take up mitigation planning very seriously. It should be planned to recover at least part of the cost spent on these mitigation projects. They can think about various business models like BOOT for implementing such projects on the similar lines of roads projects. Though the occurrence of a hazard cannot be avoided, by taking adequate mitigation measures it is possible to make sure that the hazard does not turn into a major disaster.

11. References 1.

http://en.wikipedia.org/wiki/2013_North_ India_floods

2.

http://www.dmmc.uk.gov.in/: Official website of Disaster Mitigation and Management Centre Government of Uttarakhand

3. Introduction to Disaster Management, Virtual University for Small States of the Commonwealth

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AN EXPERIMENTAL STUDY ON THE BEHAVIOUR OF STEEL PLATEANCHOR ASSEMBLY EMBEDDED IN CONCRETE UNDER CYCLIC LOADING

Deepak K. SAHU Student IIT Kanpur, Kanpur India

Saiwal KRISHNA Ph.D. Student IIT Kanpur, Kanpur India saiwal@iitk.ac.in

S.K. CHAKRABARTI Professor IIT Kanpur, Kanpur India chakra@iitk.ac.in

D.K. Sahu received his bachelor degree in Civil Engineering in 2002 from CET, OUAT (Bhubaneswar). He received his M. tech. degree in Civil engineering in 2004 from IIT Kanpur with specialization in Structural Engineering. From 2004 to 2012 he has worked in the power sector in various organisations. He has vast experience in conceptualisation, project development, design, execution and project management. Since 2012 he is running own design consultancy with primary focus in renewable energy sector.

Saiwal Krishna received his B.Tech degree in Civil Engineering from NIT Jalandhar in 2011. He received his M.Tech degree in Civil Engineering from IIT Kanpur in 2014 with specialization in Structural Engineering.

S.K. Chakrabarti received his Bachelor degree in Civil Engineering in 1972 from Bengal Engineering College (University of Calcutta), and, both Masters degree in Civil Engineering and Doctoral degree in Civil Engineering from University of Arizona, USA in 1983 and 1987, respectively. During the periods 1973-1981 and 19871988, he worked in industry (design of nuclear and thermal power plant structures) in India and USA. He held the position of Professor (2008-2010) in University of KwaZulu-Natal, South Africa, and the positions of Visiting Professor (2002) and Visiting Associate Professor (1998) in Asian Institute of Technology (Bangkok, Thailand) and Tribhuvan University (Nepal), respectively. Presently, he is a Professor of Civil Engineering at IIT Kanpur.

Summary An experimental study has been undertaken to investigate the behaviour of steel plate-anchor 128â&#x20AC;&#x192; Volume 45

Number 4 December 2015

assembly embedded in concrete under reverse cyclic loading. Three full scale specimens were tested under reverse cyclic eccentric load. Only the relative stiffness of plate and anchors were altered keeping all other The Bridge and Structural Engineer


parameters same, to study the behaviour in different failure modes. The displacement and strains at some critical points were recorded. Load-displacement and moment-rotation relationships were studied based on the test results. The capacities of the assemblies under cyclic loading have been compared with that under monotonic loading. Finally few conclusions were made regarding the behaviour of these connection systems under different failure modes. Keywords: Embedded plates; plate-anchor assembly; steel plate; steel anchor; moment resistant anchorage; moment resistant base plate

1. Introduction Steel and concrete are the two basic materials used in Civil Engineering construction. Each of the two materials has its own advantages and disadvantages and often an optimal solution is found by combining both materials giving rise to “composite steel-concrete construction” or “mixed construction”. In mixed construction separate steel and concrete members are efficiently combined to build structures. The connections between the steel and concrete members in mixed construction have always been a cause of concern for Civil Engineers due to lack of proper design guidelines for predicting behaviour of the connections between the steel and concrete members.

the plate-anchor assemblies are subjected to different types of dynamic loads such as vibrations or seismic inertias. However, from past experiences it has been observed that there is a common trend among designers to design such connections for dynamic loading conditions by considering an equivalent static load approach. The actual behaviour of the connections under the dynamic loads is rarely taken into consideration in such an approach. As a result of this, the higher values of deformations and resultant forces developing in different components of the plate-anchor assemblies due to the dynamic nature of the loads may be neglected. Therefore, due consideration should be given to the large cycles of displacements and internal forces to which the plateanchor assemblies may be subjected. Hence, there is an indispensable need to study the actual behaviour of these systems under dynamic loads to facilitate the development of a more meaningful and effective design approach for the same. The work done here is one of the attempts at studying the behaviour of plate-anchor assemblies under cyclic loads. From a review of past studies done in this field, presented below, it is evident that lot of research work has been done for individual components, however, there is limited amount of work done on the behaviour of plate-anchor assembly.

