STRUCTURE magazine - October 2018

A Joint Publication of NCSEA | CASE | SEI

STRUCTURE

®

October 2018 Bridges Inside: Pierceville Bridge, PA

2018 NCSEA Summit Chicago, IL • Oct 24-27

LeMessurier Calls on Tekla Structural Designer for Complex Projects Interoperability and Time Saving Tools

Tekla Structural Designer was developed specifically to maximize collaboration with other project parties, including technicians, fabricators and architects. Its unique functionality enables engineers to integrate the physical design model seamlessly with Tekla Structures or Autodesk Revit, and to round-trip without compromising vital design data.

“We’re able to import geometry from Revit, design in Tekla Structural Designer and export that information for import back into Revit. If an architect makes geometry updates or changes a slab edge, we’ll send those changes back into Tekla Structural Designer, rerun the analysis and design, and push updated design information back into Revit.”

Tekla Structural Design at Work: The Hub on Causeway

For over 55 years, LeMessurier has provided structural engineering services to architects, owners, contractors, developers and artists. Led by the example of legendary structural engineer and founder William LeMessurier, LeMessurier provides the expertise for some of the world’s most elegant and sophisticated designs while remaining true to the enduring laws of science and engineering.

Known for pushing the envelope of the latest technologies and even inventing new ones, LeMessurier engineers solutions responsive to their clients’ visions and reflective of their experience.

An early adopter of technology to improve their designs and workflow, LeMessurier put its own talent to work in the eighties to develop a software solution that did not exist commercially at the time. Their early application adopted the concept of Building Information Modeling (BIM) long before it emerged decades later.

While LeMessurier’s proprietary tool had evolved over three decades into a powerhouse of capability, the decision to evaluate commercial structural design tools was predicated on the looming effort required to modernize its software to leverage emerging platforms, support normalized data structure integration and keep up with code changes.

After a lengthy and thorough comparison of commercial tools that would “fill the shoes” and stack up to the company’s proprietary tool, LeMessurier chose Tekla Structural Designer for its rich capabilities that addressed all of their workflow needs. According to Derek Barnes, Associate at LeMessurier, ” Tekla Structural Designer offered the most features and the best integration of all the products we tested. They also offered us the ability to work closely with their development group to ensure we were getting the most out of the software.”

One Model for Structural Analysis & Design

From Schematic Design through Construction Documents, Tekla Structural Designer allows LeMessurier engineers to work from one single model for structural analysis and design, improving efficiency, workflow, and ultimately saving time. “Our engineers are working more efficiently because they don’t need to switch between multiple software packages for concrete and steel design. Tekla Structural Designer offers better integration of multiple materials than we have seen in any other product,” said Barnes.

LeMessurier engineers use Tekla Structural Designer to create physical, information-rich models that contain the intelligence they need to automate the design of significant portions of their structures and efficiently manage project changes.

TRANSFORMING THE WAY THE WORLD WORKS

“Tekla Structural Designer has streamlined our design process,” said Craig Blanchet, P.E., Vice President of LeMessurier. “Because some of our engineers are no longer doubling as software developers, it allows us to focus their talents on leveraging the features of the software to our advantage. Had we not chosen to adopt Tekla Structural Designer, we would have needed to bring on new staff to update and maintain our in-house software. So Tekla Structural Designer is not just saving us time on projects, it is also saving us overhead.

Efficient, Accurate Loading and Analysis

Tekla Structural Designer automatically generates an underlying and highly sophisticated analytical model from the physical model, allowing LeMessurier engineers to focus more on design than on analytical model management. Regardless of a model’s size or complexity, Tekla Structural Designer’s analytical engine accurately computes forces and displacements for use in design and the assessment of building performance.

“Tekla Structural Designer offers better integration of multiple materials than we have seen in any other product.”

Positioning a large scale mixed-use development next to an active arena, a below grade parking garage, and an interstate highway, and bridging it over two active subway tunnels makes planning, phasing and engineering paramount. Currently under construction, The Hub on Causeway Project will be the final piece in the puzzle that is the site of the original Boston Garden.

Despite being new to the software, LeMessurier decided to use Tekla Structural Designer for significant portions of the project. “Relying on a new program for such a big project was obviously a risk for us, but with the potential for time savings and other efficiencies, we jumped right in with Tekla Structural Designer. It forced us to get familiar the software very quickly.”

“Tekla Structural Designer allowed us to design the bulk of Phase 1 in a single model,” said Barnes. The project incorporates both concrete flat slabs and composite concrete and steel floor framing. “Tekla Structural Designer has the ability to calculate effective widths based on the physical model which is a big time saver,” said Barnes. “On this project, the integration with Revit, along with the composite steel design features enabled us to work more efficiently. Adding the ability to do concrete design in the same model was a bonus because we had both construction types in the same building.”

“Tekla Structural Designer helped this project run more efficiently, and in the end it was a positive experience,” said Blanchet.

“Tekla Structural Designer gives us multiple analysis sets to pull from, which gives us lots of control. Most programs don’t have the capability to do FE and grillage chase-down. For the design of beam supported concrete slabs, Tekla Structural Designer allows us to separate the slab stiffness from the beam stiffness, so if we choose to we can design the beams without considering the influence of the slab. In the same model we can use a separate analysis set to review the floor system with the beams and slab engaged,” said Barnes.

Barnes also shared similar benefits with concrete column design. “Tekla Structural Designer does grillage take-downs floor-by-floor, finds the reactions and applies them to the next floor. This allows us to view column results both for the 3-dimensional effects of the structure as a whole and from the more traditional floor-by-floor load take-down point of view.

Doing both has always required significant manual intervention, but Tekla Structural Designer puts it all in one place.” “We reduce the possibility for human error because with Tekla Structural Designer less user input is required,” said Barnes. “Tekla Structural Designer automatically computes many of the design parameters, such as column unbraced lengths. The assumptions made by the software are typically correct, but we can easily review and override them when necessary.”

“Tekla Structural Designer provided the best fit for our workflow compared to other commercially available software.”

Want to Evaluate Tekla Structural Designer? tekla.com/TryTekla

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EDITORIAL BOARD Chair Barry K. Arnold, P.E., S.E., SECB ARW Engineers, Ogden, UT chair@STRUCTUREmag.org Jeremy L. Achter, S.E., LEED AP ARW Engineers, Ogden, UT Erin Conaway, P.E. AISC, Littleton, CO

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John A. Dal Pino, S.E. FTF Engineering, Inc., San Francisco, CA Linda M. Kaplan, P.E. TRC, Pittsburgh, PA Timothy M. Gilbert, P.E., S.E., SECB TimkenSteel, Canton, OH Jessica Mandrick, P.E., S.E., LEED AP Gilsanz Murray Steficek, LLP, New York, NY

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Evans Mountzouris, P.E. The DiSalvo Engineering Group, Ridgefield, CT Greg Schindler, P.E., S.E. Sammamish, WA Stephen P. Schneider, Ph.D., P.E., S.E. BergerABAM, Vancouver, WA John “Buddy” Showalter, P.E. American Wood Council, Leesburg, VA

STRUCTURE® magazine (ISSN 1536 4283) is published monthly by The National Council of Structural Engineers Associations (a nonprofit Association), 645 N. Michigan Ave, Suite 540, Chicago, IL 60611 312.649.4600. Application to Mail at Periodicals Postage Prices is Pending at Chicago, IL and additional mailing offices. STRUCTURE magazine, Volume 25, Number 9, C 2018 by The National Council of Structural Engineers Associations, all rights reserved. Subscription services, back issues and subscription information tel: 312-649-4600, or write to STRUCTURE magazine Circulation, 645 N. Michigan Avenue, Suite 540, Chicago, IL 60611. The publication is distributed to members of The National Council of Structural Engineers Associations through a resolution to its bylaws, and to members of CASE and SEI paid by each organization as nominal price subscription for its members as a benefit of their membership. Yearly Subscription in USA $75; $40 For Students; Canada $90; $60 for Canadian Students; Foreign $135, $90 for foreign students. Editorial Office: Send editorial mail to: STRUCTURE magazine, Attn: Editorial, 645 N. Michigan Avenue, Suite 540, Chicago, IL 60611. POSTMASTER: Send Address changes to STRUCTURE magazine, 645 N. Michigan Avenue, Suite 540, Chicago, IL 60611.

STRUCTURE is a registered trademark of the National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

CONTENTS

30 RIVETS AND RELOCATION By The Historic Bridge Foundation

Cover Feature

The 113-foot Pierceville Bridge, ca. 1881, is a rare lenticular truss bridge which served as a traffic crossing over the Tunkhannock Creek. In 2005, deteriorated conditions prompted its closure. Engineers restored the structure while keeping to its origins, devising an ingenious scheme for transporting the historic bridge to Lazy Brook Park. It will serve as a pedestrian crossing in its new home.

Columns and Departments

26 Shaft Wall Solutions for Wood-Frame Buildings – Part 2

By Richard McLain, P.E., S.E.

EDITORIAL

By Tom Huempfner

7 38-Minute Missile Threat Provides an Emotional Lesson

By Kevin Nakamoto, P.E.,

12 Springfield’s Great Bridge Salutes History

By Beth McGinnis-Cavanaugh

STRUCTURAL REPAIR

16 Metal Plate Connected Wood Roof Trusses

on a side-by-side comparison of two different fill materials – a

By John P. Busel

lightweight aggregate and Geofoam. After five years in-place, the results are interesting.