Steel plate-anchor assemblies embedded in concrete are one of the commonly used connection systems used for connecting steel and concrete members. They are frequently used to support various types of equipment and machinery from reinforced concrete structures (Fig. 1). The assembly consists of castin-place steel plate, which is so positioned in the concrete form-work prior to concreting, that its one surface gets flushed with the concrete surface after concreting. Anchors welded to the other surface of the plate remain embedded in concrete. The steel member, which has to transfer load to the concrete member, is connected to the exposed plate surface. Thus the load from the steel member gets transferred to concrete through the plate anchor assembly.

2.

Depending upon the members they connect, the plate-anchor assemblies may be subjected to different types of loads such as transverse loads, shear forces, bending moments or any combination of these. It is not necessary that these loads are always static in nature. In fact, in large number of real-life situations

Salmon et al. [2] analytically studied the moment resisting capabilities and moment resisting characteristics of column anchorages keeping the axial load constant. The effects of anchor location, axial load, characteristic strength of the base, and free length of the anchors on moment resisting capacity

The Bridge and Structural Engineer

Literature Review

Brown and Whitlock [1] experimentally studied the behaviour of J-shaped bolts embedded in concrete masonry under direct tension, direct shear, and in combined shear and tension. It was demonstrated that the load-deflection behaviour of the bolts is a complex function of bolt diameter, the masonry strength and the level of pre-tightening. The basic modes of failure were identified, as (1) fracture of bolt, (2) pullout by straightening of anchor bolt and (3) fracture of masonry (wedge failure). The ductility of bolts in direct tension was shown to decrease with bolt diameter, but in direct shear the ductility increased slightly with bolt diameter.

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that some available procedures were significantly unconservative, and all methods show considerable deviation from those tests which are governed by concrete failure. The authors recommended the use of PCI Design Handbook for cases governed by steel failure with a capacity reduction factor 0.9 for design purposes in direct tension. For failure governed by concrete; the authors recommended the use of ACI committee 349-76 with inclusion of a modified capacity reduction factor 0.85 and 0.675 for sandlightweight and all-lightweight concrete in direct tension respectively.

1: Typical Use of Embedded Plate-Anchor

and rotation, were studied. Upper and lower bounds for maximum resisting moment and minimum rotation, respectively, were developed. Shear was shown to have little effect on ultimate moment but to have significant effect on maximum rotation. Behaviour of the moment resistant base plate was studied analytically using FEM techniques by Diluna and Flaherty [3] in order to see the effects of the plate flexibility on the plate assembly. It was found that the distance between tensile reaction and centroid of compressive reaction tends to decrease as the plate becomes more flexible, thereby increasing the bolt tension. De Wolf and Sarislly [4] have experimentally investigated the behaviour of steel column base plates subjected to axial loads and moments. Specimens were tested with anchor bolt size, base plate thickness and ratio of moment to the axial load, as variables. It was concluded that ultimate load design more closely predicts the actual behaviour than working stress method. Klingner and Mendonca [5, 6] reviewed the available procedures for computing nominal capacity of short anchor bolts and welded studs loaded monotonically in direct tension and shear and compared them with available experimental results. It was demonstrated 130â&#x20AC;&#x192; Volume 45

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Thambiratnam and Paramsivam [7] conducted experiments to study the behaviour of base plates under the action of axial loads and moments, by eccentric loading on the column, in the context of the thickness of the base plate and the eccentricity of the load. The results of this study showed that flexible base plates when loaded at high eccentricities failed primarily by yielding of the plate and that their behaviour was somewhat different from that predicted by the existing design methods. In case of thicker base plates, which behave more or less like rigid plate, large bearing pressure can occur due to riding up of the plate on the edge causing premature failure due to crushing of concrete by high bearing pressure at the edge. Cook and Klingner [8] conducted experimental study on multiple anchor connections loaded monotonically, with various combinations of moment and shear. They used flexible and rigid base plates with steel attachments connected to concrete with threaded cast-in-place or retrofit anchors. The authors proposed plastic analysis approach for calculation of capacity of a connection loaded in eccentric shear (i.e. combined shear and bending). Cannon [9] conducted a series of 93 tests on flexible base plate with multiple anchor bolts to study the effect of plate stiffness on the location of compressive reaction, redistribution of load to anchors, and the effect of pre-loading of anchors on performance and capacity of anchorage. It was observed that the amount of redistribution that takes place between the lines of anchors, in connections subjected to shear and moment, depends on the flexibility of the base plate and strain capacity of the anchor steel. A guide was also prepared for the design of anchor bolts and other steel embedments. The Bridge and Structural Engineer


Chakrabarti and Tripathi [10] analysed an idealized model of a typical embedded plate-anchor assembly by the finite element method for different combinations of assembly parameters such as plate thickness, concrete base stiffness and anchor stiffness and presented the effect of variation of assembly parameters on the tensions in plate-anchors, the maximum bearing pressure in the concrete base, and the maximum equivalent stress in the plate. The results obtained from the finite element analysis were compared with those obtained from the “Modified Concrete Beam Analogy Method” and finally a set of design recommendations regarding the use of the modified concrete beam analogy method was proposed. Mishra and Chakrabarti [11] developed a comprehensive analytical model to facilitate a realistic prediction of the static behaviour of embedded plateanchor assembly. In this model the tensionless concrete foundation was incorporated through an iterative scheme. The model was duly verified with results of classical theory, results from previous works as well as with experimental results. The following conclusions are made regarding the linear behaviour of embedded plates: Important design parameters such as maximum anchor tension, maximum base pressure, and maximum plate stress are influenced by plate flexibility, anchor stiffness, base stiffness, and attachment size.

Inclusion of transverse shear effects in a plate demonstrates that the base pressure is

underestimated by about 20% when thin-plate idealization is adopted. The model has the capability for prediction of the capacity (both at failure and collapse) of the assembly, with the load-deformation history, along with indication of the failure modes. Panisetty [12], using a previously developed model, developed interaction curves for computing the failure and collapse loads for plate-anchor assemblies, under different combinations of loads. The effects of anchor spacing, and the ratio of plate size to attachment size were also taken into account. Rodriguez et al. [13] comparatively studied static and dynamic behaviour of single and multiple tensile anchors and observed that the tensile capacities of most anchors tested are at least as high as under quasi-static loading. As a result it was concluded that most anchors, if designed for ductile behaviour under quasi-static loading would behave in a ductile manner under dynamic loading as well. Zhang et al. [14] studied the seismic response of multiple anchor connections to concrete and concluded that multiple anchor connections designed for ductile behaviour in uncracked concrete under seismic loading will probably still behave in a ductile manner in cracked concrete under dynamic loading. Adany and Dunai [15] conducted tests on steel bolted end plate type connections between steel members under cyclic loading and studied the behaviour of components which determine the joint behaviour. It

Fig. 2: Specimen dimensions (in mm) showing placement of various components

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was observed that local buckling of flange forming a plastic hinge results in extremely ductile behaviour with considerable energy absorption. On the other hand, whenever there is significant bolt elongation, the rigidity and energy dissipation capacity of the joint considerably decreases due to rigid body type rotation of the joint.

3.

Experimental Study

The scheme of the experimental set-up and the design of the specimens were finalised such that the actual testing simulates the real life situation of a plateanchor assembly to the best possible extent. In the present study, three specimens were cast, fabricated and tested under eccentric reverse cyclic loading. Specimens were prepared using M40 concrete, mild steel plate and anchor bars. The load was transferred to the plate-anchor assembly through steel attachments. The arrangement represents a steel section attached with a semi rigid concrete base through a plateanchor assembly. The objective of the experimental programme was to study the behaviour of the plateanchor assembly under a specified cyclic loading protocol; yielding of plate, yielding of anchors and the consequent collapse of the assembly were considered in this study by precluding the other possible modes of failure through adequate design of the relevant components of the assembly. 3.1 Specimen Design Details With the objective to achieve the intended behaviour of the specimens, preliminary calculations for prediction of the possible failure modes, and collapse load were made. As the failure of the specimen is more likely to be in the tension phase of loading, the possible failure modes under eccentric tension are considered. The mode of failure is predicted from a 2D model where the moment required for yielding

of the plate at its most critical section was compared with the moment offered by the anchor forces before yielding. The failure predicted by this method was used for the design of plate-anchor assembly. Three specimens were cast with the dimension of the concrete block as 1×0.42×0.5 m, subjected to given loading conditions and supported as shown in the test setup. Three mild steel (as per IS 2062:1992 [19]) plates of thicknesses 12 mm, 16 mm and 20 mm were used. Anchors made of 12 mm mild steel bars (fy = 330MPa as per IS 2062:1992 [19]) in J shape were welded to the steel plate. Bond length provided was such that the bond strength of the anchor bar was more than the tensile strength of anchor according to Equation 3.1 while including the effects of its J shape. (3.1) (4τ bd) Accordingly the effective length of the anchor bar provided is 770.5 mm. The attachment was designed to transfer the load to the plate-anchor assembly without any yielding or distortion of its shape. The concrete base was designed as a reinforced concrete deep beam of length 1 m and the cross section was chosen considering the bending moment and anchorage depth such that all possible modes of failures (including crushing of concrete below plate bottom) are precluded. The grade of concrete adopted is 40MPa. The section is designed as per IS456:2000 [17] limit state design. The reinforcement details are: 4-20 mm dia. HYSD bars at bottom