BUSINESS PRACTICES

44 Strengthening Your Employer Brand

TO THE EDITOR

37 THE RITZ-CARLTON RESIDENCES WAIKIKI BEACH, PHASE 1

By James O. Malley, S.E., P.E., SECB,

By Steven Baldridge, P.E., S.E., Fernando Frontera, S.E.,

Lawrence F. Kruth, P.E., and

and Anantha Chittur, P.E., S.E.

Michael D. Engelhardt

The Ritz-Carlton Residences Waikiki Beach Phase 1 project

By Jennifer Anderson

and Hussam Mahmoud

HISTORIC STRUCTURES

was settling due to deteriorating embankments. TxDOT decided

41 Preserving History

STRUCTURAL ECONOMICS

By Jennifer McConnell

The header for the U.S. 67 bridge over SH 174 (Cleburne, TX)

STRUCTURAL REHABILITATION

and Corey Matsuoka, P.E.

8 Life Cycle Assessment of Steel Bridges

Features 34 MATERIAL DIFFERENCE

STRUCTURAL COMPONENTS

46

in Honolulu, Hawaii, faced several major challenges but successfully overcame them through the creative use of post-

SPOTLIGHT

51 Portland’s Sellwood Bridge

By Eric Rau, P.E., and

By John Stewart, P.E.

David Goodyear, P.E., S.E., P.Eng.

CONSTRUCTION ISSUES

STRUCTURAL FORUM

21 Night Moves on Boylston

By Nathan C. Roy, P.E., and Ethan A. Rhile, P.E.

58 (When) Will Robots Replace Us?

By Eytan Solomon P.E.

tensioning, resulting in a stunning structure that reflects the Ritz-Carlton’s luxurious brand.

IN EVERY ISSUE 4 Advertiser Index 48 Resource Guide – Wind/Seismic 52 NCSEA News 54 SEI Update 56 CASE in Point

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, the Publisher, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

STRUCTURE magazine

5

October 2018

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Editorial 38-Minute Missile Threat Provides an Emotional Lesson

By Kevin Nakamoto, P.E., and Corey Matsuoka, P.E.

I

t was a clear, sunny Hawaiian morning on Saturday, January 13th, and • Do you really know what is important to them? Is it the schedule, I had just finished warming up my daughter’s soccer team, which was cost control, quality, responsiveness, or something else? preparing to take the field when a message flashed on my cell phone: • What do they worry about going wrong on the project? Is it the “BALLISTIC MISSILE THREAT INBOUND TO HAWAII. SEEK loss of funding, exposure to public criticism, exposure to lawsuits, IMMEDIATE SHELTER. THIS IS NOT A DRILL.” or something else?

Naturally, I did what any quick-thinking engineer would do. I Once you understand the answers to those questions, you can tailor moved the team to a shelter inside the park’s hollow tile restroom. It your project management style to meet the client’s needs and provide would be 38 minutes before state officials broadcast another message strategies to mitigate those concerns. declaring: “THERE IS NO MISSILE THREAT OR DANGER TO Of paramount importance is being able to communicate this THE STATE OF HAWAII. REPEAT. FALSE ALARM.” understanding to your client, which means identifying an effective Too late. This was the scariest 38 minutes of my life – not to men- communication plan. At a minimum, the plan should contain how, tion the lives of 11 frightened 10-year-old girls. what, and when you will communicate with your client. Larger and Meanwhile, Honolulu resident Noah Tom had just dropped off more complicated projects will need more frequent communication to his oldest daughter at the ensure a smooth delivery. airport and was picking This is especially true if up breakfast for a meetyou seek to manage your ing when he heard the client’s expectations. alert. His two younger While a client expecting children were at home, perfection might not be and his wife was already realistic, urgent responses at work. “I literally sent and constant communicaout ‘I love you’ texts to tion in response to errors as many family members are logical expectations.

It is the project manager’s responsibility to as I could. It was kind of It is the project manager’s surreal at that point,” he responsibility to ensure ensure the client’s expectations are in sync told The Washington Post. the client’s expectations When he heard the alert in sync with reality. with reality. Regular and effective communication are was a mistake, Tom had Regular and effective not yet made it home. communication is a great is a great tool to make this happen.

Instead, he pulled over tool to make this happen. to the side of the road Finally, inquire of your and cried. “I just broke down at that point. It all kind of hit me in a clients about how they treat their own clients. This should give you wave, what I had just gone through. I was unable to drive for 20 or a feel of how they, in turn, appreciate being treated.

30 minutes,” he said.

Not long ago, I asked a successful architectural client who recently I now know that scientists estimate a ballistic missile originating from retired about his firm’s approach to client service. His response: “You North Korea would take approximately 20 minutes to reach the state need to be in front of the client, showing them that you are absorbed of Hawaii – only 20 minutes until life as you know it changes forever. in the success of their project. Show enthusiasm and urgency.”

So, it is not hard to imagine the anger and uproar from my fellow In other words, to maximize your opportunity for positive project Aloha State citizens criticizing state officials for taking 38 minutes to outcomes, never leave a client confused at any stage about what is send out a message retracting the original alert. clearly going on.▪ The emergency management agency did not understand the expectation of the public and the consequences of the time it took to rescind the alert. In the public’s eye, it did not act with enough urgency to This article was originally published in the July/August 2018 issue of Engineering Inc. and is reprinted with the permission of the inform them of the mistake. In the era of social media, smartphones, American Council of Engineering Companies (ACEC). and instant gratification, an immediate update was expected.

The emergency management agency misunderstood the needs of the people it serves or its clients – whether it was terrified little girls Kevin Nakamoto (knakamoto@ssfm.com) is a Senior Structural Engineer, or a father crying on the side of the road – who needed to know an and Corey Matsuoka (cmatsuoka@ssfm.com) is the Executive Vice President important fact sooner rather than later. at SSFM International, Inc. in Honolulu, Hawaii. Corey is the chair of the The same theory works for engineering. You need to know your cliCASE Executive Committee. ent’s expectations on a project and answer these important questions:

“

”

STRUCTURE magazine

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October 2018

structural

L

ife-cycle assessment (LCA) is a quantitative means for assessing the environmental impact of an object. In structural engineering, the object of interest may be a building, bridge, or other structure. In order to provide a complete picture, the environmental impacts from the entire life-cycle of the object is considered: from the acquisition of the raw materials needed to form the members, through the energy and ancillary materials involved in the processing and transportation of these materials and members, exca- Figure 2. Birds-eye view of a steel twin-tub steel girder bridge. vation required during construction, future maintenance such as redecking or painting, not unrelated. Contrary to a common misconcepup to the end use or disposal of the members. tion, the most sustainable option is also often the Once the inventory of all of these items is com- most cost-efficient option when considering the full pleted, the associated environmental impacts life cycle. For example, the use of fly ash in place on climate, air quality, water quality, human of ordinary portland cement results in significant health, and resource reductions in CO2 and hence global warming potendepletion can be tial as well as cost savings and improved durability characterized using in typical applications. standardized methThe life-cycle of a steel bridge can be considods. For example, ered in four phases: design (resulting in choices the global warming of materials and their quantities), construction, potential of a proj- in-service (where inspection is required and ect can be expressed in terms of the equivalent maintenance is typically needed), and end-ofBy Jennifer McConnell mass of CO2. Such an analysis can be completed life (where the materials must be disposed of and Hussam Mahmoud using various methods and software to reveal the or repurposed). The most substantial environpotential for minimizing environmental impacts mental impacts come from the acquisition and Jennifer McConnell is an Associate on a project, for comparing alternative design fabrication of the materials based on the design Professor at the University of Delaware concepts, and for obtaining credits in sustain- of the structure: information is available in other in Newark, DE. (righman@udel.edu) ability guidelines such as Envision and LEED. publications to assist designers with minimizing Hussam Mahmoud is an Associate On the other hand, life-cycle cost analysis (LCCA) this impact. The environmental impacts from Professor at Colorado State University is a well-known concept for evaluating the eco- the construction and in-service phases can also in Fort Collins, CO. nomics of a structure, typically used to evaluate be minimized based on thoughtful actions from (hussam.mahmoud@colostate.edu) whether a particular investment or initial cost has structural engineers.

The authors are members of the a long-term economic advantage for reasons such ASCE/SEI Sustainability and Steel as reduced maintenance or longer useful life. While LCA Data from Bridge Committees.

LCCA and LCA originated and are typically applied Construction Alternatives with very different goals in mind – determining the most cost-effective option versus determining What is considered in an LCA of steel bridge the most sustainable option – the two concepts are construction? An LCA of the construction of a structure accounts for the environmental impacts associated with its erection.

Maintenance End of Life Construction Material Figure 1 lists the impacts directly associated Selection with construction for a typical project: site preparation (e.g., excavation or dewater•Acquistion and •Energy •Site •Raw material ing); energy (e.g., diesel fuel) consumed fabrication of consumed by preparation acquistion by construction equipment; transportation repair materials demolition •Energy •Transportation of materials, personnel, and equipment •Traffic equipment consuption of raw material disrucption •Site disruption •Tranportation •Material to the construction site; and the ancillary •Transportation •Transportation of materials, production material (e.g., formwork) used in construcof materials, of materials, personnel, and •Transportation tion. However, the construction method personnel and personnel, and equipment of produced often affects the design of the structure. For equipment for equipment •Ancillary material maintenance •Landfilling example, whether or not scaffolding will be material •Additional and inspection •Traffic fabrication used, the lifting sequences of curved steel disruption girders, and options of balanced cantilevering or launching of segmental concrete Figure 1. Stages of life and corresponding environmental impacts associated with structures.