3-20 mm dia HYSD bars at top

8 mm 2 legged stirrups @ 125 mm c/c

clear cover 40 mm

The overall dimensions of the specimen are shown in Fig. 2.

Fig. 3: Experimental setup for testing of specimens (all dimensions are in mm

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After the casting of concrete members, steel attachments were welded on the top surface of steel plate. The attachments were made of rolled I sections.

The loading protocol is shown in Fig. 4. Such a loading protocol has been chosen due to the fact that it is commonly encountered in industrial applications.

3.2 Experimental setup

The specimens were provided with strain gauges to measure strains on plate surface and anchors and LVDTs to obtain displacements on the plate surface. The instrumentation scheme and the numbering scheme for the sensors is shown in Fig. 5. The testdata acquisition system consisted of strain gauges, LVDTs, load cell of the actuator, scanner and the personal computer.

The experimental setup is shown in Fig. 3. The specimen was placed on the strong floor and supported with the help of reaction frames. The attachment was connected to the actuator head with steel plates and roller arrangement so as to provide a fixed load eccentricity. The specimens were subjected to gradually increasing reverse slow cyclic loading by a servo-hydraulic actuator. A load controlled loading history similar to ATC-24 [16] was adopted for the present study. An increment of 25kN was chosen for the loading cycles. The load was applied at an eccentricity of 200 mm from the center of the plate along the longer side so as to simulate a uni-axial cyclic moment acting on the assembly (Fig. 2). Thus the assembly is subjected to a cyclic axial load and a cyclic uni-axial moment.

In the anchors, strain gauges were fixed in pairs at each section to counter the risk of damage of the strain gauges during placing of concrete and its compaction. In Specimen 1, ten strain gauges were used to measure the strains in the anchors. The strain gauges were fixed at different sections of one end-anchor and one mid-anchor to observe the strain-variation along the length of anchor. The locations were decided based on previous experimental works [11][12]. Twelve strain gauges were used in Specimen 2 and Specimen 3 each. Out of these twelve gauges six were placed on the plate-surface and the remaining six were placed on one row of anchors close to the bottom of plates. All the strain gauges were oriented in the directions along which the maximum strains were expected. The placement of the strain gauges in Specimen 2 and Specimen 3 were decided based on the observed deformation and failure patterns of the Specimen 1.

4. Testing Fig. 4: Loading Protocol

The three specimens were tested according to the loading protocol until collapse of the specimen or up

Fig. 5: Instrumentation

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to the maximum limit of the available capacities of the equipment. The data was recorded from the data acquisition system for strain gauges and LVDTs. The actuator was supported by a reaction block rigidly attached to the strong floor. Physical observations

were made during the test regarding failure modes.

5.

Test Results

The results obtained from testing of the three specimens consist of displacement of plate at selected

Fig. 6: Transverse displacements of specified points on specimen 1

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Fig. 7: Strains in anchor 1 and anchor 2 for specimen 1

Fig. 8: Transverse displacements of specified points on specimen 2

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Fig. 9: Strains at specified points on the plate of specimen 2

points (i.e. the data obtained from LVDTs), strains in anchors and plates as recorded by the strain gauges and the failure/collapse loads. Based on the observations on the average cube strength of concrete of the three specimens, the strength of concrete was found to fulfil the target mean strength criteria for M40 grade of concrete as per IS456:2000 [17]. Based on the results of the tensile tests conducted on plate and anchor materials as per IS 1608:1995 [18], the yield strength for both plate and anchors satisfied the criteria of mild steel as per IS 2062:1992 [19]; but, the ultimate strength of the anchor material was found to be on the higher side of the prescribed limit. The average stresses at yield of the anchor-material and plate material were found to be 365.105 MPa and 299.75 MPa, respectively. The average values of

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Youngâ&#x20AC;&#x2122;s Modulus for the anchor material and plate material were observed to be 2.11x105 MPa and 2.18x105 MPa, respectively. The sign convention adopted for load is: compression is positive and tension is negative. For the displacement of different points on the plate, outward displacement is considered positive. Tensile strains are positive and compressive strains are negative. Representative test data on plate-displacements (as obtained from LVDTs) and on strains (in anchors and in plates, as recorded by the strain-gauges) are presented selectively in Fig. 6-11. The LVDT data is used for further calculating the rotation of the plate and in determining the deflected shape of plate. Excessive local rotational displacement (as identified by local kink-type plate bending) of plate representing yielding The Bridge and Structural Engineer


can also be determined from plate displacement data (Fig. 6, 8 and 10 for specimen 1,2 and 3 respectively). Fig. 7 shows the representative strain in anchor 1 and 2 of specimen 1. The loading cycle which leads to yielding of anchors can be identified and also the ultimate load carried by anchors can be seen from this anchor strain data. Anchor strain data from specimen 2 and 3 was not of much use due to damage and excessive noise. For specimen 1 plate yielding was determined from LVDT data aided by physical visual observation. However, for specimen 2 and 3 strain data from strain gauges attached to plate was also used to determine yielding of plate (Fig. 9, 11). These together help in determining the mode of failure as either plate bending or anchor yielding or a combination of both.â&#x20AC;&#x192;

6. Derived Test-Data and observed failure pattern The area of the plate under the attachment was

considered as a rigid part and the rotation and the displacement of the assembly were taken as the average rotation and the displacement of the rigid part. LVDT readings on the plate were used to obtain the rotation and displacements of the rigid part of plate as shown in Fig 12. The moments were calculated from the values of loads and the sign convention adopted for the load was followed for the moments. The fixed moment on the attachment due to the self weight of the actuator was included in the applied moment readings and instantaneous eccentricity calculated from rotation of assembly was used in calculation of moments. The sign convention adopted for load is: compression is positive and tension is negative. The moment applied (along-with associated rotation) by the compressive load is taken as positive and that applied by the tensile load is taken as negative. For the displacement of different points on the plate, outward displacement is considered positive. The load-displacement curves thus obtained for the three

Fig. 10: Transverse displacemnts of specified points on specimen 3

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Fig. 11: Strains at specified points on the plate of specimen 3

specimens are shown in Fig. 13-15 and the moment rotation curves are shown in Fig. 16-18.

Fig. 12: Rotation and Displacement of the assembly

In the first specimen, cracking and chipping of concrete from around the bottom edge of plate started in the first cycle of 100kN at about -80kN load along with uplift of plate from the concrete base. There was visible bending of the plate near the bottom edge of attachment in the compression phase of first cycle 138â&#x20AC;&#x192; Volume 45

Number 4 December 2015

of 150kN. The bending and uplift of plate is shown in Fig. 20, 21 and 22. In the next tension phase, the deformation and rotation of the assembly increased rapidly with the increase of load. The experiment was then halted to prevent damage to measuring equipment. In the second specimen, cracking and chipping of concrete from around the bottom edge of plate started in the first cycle of 125kN (at about -120kN load) along with uplift of plate from the concrete base. The plate was rotating like a rigid body as shown in Fig. 23. In the tension phase of first cycle of 150kN the deformation increased further. In the next compression phase the attachment started moving out of plane. The experiment was then halted. In the third specimen the first visible deformation of the plate was observed at a load of -120kN in the The Bridge and Structural Engineer


first cycle of 125kN near the bottom flange of the attachment. The plate was bulging out near the flange of the attachment. In the tension phase of first cycle of 150kN the bulging out was clearly visible with some deformations near the loads decreased. This was due to breaking of an anchor. The bulging of the plate is shown in Fig. 24. In the next compression cycle the specimen moved out of plane. The test was then halted. The assembly rotational stiffness degradation curves are obtained from moment-rotation curves. The curves for the three specimens are shown in Fig.19 .

Fig. 15: Load -Displacement curve for specimen 3

Fig. 16: Moment -Rotation curve for specimen 1 Fig. 13: Load -Displacement curve for specimen 1

Fig. 17: Moment -Rotation curve for specimen 2 Fig. 14: Load - Displacement curve for specimen 2

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near the bottom flange of attachment at -86.023kN in first cycle of 100kN load. The top anchors yielded in the first cycle of 125kN at -107.8kN. The failure of specimen was in combined mode of failure of plate and anchors. The failure pattern for specimen 1 is shown in Fig. 19-21.

Fig. 18: Moment -Rotation curve for specimen 3

7.