ECONOMICS

Life Cycle Assessment of Steel Bridges During Construction and in Service

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October 2018

after considering the design modifications needed to resist the unique stresses resulting from the launching process.

Optimizing In-Service Performance How can LCCA impact the scheduling of inspection and repair of bridges to extend the life-span of bridges and save taxpayer’s money?

In the United States, 11% of the bridges are classified as fracture critical, 83% of which are two girder steel bridges. Steel twin-tub girder bridges, similar to that shown in Figure 2,

STRUCTURE magazine

are a common example of a two-girder steel bridge. The cost implications of the bi-annual inspection mandate are enormous and impose financial strain. This problem is further aggravated by the fact that most bridges in the United States were built around the 1960s and many of them have shown significant signs of aging and deterioration. Demolition and replacement of these bridges is an expensive alternative. The development of these cost-effective strategies requires the formulation of LCCA models, which need to include both epistemic and aleatory uncertainties associated with the specific bridge in question. Several researchers have

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bridges affect the material quantities used in the structural design. Furthermore, options for accelerated bridge construction (ABC) techniques, such as prefabricated versus cast-in-place deck elements and slide-in construction, also affect environmental impacts related to both construction and design. Thus, the LCA is not exclusive to the environmental impacts occurring during construction, but also includes the environmental impacts from the material quantities used in the design of the structure.

How can sustainability impacts from construction variables be assessed? With a cognizant perspective of the basic reducereuse-recycle concept, many qualitative inferences can be made. For example, the essential principle of ABC is to reduce construction time, from which it logically follows that diesel fuel consumed by construction vehicles and traffic impacts (and the associated natural resource depletion and air pollution) would also be reduced.

However, in some cases, this could be offset by additional site clearing or preparation.

Other logical ways to reduce the environmental impact of construction choices include reusing formwork and diverting construction waste from landfills into functional purposes.

How can sustainability impacts from construction variables be quantitatively assessed? LCA provides the framework for a quantitative assessment to more rigorously evaluate the relative impacts of the different components of the construction process and compare construction (or other) alternatives; recent studies have carried out such work. One such example is the work of Dequidt (2012), who quantified the global warming potential (GWP) associated with bridge construction for a sample project.

This work showed that approximately 80% of the GWP from the construction phase was associated with transportation to the site, 20% was associated with diesel fuel consumed by construction equipment, and the remaining factors were negligible. In a case study comparing five different construction alternatives, diesel fuel consumption was again found to be a significant variable.

In this case, varying by as much as 65%, depending on how much work was done onsite versus prefabricated. Furthermore, the results of three different LCA studies can be compared to provide information regarding conventional construction versus launching. This comparison shows that for bridges with similar lengths (1000 ft.), launching reduced the GWP by 25% and the acidification potential by 64%, even

9

October 2018

Figure 3. a) Convergence of the stress field around the crack; b) Finite element mesh of the bridge; c) Zoom in on the cracked region.

developed comprehensive probabilistic LCCA frameworks for optimal maintenance budget allocation regarding deteriorating structures.

These studies have quite effectively highlighted the importance and need for efficient LCCA strategies to establish well-balanced intervention schedules that consider various economic and safety requirements while taking into account uncertainties associated with the time-dependent structural performance. Implementation of these strategies is likely to result in a more extended period between inspections for most bridges. More quantitative-based inspection and maintenance intervals may also reduce the environmental impacts of these actions, for example by reducing the emissions associated with inspections and reducing the need for future materials for more extensive rehabilitations or bridge replacements.

What is considered in the LCCA of bridges?

Assessment of life-cycle cost requires sufficient understanding of all factors involved in maintaining and prolonging the life of the structure. This may include, but is not limited to, inspection, repair, and maintenance.

Depending on the nature of the method adopted for each activity, the life-cycle cost could vary significantly. For example, the inspection cost is heavily dependent on the method used (e.g., ultrasonic versus visual inspection). Since the life-cycle cost calculated

is an expected value, the inspection cost for a specific inspection type will have to be multiplied by the probability of detection, which could have a significant impact on the overall cost. There exist several types of inspection methods, each with a distinct accuracy.

Depending on the crack size and inspection type, sometimes critical cracks would not be detected during the inspection phase. In addition to the expected direct cost associated with the inspection method, indirect cost should be included in the LCCA. For inspection, the indirect cost will mainly depend on the time taken by the specific inspection method and the closure cost per day, which is independent of the method type. In case of repair costs, both direct and indirect components should be considered and are typically related to the crack length, which in turn vary with time probabilistically. The direct repair cost is proportional to the crack length and could be defined as the cost per unit length of the crack. The indirect cost, on the other hand, could be defined as the product of closure cost per day, crack length, and the amount of time taken to repair a crack of a specific length.

This general formulation has been established and applied to studies on various structures.

What is expected from LCCA results? The intervals at which these inspections and maintenance activities should be specified require careful deliberation to minimize lifecycle cost of structures while ensuring structural safety. Therefore, optimization is required where the life-cycle cost information, along with the probabilistic curves of fatigue crack growth, can be used to calculate optimal inspection/ repair routines. The probabilistic curves of fatigue crack growth can be developed using simplified analytical fatigue growth models or using comprehensive finite element models similar to that shown in Figure 3. The advantage of using comprehensive finite element modeling is the minimization of uncerFigure 4. Schedule pattern for inspection and repair of the tainties associated with modeling the crack at the location of interest for minimum life cycle cost.

STRUCTURE magazine

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October 2018

structure or the crack. The optimization is performed to obtain an optimal inspectionrepair schedule, such that the service life can be extended to a desired target life (e.g., 70 yrs). Figure 4 shows an example of a schedule pattern for inspection and repair.

Conclusion Both LCA and LCCA represent optimization problems. In the former, the goal is to minimize environmental impact; in the latter, the goal is to minimize cost. The above discussion reveals that minimizing certain metrics can have a significant influence on both of these goals. Based on the data presently available, the most significant of these appears to be time on site. From an environmental perspective, reducing the time on site translates into reduced diesel fuel emissions from construction equipment, reduced emissions from personal transportation of workers, reduced emissions from traffic delays, and the associated reduction in fossil fuel consumption. For the same reasons, reduced time on site can also translate into reduced cost for both public agencies and the traveling public. Other strategies for minimizing the environmental impact of steel bridge construction, supported by LCA data, include the selection of local materials and launching longer span bridges. During the service life of a bridge, most of the available work focuses on LCCA, which highlights that understanding problematic locations in bridges is key to minimizing direct costs associated with inspection and indirect costs associated with bridge closures. This includes, for example, identification of fatigue-prone details before implementation of the inspection program. Furthermore, one maintenance strategy supported by LCA is the design of selective retrofits to improve natural hazard resistance.â–Ş The online version of this article contains references. Please visit www.STRUCTUREmag.org.

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historic

STRUCTURES

“H

ere...is the gateway to Springfield and the towns to the east for almost an entire nation,” proclaimed Massachusetts Governor Channing Cox on August 2, 1922. It was Dedication Day for the new Hampden County Memorial Bridge, which spans the Connecticut River between the City of Springfield and Town of West Springfield in Western Massachusetts.

Boston engineers Fay, Spofford & Thorndike, with architects Haven & Hoyt, designed the bridge, deemed a “finely-engineered example of a rare self-supporting arch rib reinforcement technique derived from the Melan tradition” [HAER Ma-114]. Builder H. P. Converse & Co. of Boston completed the bridge ahead of schedule on July 31, 1922, after 28 months of construction. At 1,515 feet long and 80 feet wide, it was designed to support pedestrian, vehicular, street rail, and heavy armament traffic. A bridge big in size and cost ($4 million) for its time, it remains the longest reinforced concrete deck arch span in Massachusetts.

Springfield’s Great Bridge Salutes History

By Beth McGinnis-Cavanaugh, MSCE Beth McGinnis-Cavanaugh is a Professor at Springfield Technical Community College in Springfield, Massachusetts, where she teaches courses in physics, engineering mechanics, and structures.

She is particularly interested in the engineering and social significance of historic structures. (bmcginnis-cavanaugh@stcc.edu)

The Memorial Bridge sits at the nexus of three rivers: the 410-mile north-south Connecticut River, New England’s longest, and major east and west tributaries Chicopee and Westfield Rivers.

With river access to New York and Canada and the rich soil of the Connecticut River Valley, Springfield was founded as a trading and farming community when Puritan William Pynchon purchased land from the Agawam Indians in 1636. The river was not bridged until 1805 when a wooden toll bridge was built. A covered wooden toll bridge followed in 1816, but the motorized cars and trucks of the 20th century and a burgeoning Springfield population made a new bridge imperative in the early 1900s.

Hampden County Memorial Bridge looking west from Springfield, August 1922. Springfield viaduct in the foreground. 1816 covered wooden toll bridge upstream, in the process of deconstruction. A segment of railroad bridge is visible beyond. Courtesy of the Lyman and Merrie Wood Museum of Springfield History.

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Looking east, showing piers and arches. Retrieved from the Library of Congress.

Years of dispute forced the Commonwealth to appoint an independent commission to finalize its design and location. The Connecticut’s soft riverbed precluded solid concrete and masonry structures, and exposed steel arches were deemed unpleasing for what principal Charles M. Spofford called “this important new artery of commerce spanning a great New England river.” In 1919, commissioners selected a reinforced concrete deck arch bridge named to honor “those who had died as pioneers, and soldiers in the Revolutionary, Civil and Foreign Wars.” The bridge, designed in the Beaux-Arts style, boasted seven parabolic concreted rib arches on six piers and two abutments that spanned 1,200 feet across the river. A nine-span viaduct of 314 feet over railroad tracks on the Springfield (east) side formed the Springfield approach. The bridge was located 400 feet downstream of the 1816 covered bridge at right angles to the river – just north of the river’s widest point.

The Memorial Bridge opened to great fanfare in 1922, “beautiful in the sweep of its lines, the last word in engineering science...a symbol of that progressiveness that has been characteristic of the valley” [Springfield Republican, July 1922].

Springfield had shed its colonial past, surpassing neighboring Hartford in size and status and emerging as a hub of industry, innovation, and intellect. General George Washington had established the Springfield Armory in 1777, where the first American musket and famous Springfield rifle were produced. After the War of 1812, the Armory pioneered the use of interchangeable parts and assembly line production, making Springfield the nation’s epicenter of precision manufacturing – the “Silicon Valley of the 19th century.” This catalyzed industry of all types.

Springfield was the birthplace of the Duryea car, America’s first gas-powered automobile,

and the Indian motorcycle. The “City of Firsts” was home to Knox fire engines, Wason railroad cars, Goodyear vulcanized rubber, Rolls Royce automobiles, Smith and Wesson firearms, and Merriam-Webster’s dictionary, among many other firsts. Naismith’s game of basketball, Milton Bradley board games, four Carnegie libraries, renowned museums, and a young Dr. Seuss also called Springfield home in 1922.

Thus, upon its opening, the Memorial Bridge was more than it appeared – much more than just a river crossing. It was an announcement.

The frontispiece of a confident city, the bridge exuded strength, permanence, and promise.

It was the very embodiment of Springfield in 1922. Like the City, it was “practically imperishable” according to H. P. Converse, who stated, “I can’t think of anything that will prevent the bridge from standing as firmly 500 or a thousand years from now as it does today” [Springfield Republican, July 1922].

The Melan System

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The introduction of the Melan system spurred the construction of reinforced concrete bridges in the U.S. in the late 1800s and early 1900s.

The system called for parallel, self-supporting steel arch ribs – curved I-beams – encased in concrete to support traditionally reinforced superstructures. Ribs were placed along the centerline of the arches, not where tensile stresses occurred as would be typical with traditional reinforced concrete. Steel and concrete were used in parallel to support loads but did not act as a composite material. By 1924, over 5,000 Melan or Melan-style bridges had been built in the U.S. Patented by Austrian engineer Josef Melan (1854-1941) in 1893 in the U.S., the Melan system was originally a design for suspended floors and roofs in warehouses and other large span buildings.

Melan, a renowned bridge engineer and professor of structural mechanics, adapted it for bridge use after testing showed that it was 3 to 4 times stronger than other bridge designs, including those using Monier’s wire mesh. The system was championed in the U.S. by former Melan student Fritz von Emperger, who patented several variations, including the use of lighter latticed (trussed) ribs instead of I-beams.

Longer and wider spans, greater loads, ease and speed of construction, and economy made the Melan system popular in the U.S. The use of steel arch ribs minimized the use of concrete or masonry as in barrel arches, which reduced dead load.

Concreting of the ribs added strength and stiffness. This permitted longer spans with

fewer piers, and greater live loads.

Further, the use of ribs to support formwork and concrete induced stresses in the ribs that allowed the more efficient use of the steel and maximized the steel’s strength. With traditional reinforcement, the capacity of the steel was limited by the concrete modulus, which was typically 10 to 15 times less than that of steel.

The Melan system promoted ease and speed of construction, which meant Bridge construction. Concrete hoisting tower (130 feet) shown.

Concrete was transported from mixing plant on West Springfield fewer laborers, less skilled labor, and side along a temporary wood trestle 70 feet upstream. Courtesy less time. Unlike concrete or masonry of the Lyman and Merrie Wood Museum of Springfield History. arches that could not be prefabricated or labor-intensive bar reinforcement, arch ribs were delivered in two or four sections ready for erection. The ribs were stable during erection with minimal support and equipment and designed to support the formwork for concrete, which was hung on the ribs. This eliminated vast amounts of falsework, which minimized the use of timber and simplified construction over terrain that threatened the stability of the falsework. Further, the system allowed multiple facets of construction to be done simultaneously; for example, ribs in one span could be concreted while ribs in another span Arch erection in span 7, Springfield side. Courtesy of the were erected.

Lyman and Merrie Wood Museum of Springfield History.

As spans lengthened and live loads increased due to the growth of vehicular I-beams were replaced with lighter latticed and street rail traffic, larger rib sections were or trussed hinged ribs. Often, these ribs required. To reduce increasing amounts were reinforced with hoops and traditional of steel and dead load as well as pier size, bars. The use of hinges minimized bending

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October 2018

stresses, temperature stresses, and stresses due to shrinkage of concrete, and made for more straightforward analysis. Crown hinges were fixed during concreting, reducing deflections of the arches.

The Melan system was an alternative to traditional reinforcement and the prevailing uncertainty about composite action, concrete quality, and construction methods. Swiss bridge engineer Robert Maillart criticized the system because the design could not rely on the bond between concrete and ribs. Maillart believed the lack of bonding would lead to separation and ultimate corrosion of steel.

Moreover, as Melan bridges were overbuilt to inspire confidence, the system was surely an affront to the efficiency and elegance of Maillart’s three-hinged deck-stiffened concrete arches. Also problematic was the concrete encasement of the steel, which prevented proper drainage and led to corrosion.

The Bridge “No other bridge in the country is just like it…” boasted the bridge souvenir edition of the Springfield Republican in July 1922, adding “the structure will long be of interest to engineers.” The bridge was “structurally and architecturally significant” and the engineering “sophisticated” [HAER MA-114].

A total of 10,500 pine piles, 20 to 40 feet in height and spaced 20 feet on center on hard clay, form the foundation for six river piers and two abutments. Under the channel span piers, there are 2,263 piles; under the smallest pier on the West Springfield (west) shore, there are 700. An average of 110 piles per day were placed with two steam-powered

pile drivers. Concrete piles were used for the viaduct spans.

The hollow concrete piers support 5 arches per span. The piers, constructed using cofferdams, vary in size. The channel span piers, designed to accommodate a potential draw span per order of the Army Corps of Engineers, are the largest at 65 feet by 179 feet. Each pier has ten skewbacks – two for each arch – and does not extend above the springing. All are faced with 10 courses of cut granite from just below the water level that protects the piers from river current and winter ice flows. Dredging was done to maintain the natural flow of the river.

The arch span lengths vary from 110 to 209 feet; the span rises from 19.1 to 29.7 feet.

Marked by four 80-foot beacon towers, the channel span is 176 feet in width and 40 feet above low water over 60 feet, “fixed in accordance with the requirements of the War Department” [ENR 88, 13] for all “navigation necessities” [ENR 88, 13]. The bridge is asymmetric on the river to follow deep water; that is, the channel span is the third span from the Springfield side. The two spans that flank the channel span on either side, 154 feet and 146 feet in length, are sized to give the bridge symmetry. Smaller beacon towers embellish these spans, which furthers the illusion of symmetry.

The remaining two smaller spans on the West Springfield side balance the Springfield viaduct.

The nine viaduct spans are equal in width.

In all arch spans, there are five parallel arches: two exterior and three interior arches, which are centered under the critical street rail load in the middle of the road deck. Each arch is a steel arched Warren truss rib encased in concrete. All arches are 5.5 feet wide but vary

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October 2018

in height with span length, ranging from 4 feet 9 inches to 7 feet. The arches support a traditionally reinforced concrete deck on reinforced columns. The inner arches are not filled, and exterior spandrel walls hide the arches and columns to give the bridge the appearance of solid masonry.

All of the 35 steel arch ribs, each weighing between 20 and 70 tons, were initially three-hinged, transported in four sections.

They were erected in only 10 days with an erection sequence designed to ensure the stability of the piers. Falsework was used to support the arches at the crowns and quarter points on the four larger spans during erection. The ribs were encased in 593 psi concrete, as compared to a working stress of 16,000 psi for the rib steel. The crown hinges were fixed after concreting, leaving the ribs as two-hinged. On the outer arches, the crown hinges were fixed before concreting; the interior crown hinges were fixed after the roadway deck was placed to offset deformations due to dead load and shrinkage. Within each span, the ribs are connected with wind bracing and reinforced with traditional bars along the arch and hoops around the rib.

Conclusion The Melan system fell out of favor in part as steel became more expensive and less available. More so, a better understanding of cement and concrete technologies, composite behavior, and the development of uniform codes and construction methods moved structures towards more efficient and economical bar reinforcement. The system experienced a rebirth in the 1970s which continues in Japan and China, where self-supporting arches are used to construct bridge spans in mountainous regions.