Discussion of test results

The area of the plate under the attachment is considered as a rigid part in the plate. The rotation and displacement of the assembly are taken as the average displacement and rotation of the rigid part of the plate. It is observed that for specimen 1, uplift of plate from concrete base occurred in the first cycle of 100kN (at about -80kN). In the compression phase of 150kN, there was visible bending of the plate near the bottom edge of the attachment and also between the bottom row anchors. In specimen 2, cracking and chipping of concrete from around the bottom edge of plate started in the first cycle of 125kN (at about 120kN) along with uplift of plate. The third specimen showed visible deformation of plate at a load of 120kN in the first cycle of 125kN near the bottom flange of attachment. The plate was bulging out near the flange of the attachment. In the tension phase of first cycle of 150kN the bulging was clearly visible with some deformations. This was due to the breaking of anchor.

In specimen 2, 20 mm thick plate was used and anchor diameter was kept same as specimen 1. In this case the anchor yielding in the bottom row started at 101.2kN load of first cycle of 125kN. The top row anchors started yielding at 149.6kN in the first cycle of 150kN. The extreme anchors failed earlier than the middle anchors and plate, as the plate was much thicker. The test was stopped in the 150kN second cycle when the attachment with the plate moved out of plane as a rigid body. Most of post yield behaviour could not be studied in this specimen. The failure pattern for specimen 2 is shown in Fig. 22. In the third specimen, 12mm thick plate was used keeping anchor size constant. As expected the plate yielded at a lower load. The plate started yielding at 74.8kN load near the top flange of the attachment in the first cycle of 75kN. The plate yielded near the bottom flange of the attachment at -56.6kN load in the first cycle of 75kN load. The bottom row and middle row anchors yielded simultaneously at -109.4kN in the first cycle of 125kN. The failure was a pure plate failure. The structure was not identical in resisting tension and compression. The failure pattern for specimen 3 is shown in Fig. 23.

7.1 Modes of failure In general there are three primary modes of failure for plate-anchor assembly: (1) Plate yielding, (2) Anchor yielding and, (3) crushing failure of concrete. In specimen 1, 16 mm thick plate and 12 mm diameter anchors were used. Yielding occurred first in the anchor and in subsequent cycles the plate yielded. First the bottom row anchor reached yield stress at -83.7kN load in the first cycle of 100kN. The middle row anchor reached yield at 97.76kN load in the first cycle of 100kN load. The plate reached yield stress 140â&#x20AC;&#x192; Volume 45

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Fig. 19: Failure pattern for specimen 1, view from bottom

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Fig. 22: Plate uplift in specimen 2

Fig. 20: Plate failure in specimen 1 form side

Fig. 23: Plate bulging between anchor rows in specimen 3

7.2 Ductility Ratios of Assemblies Fig. 21: Failure pattern for specimen 1, view from top

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Ductility ratio is one of the important parameters in establishing the cyclic characteristics. It is the ratio of

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the rotation at which failure starts to that at collapse. It was observed that in case of combined mode of failure, though the rotational values are more, the ductility ratio is highest in case of plate failure. The ductility ratios for different specimens are presented in Table 1. Table 1: Ductility ratios for different specimens Specimen

Rotation at failure start (rad.x105)

1 (t=16mm) -0.42 2 (t=20mm) -0.54 3 (t=12mm) -0.28 * Till the test continued

Final Ductility rotation at Ratio collapse (rad.x105) -7.49 17.83 -3.24* 6.00 -5.05 18.04 t= Plate thickness in the specimen

7.3 Stiffness Degradation of the Assemblies The assembly has two kinds of stiffness : translational stiffness and rotational stiffness. In cyclic loadings the rotational stiffness is more significant than the translational one. The stiffness degradation is an important parameter in the cyclic tests which is caused due to the progressive change in the stiffness of the structure. The rotational stiffness obtained in tension phase are only considered for inspecting the stiffness degradation. The rotational stiffness in different cycles for the three specimens is shown in Fig. 24.

increments the stiffness gradually decreases. The stiffness further decreases because of failure of plate and again after cycle 10 due to failure of an anchor. Similar decrease in stiffness can be seen for specimen 1 initially and when there is an anchor failure after cycle 10. For specimen 2 there is an initial decrease similar to specimen 3 possible due to loss of bond and after cycle 13 a drastic drop when failure of anchor takes place. 7.4 Comparison of results The failure loads have been compared with those obtained from interaction curves [12] for monotonic compressive and for monotonic tensile loading at eccentricity of 200mm. The values from the experiment and interaction chart are presented in Table 2. It is seen that the predicted failure load is lower than experimental failure load only for tension phase whereas the compressive failure load is lower than the predicted value. This can be explained by the fact that the neutral axis shifts towards the centre and the plate loses complete contact with base after the plastic deformation of anchors. This induces more force in the anchors causing them to fail earlier in compression. Table 2: Comparison of results Specimen