As a structure, then, the Memorial Bridge was technically obsolete almost upon its execution – a harbinger of things to come for Springfield. Alterations to the bridge were made in 1950 and 1966, and it was obfuscated by the opening of an elevated viaduct in 1970. A complete rebuilding of the superstructure was completed in 1996 after extensive corrosion was discovered. An EF3 tornado rendered it unscathed in 2011, yet the recent construction of a downtown casino has cast further shadow. But the bridge will not be denied. Transcendent, it remains a vital channel and landmark in Western Massachusetts. As it nears its 100th year, it continues a touchstone for a City struggling to reclaim itself and persists in bearing witness to Springfield’s illustrious past.■

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structural

REPAIR

T

he use of Metal Plate Connected (MPC) wood roof trusses is a common structural framing system used in residential and light commercial structures. Low initial cost of fabrication, ease of installation, and ability to accommodate varying profiles are some of the advantages of this type of roof framing system. However, when these trusses are damaged by fire, evaluation and repair of the trusses can be challenging. This article presents methods for evaluation and repair of fire-damaged MPC wood roof trusses and discusses the factors that go into the decision whether to remove and replace or repair-in-place, as illustrated by several real-life case studies.

Design Repair Phase

Metal Plate Connected Wood Roof Trusses Evaluation and Repair of Fire Damaged

By John Stewart, P.E. John Stewart is the Structural Practice Area Leader at Rimkus Building Consultants. (jstewart@rimkusbc.com)

It is important to keep the client’s overall cost and schedule in mind, as they may take precedence over a sophisticated repair design that minimizes the number of repair materials used. The remediation contractor usually likes simple repetitive details and more repetition, which typically lowers the unit costs.

Primary considerations in the repair versus replacement decision include: • From a global perspective, if a high percentage of the trusses are damaged, and only a few trusses can be salvaged, it is better just to replace all the trusses. Matching up the alignment of new and pre-existing trusses that may have deflected or sagged over time can be problematic and difficult. Where possible, group replacement trusses by configuration, slope, function, or location. Extend the truss replacement area to an expansion joint or a

convenient architectural breakpoint such as a roof valley or ridge.

• Within a truss, determine if there are enough undamaged members that can remain in place to allow the repair of the damaged members.

Generally, if more than 25% of the members in a truss need repair, then it might be better to replace the entire truss.

• The repair versus replace decision must also take into account non-structural factors such as mechanical and electrical utilities that run through the trusses. Total removal and replacement of damaged trusses may be more costly because of the need also to remove and replace/ reinstall the utility lines. This idea also applies to expensive roof coverings and ceiling finishes.

More extensive in-place repairs and temporary shoring of the damaged trusses may be justified by these other costs.

• Consider how the remediation contractor will access tight working spaces, such as attics, for both workers and materials. Temporary openings can be cut in end walls or holes cut in ceilings and the roof to bring in materials. If new replacement trusses are needed, a small crane will be required to place the trusses.

Access to set the crane within a reasonable reach of the delivery truck and the placement location can be a critical factor.

• Consider the costs of cleaning smoke residue and sealing odors on the trusses that will remain in place. Soda blasting, dry ice, and sanding are typical cleaning methods with a final application of a sealant used to provide odor protection.

Many times the answers to the above items are easy to make, and the repair versus replacement decision is an obvious one. However, for more complex situations, the decision-making process can be complicated because it involves issues beyond structural calculations.

Concerning the actual repair of the MPC trusses, there is often a tendency to go with full truss replacement because it is felt that the repair of the trusses is inherently a costly item and requires an extensive amount of technical expertise. Some contractors claim that local building officials mandate that the only person that can modify an MPC truss is the original designer of the truss.

This is not accurate. As professional engineers, we can modify existing structural elements provided we have the required technical expertise, and we take responsibility for the entire truss that we are repairing.

Truss Repair Techniques A structural analysis of the truss must be performed to understand the distribution of the forces, moments, and stresses. Generally, it is sufficient to assume all members are pinned and that

Figure 1. Charred bottom chord truss members.

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October 2018

Figure 2. Repair detail for bolted sister members.

secondary moments are neglected if panel point spacings are reasonable. Generally, conservative sizing of repair members and connections is not a significant cost driver because the majority of the repair cost and schedule may be associated with accessing the repair area and getting materials to the repair location. An additional 15% to 20% increase in the repair design factor of safety is reasonable, given the unknowns in the conditions of the members that remain, the difficult conditions that the contractor has to work in, and construction tolerance and fit-up issues.

The most common and most straightforward repair is to use sisters to existing members that have sustained minimal damage. This usually involves localized charring of a member that is not located near a panel point or a joint.

The sisters will generally be the same size as the main member, with sisters on both sides to maintain symmetry. If access to only one side of the damaged member is possible, the eccentricity of a one-sided connection is generally not a problem for members with low forces. Be sure there is enough good remaining material to fasten the sisters to and provide a long enough lap splice length beyond the

damaged area to develop the forces required.

The cost difference between using a 6-foot long or an 8-foot long 2x4 is negligible, so it is important to ensure you have the appropriate length of the sisters.

Many designers use 10d or 12d nails in single shear because framers typically have nail guns that use these nails. If spacing limitations are present, Simpson metal straps, steel side plates, or through-bolts can be used. Through-bolts in double shear have higher allowable shear forces that can be up to 10 times greater than nails.

The second most common repair is the replacement of a damaged member between

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9/13/2018 2:08:24 PM

Figure 3. Damaged truss members and the HVAC plenum.

two-panel points utilizing plywood or OSB gusset plates for connection at the panel points. It is preferred to use gusset plates on both sides of the truss to keep the forces symmetrical. The connection forces to size the gusset plate fasteners will be determined from the structural truss analysis. Once again, 10d or 12d nails in single shear are used through the gusset plate on each side of the truss.

As an alternative detail, one can use 14d or 16d nails in double shear by driving the nail completely through both gusset plates and clinching the end of the nail on the back side. This will result in fewer nail holes in the truss members. However, for both methods, holes should be pre-drilled in the members if splitting becomes a problem. The spacing of the nails at 3 to 4 inches is usually preferred to help avoid splitting of wood.

The size of the gusset plates will generally be large enough such that shear or tension stresses in the gusset plates are not a problem. The size of the gusset plates will also be

much larger than the existing MPCs. MPCs are a proprietary product with high allowable stresses based on extensive testing. The factor of safety used for the design of the plates is also reduced compared to the typical factors of safety applied to wood members connected with nails or bolts. Therefore, do not be surprised at the large number of nails required and the larger size of the gusset plates for members with high loads.

The final type of repair to be discussed is a full depth gusset plate. Rather than trying to repair individual panel point connections and members, plywood or OSB side plates are installed from the top chord to the bottom chord creating a box beam. This situation can occur for trusses with minimal depth where member forces are high, and there is not enough room to develop the full force of the member in the room allotted for the gusset plate. This can also occur where the truss members are installed flatwise. The internal web members could be replaced, but there

Figure 4. Repair detail using bolts for select connections.

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October 2018

will not be enough room for the required number of nails to develop the full force of the web members. Therefore, the web members are omitted, and the beam is designed with the side plates carrying the entire shear forces and the top and bottom chords forming a tension-compression couple to carry the bending moment. The connection between the side plates and the chords is designed using classical shear flow theory which generally results in a reasonable nailing pattern.

Normally, the side plates are extended one to two panel points beyond the damaged/ repaired area to provide a “transition zone� between the original truss section with discrete web members/forces and the box beam section with shear stresses in the side plates.

Normally, a two-panel point transition zone is used near the ends of a truss where the shear stresses are high. Near the middle of a truss, a one-panel point transition zone beyond both ends of the damaged/repair area is needed.

If there are concerns about high member forces in the webs and the transition of these forces to shearing stresses in the side plates, a finite element analysis can be performed to evaluate this. In general, when this type of analysis is performed, the transition zone from discrete web and chord member forces to distributed shearing stresses in the side plates occurs within the first half of the first adjacent panel point because the stiffness of the side plates is greater than the discrete web members. Another method of confirming the adequacy of the repair is to load test the truss.

Case Studies Sister Repair Detail There was minimal damage to the bottom chord of three trusses due to a fire from below.

The Howe trusses spanned 40 feet and were composed of 2x4 members. There were numerous ducts, water pipes, and electrical conduits that ran through the openings in the trusses. There was sprayed-on foam insulation on the underside of the roof sheathing that was not damaged.

Because the damage to the three bottom chords only included minor charring of the 2x4 bottom chord member at two locations and the nearby metal plate connectors were not damaged, isolated repair of the damaged chord members was selected (Figure 1, page 16 ). This eliminated the need for the full replacement of the trusses, finishes, insulation, and utility lines.

The force in the damaged bottom chord was 3,240 pounds tension. It was conservatively decided to sister the bottom chord with an eight-foot-long 2x4 on both sides to provide an additional allowance for fit-up.

Because of the relatively high member force, the use of 5⁄8-inch diameter through-bolts was selected for the fasteners. Three bolts at each end of the sisters were sufficient, but a fourth bolt was added near the panel point to provide additional capacity and stability (Figure 2, page 17 ).

Nailed and Bolted Gusset Plates Portions of the top chord, bottom chord, and a web member were partially consumed by a fire from below (Figure 3). The metal plate connectors at the ends of the damaged members were warped from the heat of the fire. The Howe trusses spanned 45 feet and were composed of 2x4 web and bottom chord members with 2x6 top chord members. The damaged trusses sandwiched a plenum above an HVAC unit below. Large horizontal ducts came out of the plenum and were routed between the web members of the truss. The underside of the roof sheathing above was covered with smoke residue but was not structurally damaged. Isolated replacement of the damaged truss members was deemed to be

more cost-effective than total removal and replacement.

The force in the damaged top and bottom chord members was 4,500 pounds. The maximum force in the damaged web members was 1,300 pounds.