Failure Loads (kN) at 200mm Eccentricity Experimental

From Interaction Chart

Tension Compression Tension Compression 1

83.7*

77.74

57.6

144

2

101.2*

149.6

75.2

192

3

56.6

72.6*

44.8

96

* indicates the failure that occurred first

8. Conclusions

Fig. 24: Rotational Stiffness Degradation curves for specimen 1,2 and 3

The stiffness of specimen 3 decreases suddenly after the 25kN cycle, possible due to failure of bond between plate and concrete. In subsequent load 142â&#x20AC;&#x192; Volume 45

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The present study gives an overview of the behaviour of plate-anchor assembly when subjected to uni-axial cyclic moment. The results from this study can be used in combination with other parametric studies to develop design guidelines for such structures. Following conclusions are made based on the study : 1.

The relative stiffness of the plate with respect to the anchors plays a dominant role in governing the mode of failure of the assembly. The Bridge and Structural Engineer


2. The rotational stiffness of the assembly drops abruptly after the anchor failure, unlike the case of plate failure. 3. Residual deformation in anchors in tension phase can reduce the failure load in compression phase by a substantial amount. 4. Combined failure mode of the plate-anchor assembly gives the maximum rotation and transverse displacement in post yielding cycles. However, the ductility ratio in case of platefailure mode, is marginally higher (by about 2%) than that for the combined failure. 5. It was observed that, in the plastic range the third cycle repeats over the second cycle and does not give additional information regarding the structural behaviour, relatively to the first two cycles. This is important in the context of adoption of three cycle loading (reverse cyclic) as a common accepted practice. The present study focuses on few parameters which play a key role in governing the behaviour of plateanchor assemblies subjected to cyclic load. However more studies are needed to develop a comprehensive set of design guidelines so that such structures can be designed efficiently. The parameters which should be considered for further studies include the different load combinations encountered in industrial applications, variations in anchor diameter, different shapes of attachments and how these affect the behaviour of the assembly.

9. References 1.

BROWN R.H., WHITLOCK E.G., “Strength of Anchor Bolts in Grouted Concrete Masonry”, Journal of Structural Engineering, ASCE, Vol.109, No. 6, 1983, pp.1362-1374.

2. SALMON, CHARLES G., SCHENKER L., JOHNSTON B.G., “MomentRotation Characteristics of Column Anchorage”, Transaction, ASCE, Vol.122, 1957, pp.132-154 3.

DILUNA L.J., FLAHERTY J.A., “An Assessment on the effect of Plate Flexibility on the Design of Moment Resisting Base Plates”, Proceedings of the Pressure Vessels and Piping

The Bridge and Structural Engineer

Conference, San Francisco, California (USA), pp 25-29 4.

DEWOLF J.T., SARISLLY E.F., “Column Base plates with axial loads and moments”, Journal of Structural Mechanics Division, ASCE, Vol 106, No. ST11, 1980

5.

KLINGNER R.E., MENDONCA J.A., “Shear Capacity of Short Anchor Bolts and Welded Struts: A Literature Review”, ACI Journal Proceedings, Vol. 79, No. 5, pp.339-349, 1982

6.

KLINGNER R.E., MENDONCA J.A., “Tensile Capacity of Short Anchor Bolts and Welded Struts: A Literature Review”, ACI Journal Proceedings, Vol 79, No. 4, pp.270-279, 1982

7. THAMBIRATNAM D.P., PARAMSIVAM P., “Base Plates Under Axial Loads and Moments”, Journal of Structural Engineering, ASCE, Vol.112, No. 5, pp 1166-1180, 1986 8.

COOK R.A., KLINGNER R.E., “Ductile Multiple Anchor Steel To Concrete Connections”, Journal of Structural Engineering, ASCE, Vol.118, No.6, 1992, pp.1645-1655

9.