Plywood/OSB gusset plates were used to connect the new members at the panel points.

Because of the relatively high member forces in the top and bottom chords and interferences with the ducts, 5⁄8-inch diameter through-bolts were used to minimize the size of the gusset plates (Figure 4).

Figure 5. Damaged top chords of the hybrid truss.

Box Beam Design Repair Portions of the top chord, bottom chord, and web members at the ends of four 60-footlong hybrid Warren trusses were partially consumed by fire (Figure 5). The trusses were a parallel wood chord truss 4 feet deep with 1-inch diameter hollow steel pipes for the web members. The top and bottom chords were 2x4 stress rated wood members laid flatwise. The web members were connected to the top and bottom chords with a single ½-inch diameter through-bolt at each panel point. Approximately 8 linear feet of the four subject trusses were damaged at one end.

Total replacement of the trusses would have required removing and replacing numerous ducts, pipes, and electrical conduits that were threaded through the trusses. Also, there was a long lead time to fabricate the trusses, and a small crane would have been required to set the new trusses. Even though the construction of the trusses was unique and the member forces were very high, it was decided to temporarily shore the trusses and perform isolated repairs.

The vertical reaction at the ends of the trusses was 2,100 pounds. The force in the first diagonal web member was 2,200

Figure 6. Repair detail at the end of the hybrid truss.

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pounds. Trying to pass this force through nailed gusset plates was not workable. Using through-bolts in the flatwise chords similar to the existing connections was also not workable. Hence, it was decided to make a box beam at the end of the truss using full height plywood/OSB side plates. The box beam concept extended the 8-foot damaged length of the truss and then another 8 feet to create a two-panel point transition zone (Figure 6 ). A finite element model was run to investigate the stresses. Because the plywood/OSB side plates are stiffer than the discrete web members and chords, there was a localized area of high shear stress in the side plates that only extended about 8 inches from the ends of side plates. This was deemed acceptable because minor slippage of the nails would redistribute the high localized shear stresses over a large area.

Conclusion There are various methods for evaluation and repair of fire-damaged MPC wood roof trusses and many factors that go into the decision whether to remove and replace or repair-inplace. Consider the repair solutions presented as an alternative to the replacement of firedamaged trusses.▪

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T

he recent $78 million transformation of the Boston Public Library by William Rawn Associates of Boston, Massachusetts, opens up the Library’s Johnson Building to become a more inviting public space. It improves the connectivity between the Johnson Building, 1972, and the Library’s original McKim Building, 1895. The structural scope to accomplish the transformation included the removal of an existing floor to create a grand two-story lobby along Boylston Street and removal and replacement of a portion of a 7-story column for the installation of a new glass elevator. Also included was the support and removal of concrete walls and removing a segment of a heavily loaded masonry bearing wall to create a long-missing link between the Johnson and McKim Buildings. The renovation occurred while the building remained occupied and open to the public. The article Handle with Care published in STRUCTURE (September 2018), summarized the main challenges and structural solutions to accommodate the ambitious renovation.

The primary structure for the transformation, design by LeMessurier as the Engineer of Record, considered the “means and methods” of how the structure might be altered. Drawings presented proposed construction schemes including shoring, bracing, jacking, and sequencing. The contractor, Consigli Construction, was ultimately responsible for the temporary conditions during construction.

Consigli Construction engaged Becker Structural Engineers to facilitate the details of the construction means and methods. Engineered shoring, erection, and jacking sequences required early collaboration and communication among the design and construction teams.

McKim Wall Opening Connecting the Johnson Building to the original McKim building required a 36-foot opening in the McKim Building’s century-old brick masonry bearing wall. The existing wall consisted of a 30-inch-thick unreinforced masonry brick wall supporting a load of over one million pounds.

The wall supported terracotta arch masonry floors spanning between iron beams. Foundations for the wall consisted of unreinforced granite walls and pile caps supported by timber piles, typical for the Boston Back Bay at the time.

Rigid steel framing consisting of double W18x158 beams above and below the opening, supported at the ends and approximate beam third-points with W12 columns, were used to support the wall above. The 3-span continuous beams were designed to limit live load deflection to 1⁄16 of an inch. Stiff framing was provided so as not compromise the masonry wall and brittle terracotta masonry floor arches, as well as to adequately distribute the load to the original granite foundation wall and timber piles below.

The structural drawings showed recommended sequencing along with conceptual shoring for the general scheme used in the design. The existing wall was cored above and below the opening to allow for W10 needle beams to be inserted and fully grouted into the wall. The temporary shoring consisted of W10 spreader beams engaging the needle beams, each side of the wall spanning between proprietary shoring posts. To insert the new W18 beams at the base of the wall, a portion of the wall needed to be removed which compromised the support of the existing first floor McKim beams. The W10

ISSUES

Night Moves on Boylston

McKim wall removal construction sequencing.

STRUCTURE magazine

construction

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October 2018

Means and Methods for Renovation of the Boston Public Library Johnson Building

By Nathan C. Roy, P.E., and Ethan A. Rhile, P.E. Nathan C. Roy is an Associate at LeMessurier and teaches at the Boston Architectural College. Nathan was the project manager for LeMessurier, the Structural Engineer of Record for the BPL transformation. (nroy@lemessurier.com) Ethan A. Rhile is a Senior Associate with Becker Structural Engineers, Inc. Ethan was the project manager acting as the Specialty Shoring Structural Engineer for the BPL transformation. (ethan@beckerstructural.com)

McKim wall removal temporary shoring. Temporary needle beams and shoring at McKim wall (left); Temporary needle beams, jacks, and jack boxes (upper right); 14 jacks used to simultaneously preload McKim Shoring (lower right).

spreader beam provided temporary support for these beams until they could be reattached to the structure.

Complete drawings for the existing McKim structure were not available. Exploratory work performed between concept and final shoring design resulted in changes, including coordination of needle beam and post locations to avoid existing framing above and below the opening. A lintel beam above an existing double door opening was discovered at an elevation higher than anticipated. The lintel conflicted with the temporary needle beams.

Pre-shoring of the header within the existing opening was provided to support the wall and allow for coring of the lintel and installation of needle beams.

Both the final structural steel framing and temporary shoring were preloaded with jacks to limit deflections of the existing McKim Building unreinforced masonry walls and terracotta masonry floor arches. Fourteen independently controlled hydraulic jacks were used during preloading of the shoring.

Custom jack boxes at the top of the posts were designed to accommodate hydraulic jacks to allow for preloading of the framing. The tops of shoring posts had to be temporarily braced to each other and to the wall to provide stability of columns during preloading of the wall.

Load requirements were coordinated between the design and construction teams.

Loads were independently calculated, with dead loads of approximately 25.5 kip/ft and unreduced live loads of 6.5 kips/ft. It was agreed to preload the frame to 75% of calculated dead loads. Fourteen independently controlled hydraulic jacks were used

during preloading of the shoring while nine inflatable air jacks were used to preload the final framing.

One of the most critical components of the construction sequence was monitoring during preloading. Representatives from LeMessurier, Becker, and Consigli were present for all jacking operations. Preloading occurred at night when floors above were closed to the public.

Deflection monitoring points at the top, along the length of the opening, were set up to continuously monitor for movement. No measurable movement was observed during jacking operations, demolition, and after temporary shoring removal.

Johnson Wall Removal Between the new Boylston Hall and the new opening to the McKim building was an existing stair servicing all floors. Two 10-inchthick by 8-foot-long reinforced concrete walls on either side of the stair supported both the stair and surrounding floor structure. The walls not only supported vertical loads but also acted as cantilevers extending from the mat slab foundation. The transformation required the relocation of the stair and the removal of the walls to provide an open and clear connection between spaces. Temporary lateral wall bracing was provided between walls to allow for removal of the mezzanine slab prior to removal of the wall. Also, a 4-foot 2-inch opening was made in each of the walls above the second floor to accommodate the relocated stair.

A steel framing scheme was developed to allow for erection, jacking, and demolition without the need for temporary shoring.

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Cantilever steel framing was provided to support the wall above the second floor.

Two W24x250 beams, one on either side of each wall, cantilever 10 feet with a 10-foot backspan and are both supported by a shared system of W36 girders and W14 columns.

One column was required to be discontinuous and supported by a transfer girder at the first floor to accommodate the architectural program. The original 7-foot-thick reinforced concrete mat slab provided flexibility in locating new columns as required to meet the architectural layout without the need for new foundations. The wall sections to remain on either side of the new opening above the second floor were reinforced with concrete sections on either side of the wall, creating two reinforced concrete columns per wall. W12 needle beams centered below the newly created columns (i.e., two W12s per wall) were designed to transfer the load between the walls and steel framing. The framing was designed to limit live load deflections to L/800.

Walls were estimated to each support a load of over 450 kips, with unreduced live load representing approximately 15 to 20 percent of the total. The steel framing was preloaded for the estimated dead load with hydraulic flat jacks to limit deflection, reduce the required steel tonnage, and protect the existing Johnson Building. Jacks were located between each end of the needle beams and W24 cantilevers, a total of four jacks per wall.

The initial Contractor-proposed scheme was to preload a pair of W24 beams at one wall, remove the jacks and then preload the second pair of W24 beams at the other wall. The design team demonstrated that over 25% of the force would be lost in the first W24 pair when the second pair was loaded. Therefore, it was agreed that both pairs of W24 beams would be loaded simultaneously to eliminate any preload losses.

Deflections were monitored during preloading to ensure that calculated wall weight was accurate and the building was not raised.

Preloading resulted in approximately 1Âź inches of deflection in the steel framing with less than 1â „8-inch of deflection in the framing after jack removal.

Johnson Column Replacement One architectural feature of the renovated space included a new glass-enclosed elevator adjacent to and in front of the McKim opening that replaced a double elevator bank in the existing Johnson Building. The new elevator was located directly next to an existing concrete column that was braced

Johnson Building wall removal. Partial plan of second floor framing at Johnson wall removal and McKim Link. Green – removed walls below. Red – cantilever and backspan W24. Blue – support framing. Orange – needle beams. McKim wall let; (left) Wall to W24 detail (right).

by the existing mezzanine. The mezzanine was removed as part of the project to create Boylston Hall. The existing concrete column was replaced with a new, slender, built-up steel tube column, ten inches square with 1½-inchthick walls to minimize the visual impacts of the column in proximity to the glass elevator.

The column replacement occurred at about the third point of the total column height; two levels of sub-basement located below the main level and the column extended upward another 6 stories above where the column was replaced. The major challenge in replacing the column was how to shore the loads above the work area, remove the existing column and then install the new column.

The original construction concept was to shore around the column for the full height of the building. However, the spaces above the column were occupied and some of them recently renovated. Further, mechanical and electrical support spaces located in the sub-basement below the column limited the opportunities to shore. It quickly became clear that the vertical extent of shoring would need to be limited to reduce the impact on the other spaces outside the scope of the renovation.

The shoring scheme utilized to replace the existing column was to construct a steel collar around the existing column above the second floor. The collar was constructed of steel channels and plates, with adhesive anchors utilized to transfer the column loads to the collar.

Just above the second floor, two steel girders were utilized as jacking points. Jacking of the column above was considered critical to assure that the load would be removed from the existing column and transmitted to the shoring system. The steel girders were supported on temporary W12 steel columns. The columns were grouted to the floors that were to remain, but cored through the existing mezzanine so that the mezzanine could be removed once the column was unloaded. Column locations were coordinated in extremely tight spaces, avoiding active electrical conduits and other utilities. In one location, a column flange had to be notched and the remaining column reinforced for the column to fit in the proper location. At another level, the temporary column forces had to be redirected with a transfer girder cantilevering at each end to avoid interferences. The bases of the columns were founded on the existing 7-foot-thick foundation mat.

The column load supporting the existing two-way concrete flat plate was estimated at 550 kips with full live load, which was included when sizing shoring elements.

Only dead load was considered when determining jacking forces for shoring preloading. Because some of the floors supported by the column that was being shored were occupied during regular library operations, the jacking and cutting of the existing column were performed at night, during off hours. To ensure that loads were

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properly removed from the existing column and transferred to the shoring system, monitoring points were established on the jacking collar. The points were observed for movement during jacking using a construction auto level located away from the jacking site.

Two hydraulic rams, with a common manifold, were used to jack the collar from the steel girders. The common manifold was needed to ensure that the jacks on either side of the column were lifting at the same pressure and, therefore, a balanced force. The jacking was stepped at 25-kip intervals, with a 10-minute hold at 50-kip intervals to allow for monitoring of the structure and to observe any slippage that might occur in the column collar. The maximum predicted jacking force was 320 kips. At 275 kips, the collar had moved up 1⁄16 inch, indicating the column was raised. At that jacking increment, it was decided to lock off the hydraulic rams.

It should be noted that the original column load calculation included an allowance for the existing Concrete Masonry Unit (CMU) elevator shaft above the jacking site to be grouted solid. The in-place shaft was constructed with un-grouted CMU. The maximum predicted jacking force would be 292 kips, or within 6 percent of the actual jacking force after adjusting the calculations to remove the grout.

Also, note that no live load was considered for jacking, even though the areas above the column contain spaces that would normally

Temporary shoring for McKim wall removal, Johnson concrete walls remain for new link (left); Finished McKim Link (right). Courtesy of Bruce T. Martin.

be designed for 100 psf or more. The decision to exclude live load was based on field observation which found many open spaces which were not occupied during the night operations and therefore had little contribution to the actual column load.

To install the new column, the demolition of the existing column required a clean cut where the column met the structure above and below. The initial cutting was performed at the level directly below the jacking site at the top of the column. A 42-inch-diameter hydraulic rotary cutting saw was utilized to make the cut immediately after jacking. After demolition, the new steel column was set, grouted in place, and anchored top and bottom with adhesive anchors completing the column replacement.

Summary Contributing to the success of the transformation of the Boston Public Library Johnson Building was early consideration of the construction process by William Rawn and LeMessurier. Consideration of the “means and methods” within the design, and providing concept shoring and erection procedures, gave the Contractor and their subs confidence to bid and execute the ambitious structural renovation and ensure the Architect’s vision was realized. This level of detail and collaboration continued into the construction phase with well planned, open, and professional dialogue between the design and construction teams, with Becker Structural Engineers

(a)

serving as Consigli’s means and methods engineer. In-house design charrettes, and design and pre-construction meetings, were integral in developing the final scheme and executing the work. Final temporary shoring, bracing, and preloading calculations and design were submitted, reviewed, and discussed with the design team, ensuring transparency in the design approach, expected outcome, and concerns. Shoring, bracing, jacking, demolition, and installation of the new steel framing went smoothly as a result of rigorous design, good planning, frequent field observations, and strong collaboration between the Architect, Engineer of Record, the Contractor’s means and methods Engineer, and the Contractor.▪

(b)

(c)

Column replacement at glass elevator. a) Section of temporary shoring; b) Collar to concrete column connection; c) Temporary shoring above second floor during preloading.

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structural

COMPONENTS

P

art 1 of this article (STRUCTURE, September 2018) addressed code considerations and detailing related to wood-frame shaft walls in multi-family and commercial buildings that are also woodframe. Building on those fundamentals, this article examines fire design requirements, construction constraints, and other potential differences associated with applications such as stairs, elevators, and MEP shafts. With a greater understanding of the nuances, the goal is to better equip engineers to realize the cost, schedule, and other benefits of this increasingly common approach to shaft wall design.

Shaft Wall Applications The three main types of shafts in commercial and multi-family construction are stairs, elevators, and mechanical. Some of the following principals apply to all of these shafts, while some are unique to each.

Shaft Wall Solutions for Wood-Frame Buildings Part 2: Applications

By Richard McLain, P.E., S.E. Richard McLain is a Senior Technical Director in the Project Resources and Solutions Division of WoodWorks. He is Executive Director of the Structural Engineers Association of Vermont and a member of the ASCE Structural Wind Engineering Committee, SEI Blast Protection of Buildings Standards Committee, and NIBS Offsite Construction Council Board. (ricky.mclain@woodworks.org)

Figure 1. Floor framing ledger attached to shaft wall through two layers of gypsum.

Stair Shafts Stair shafts are unique when compared to elevator shafts and mechanical shafts in that they have framing within the shaft (stair and landing framing) that must be accommodated.

Once the typical wall assembly and main floorto-shaft wall detail have been selected, the next detailing considerations involve attaching the stair framing – stringers and landing framing – to the shaft walls. Many of the same considerations exist for main floor-to-wall detailing at this stair framing-to-wall detail. The difference is that a break/ joint in the wall studs is typically not present at the stair and intermediate landing framing-to-wall attachment. Due to this, it is common to run one or two layers of wall gypsum up the face of the wall and attach the stair and landing framing to the shaft wall through the wall gypsum.

To accomplish this detail, a ledger is typically attached to the shaft wall through the layer(s) of gypsum that extend continuously up the shaft with the stair/landing framing hung from the ledger. Note that this configuration requires particular attention to the design of the fasteners attaching the ledger to the wall (Figure 1).

Fasteners installed through gypsum wallboard can be large and difficult to accommodate when supporting larger loads because of the eccentricity on the fastener and the compression capacity of the gypsum. In addition to fastener requirements, regardless of the magnitude of loads, construction sequencing is a significant concern. To address this, some contractors begin by installing STRUCTURE magazine

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Figure 2. Stair landing framing attached to shaft wall through two layers of gypsum.

a strip (or strips) of moisture-resistant gypsum wallboard only where the structure will attach to the shaft wall. After all the framing is installed, the remainder of the shaft gypsum is installed (Figure 2).

Elevator Shafts Many of the same design considerations and wall assembly options that exist for stair shaft walls also apply to elevator shaft walls. Acoustical design considerations are perhaps more pronounced in elevator shaft walls than they are in stair shafts and mechanical shafts. The distinguishing factor in elevator shafts is the design of the rail supports.

In some instances, elevator rails are attached to the structure at each floor level. For this option, a rim joist is typically implemented in the adjacent floor framing for rail bracket attachment. These rim joists provide backing to bolt the connecting plates to the shaft. Additional blocking and strapping are provided around the perimeter of the shaft to transfer the elevator’s horizontal forces into the floor diaphragm. The bracket attaching the elevator rail to the connecting plate should be vertically slotted at each floor level to compensate for shrinkage of the wood framing.

In other instances, the rails can attach at any elevation in the shaft. For this situation, vertical wood posts composed of wood members oriented with their wide face parallel to the wall are typically used for rail bracket attachment. Regardless of the situation, the elevator manufacturer should be consulted for input on the proposed detail.

Most elevator shafts are required to have a hoist beam at the top for installation safety.