CANNON, R.W., “Flexible Base Plates: Effect of Plate Flexibility and Preload on Anchor Loading and Capacity.”, ACI Structural Journal, Vol. 89, No.3, pp 315-324, 1992

10. CHAKRABARTI S.K., TRIPATHI R.P., “Design of Embedded Steel Plates in Reinforced Concrete Structures”, Structural Engineering Review, Vol. 4, No. 1, pp 81-91, 1992 11. MISHRA R.C., CHAKRABARTI S.K., “Comprehensive Model for Embedded Plates”, Journal of Structural Engineering, ASCE, Vol.126, No.5, pp. 560-571, 2000 12. PANISETTY R., “Non-Linear Modeling of Embedded Plate: Prediction of Capacity and Behavior”, M.Tech Thesis Report, Department of Civil Engineering, IIT Kanpur, 1997 13. RODRIGUEZ M., LOTZE D., GROSS J.H., ZHANG Y.G., KLINGNER R.E., and GRAVES H.L., “Dynamic Behaviour of Tensile Anchors to Concrete”, ACI Structural Journal, Vol. 98, No. 4, pp 511-524, 2001 14. ZHANG

Y.G.,

KLINGNER

R.E.,

and

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GRAVES H.L., “Seismic Response of MultipleAnchor Connections to Concrete”, ACI Structural Journal, Vol. 98, No. 6, pp 811-822, 2001 15. ADANY S., DUNAI L., “Modeling of Steel-to-Concrete End-Plate Connections under Monotonic and Cyclic Loading”, Periodica Polytechnica Civil Engineering, Vol. 41, No. 1, pp 3-16, 1997 16. ATC-24, “Guidelines for Cyclic Seismic Testing of Components of Steel Structures”, Applied

Technology Council, Redwood City, CA, 1992 17. IS 456:2000, “Plain and Reinforced Concrete – Code of Practice”, Bureau of Indian Standards, Bahadur Shah Zafar Marg, New Delhi. 18. IS 1608:1995, “Mechanical Testing of Metals – Tensile Testing”, Bureau of Indian S tandards, Bahadur Shah Zafar Marg, New Delhi. 19. IS 2062:1999, “Steel for General Structural Purposes – Specification”, Bureau of Indian Standards, Bahadur Shah Zafar Marg, New D

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INDIAN NATIONAL GROUP OF THE IABSE OFFICE BEARERS AND MANAGING COMMITTEE – 2015 Chairman 1. Shri DO Tawade, Chief Engineer (Coordinator-II), Ministry of Road Transport and Highways Vice-Chairmen 2. Shri Divakar Garg, Director General, Central Public Works Department

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Members of the Executive Committee

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Shri Ninan Koshi, Former DG (RD) & Additional Secretary

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MEMBERS OF THE MANAGING COMMITTEE – 2015 Rule-9 (a): A representative of the Union Ministry of Road Transport and Highways 1.

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36. Shri AK Banerjee , Former Member (Technical), NHAI

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Technical Presentation on “Maintenance and Rehabilitation of Bridge Structures” held at New Delhi in November 1994 Seminar Report on “Elevated Transport Corridors” held at Maysore (Karnataka) on 27th and 28th June 2014 Themes of the Seminar are as under:Technical Session - I Planning Preparation, Economic benefits & Value Engineering Technical Session - II Design and Construction Technical Session - III Operation and Maintenance Technical Session - IV Case Studies Hard Copy Out of Stock – CD is available 36th IABSE Symposium on “Long Span Bridges and Roofs – Development, Design and Implementation” held at Kolkata from 24th to 27th September 2013 Vol.43 No.3, September 2013 Special Issue – Urban Flyovers (Structure, Architecture, Sustainability) Hard Copy Out of Stock – CD is available Vol.43 No.4, December 2013 Special Issue – Bearings, Expansion Joints & STUs for Bridges (Selection, Design, Testing Installation, Maintenance) Hard Copy Out of Stock – CD is available Vol.44 No.1, March 2014 Special Issue – Building Structures Vol.44 No.2, June 2014 Special Issue – Codes & Standards in Structural Engineering (Developments & Need for Improvement) Vol.44 No.3, September 2014 Special Issue – Reinforced Soil Walls (Current Practices & Future Directions) Vol.44 No.4, December 2014 Special Issue – Structural Failures (and Lessons Learnt) Vol.45 No.1, March 2015 Special Issue – Earthquake Resistant Design of Structures Vol.45 No.2, June 2015 Special Issue – Strengthening, Repair & Rehabilitation of Structures Hard Copy Out of Stock – CD is available Vol.45 No.3, September 2015 Special Issue – Aesthetics of Structures Annual Subscription Charges for Quarterly Journal - 2016 “The Bridge and Structural Engineer” Published in March, June, September & December

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148  Volume 45

Number 4 December 2015

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The Bridge and Structural Engineer

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150â&#x20AC;&#x192; Volume 45

Number 4 December 2015

The Bridge and Structural Engineer


The Bridge & Structural Engineer  

Vol. 45 No. 4, December 2015

The Bridge & Structural Engineer  

Vol. 45 No. 4, December 2015