The elevator manufacturer specifies the location and required load resistance. In masonry and steel-frame shafts, the hoist beam is typically a structural steel wide flange beam. In wood frame elevator shafts, the hoist beam can be

structural steel or in some cases wood. The elevator manufacturer should be consulted to determine the compatibility of their product with different hoist beam options.

Mechanical Shafts Many of the same design considerations and wall assembly options that exist for stair and elevator shaft walls also apply to mechanical shaft walls.

The main difference is that mechanical shafts are often small enough such that physically getting into the shaft to finish the gypsum is not possible. A common solution includes framing some or all sides of the shaft with shaftliner panels to deal with this situation.

Shaft Walls That Are Also Exterior Walls

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In building types such as multi-family, it is common to have stair and elevator shafts located at the ends and corners of the building. When a shaft wall also forms a portion of the perimeter of the building, the following code provisions in the International Building Code (IBC) apply.

Section 713.6 Exterior walls. Where exterior walls serve as a part of a required shaft enclosure, such walls shall comply with the requirements of Section 705 for exterior walls and the fire-resistance-rated enclosure requirements shall not apply.

Exception: Exterior walls required to be fire-resistance-rated in accordance with Section 1019.2 for exterior egress balconies, Section 1022.7 for interior exit stairways and ramps and Section 1026.6 for exterior exit stairways and ramps.

Section 1023.7 Interior exit stairway and ramp exterior walls. Exterior walls of the interior exit stairway and ramp shall comply with the requirements of Section 705 for exterior walls. Where nonrated walls or unprotected openings enclose the exterior of the stairway and the walls or openings are exposed by other parts of the building at an angle of less than 180 degrees (3.14 rad), the building exterior walls within 10 feet (3048 mm) horizontally of a nonrated wall or unprotected opening shall have a fire-resistance rating of not less than 1 hour. Openings within such exterior walls shall be protected by opening protectives having a fire protection rating of not less than 3â „4 hour. This construction shall extend vertically from the ground to a point 10 feet (3048 mm) above the topmost landing of the stairway or to the roof line, whichever is lower.

As noted in these code sections, shaft walls that are also exterior walls can be rated per the exterior wall requirements. IBC Tables 601 and 602 provide the fire-resistance rating requirements for exterior walls. It is important to note that exterior walls with a fire separation distance of greater than 10 feet are only required to be rated for exposure to fire from the inside face of the exterior walls per IBC Section 705.5. A definition of fire separation distance is provided in IBC Section 202. Following the provisions of the code sections cited above, it is not uncommon to have a non-rated shaft wall along the perimeter of the building. Under this circumstance, the sections of the exterior wall adjacent to the shaft must be rated for a minimum of 1 hour for a minimum of 10 feet away from Figure 3. Stair/exterior walls with options for bracing wall the shaft. The intent of the code plates at stud joints. here is to prevent a fire in the main area of the building from running through Unbraced Joints in Wall Studs at Shafts the unrated exterior wall and then over and into the shaft. See Figure 1023.7(1) of the When a shaft wall is also an exterior wall, the IBC Commentary for example illustrations typical floor framing is not in place on the of this condition. non-shaft side of the wall to brace it against

MCI_5x3.5_02-18.indd 1

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2/19/18 8:44 AM

To address this requirement of using a lower R factor associated with the masonry walls, as well as to address the requirement of SDPWS Section 4.1.6, many engineers are recognizing the benefits of switching shaft walls to wood.

This reduces the seismic forces (lower wall mass) and allows the entire building’s lateral system to use an R of 6.5, while also dealing with issues such as differential movement/shrinkage which can occur between a wood-frame floor and its supporting masonry shaft wall. Switching to wood shaft walls may also be beneficial from the perspective that it eliminates the need for two construction trades and has the potential to speed the construction schedule and reduce cost.

Figure 4. Masonry shaft wall in wood frame building.

out-of-plane wind and seismic forces. Due to this, hinge effects in the wall framing should be considered. A few options exist to address this condition. One is to use the wall plates as continuous, horizontally-spanning members to resist out-of-plane loads. If using this option, the designer should specify that the plates not be jointed in the shaft area. Another option would be to install a structural rim member between the plates with the purpose of spanning horizontally and resisting out-ofplane loads. Where this rim member ends at the end of the shaft, it would be attached to the diaphragm and shear walls to resolve its out-of-plane force reactions. See Figure 3, page 13 for an example of this detail.

Masonry Shaft Walls In some regions of the country, masonry shafts are commonly used in buildings that are otherwise wood-frame (Figure 4 ). In addition to acting as shaft enclosure walls, these masonry walls are often used as shear walls. While this is common practice, there are several issues with mixing masonry shear walls at the shafts with an otherwise light-frame wood structure, notably seismic compatibility of the systems and differential shrinkage.

Figure 5. Masonry shaft wall isolated from wood floor framing by using wood bearing wall.

Differential Movement

Seismic Compatibility ASCE 7-10, Minimum Design Loads for Buildings and Other Structures, Table 12.2-1, lists design coefficients and factors for seismic-forceresisting systems. This table does not include a lateral load-resistance combination for both light-frame wood shear walls and masonry shear walls. Each is categorized separately, and they have significantly different seismic-resistance properties. The seismic response modification coefficient, R, of light-frame wood-sheathed shear walls is 6.5, while the R of masonry shear walls can vary from 2 (ordinary reinforced masonry shear walls) to 5 (special reinforced masonry shear walls). Regardless of which type of masonry shear wall is used, the associated lower R of masonry shear walls will produce higher seismic forces when compared to a wood shear wall system. When using more than one type of lateral-force-resisting system in the same force direction, ASCE 7-10 section 12.2.3.3 requires that the least value of R be used for all systems in the direction under consideration.

Although there are a few conditions that allow this requirement to be waived, in most commercial and multi-family buildings the lower R factor of the masonry shear walls would need to be used throughout the building for the loading direction considered, even for the design of the wood shear walls.

Wood shear walls and masonry shear walls also have inherently different stiffness properties. When using a flexible diaphragm analysis, the diaphragm forces are distributed to vertical-resisting elements based on their tributary area, regardless of their relative stiffness. A flexible diaphragm analysis is typically done for light-frame construction. If accounting for the difference in relative stiffness of the vertical-resisting elements (shear walls) is desired, a semi-rigid or rigid diaphragm analysis would be required. Section 4.2.5 of the American Wood Council’s Special Design Provisions for Wind and Seismic (SDPWS) discusses this in further detail.

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When mixing materials, detailing best practices include consideration of how each of these construction materials will move relative to each other. Wood framing will likely shrink, with the amount varying depending on how the building is detailed, the moisture content of the wood before construction, and the equilibrium moisture content of the site. Masonry will shrink very little if at all (in some instances it can expand). Also, the differential movement between the wood walls supporting the wood-frame floor and the masonry shaft wall may cause floors to slope, finishes to be damaged, or issues at door thresholds to occur.

If using masonry shaft walls in a wood-frame building, the best option for detailing to avoid issues is to isolate the wood framing from the masonry shaft walls. See Figure 5 for an example of this detail. For further information on differential material detailing, see the WoodWorks publication, Accommodating Shrinkage in Multi-Story Wood-Frame Structures.

Conclusion Shaft wall assembly and detail selection should be carefully considered regardless of the material used. The IBC provides ample opportunities for wood-frame shaft walls that should be explored before assuming that other materials are necessary for an otherwise wood-frame structure. A variety of detailing options exist at assembly intersections. This is a positive, as it allows for flexibility in shaft wall solutions and enables the designer and building official to explore options and determine the most appropriate solution for a given project.▪ This article is excerpted from the WoodWorks paper, Shaft Wall Solutions for Wood-Frame Buildings, available at www.woodworks.org.

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RIVETS AND RELOCATION New Life for the Pierceville Bridge

By The Historic Bridge Foundation

T

he 113-foot Pierceville Bridge was built in 1881 to carry SR-1029

over Tunkhannock Creek in Wyoming County, Pennsylvania. The Pierceville Bridge is a rare lenticular truss bridge, so-called because the top and bottom chords of the trusses create a distinctive “lens” shape. The single span bridge is 113 feet 8 inches long with a width of 14 feet 9 inches between centers of trusses. Only five lenticular truss bridges remain in Pennsylvania, including the well-known Smithfield Street Bridge in Pittsburgh. The Pierceville Bridge is also noteworthy because it is the oldest lenticular truss bridge in Pennsylvania and one of the oldest lenticular truss bridges in the United States. It was built by the Corrugated Metal Company of East Berlin, Connecticut, which later changed its name to the Berlin Iron Bridge Company and was a promoter of this bridge type. The company held rights to inventor William Douglas’ patent for a distinctive variety of lenticular trusses. Except for the Smithfield Street Bridge, which was designed

by Gustav Lindenthal, all other surviving lenticular truss bridges in America were designed and built by the Corrugated Metal/Berlin Iron Bridge Company.

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off its abutments and moved to a nearby laydown and disassembly area. The process involved using a 500-ton crane and a specialized steel lifting frame that had been designed and fabricated to act both as a spreader beam system for the lift and as a disassembly/assembly jig (Figure 1). After the crane had moved the bridge and lifting frame to the laydown area, vertical steel columns were attached and the whole structure was laterally braced with guy wires anchored to concrete deadmen. This transformed the lifting frame into a supporting structure from which the truss hung, enabling safe disassembly of the truss. The third and final use of the steel support structure was to provide the structural support for the Figure 1. Removal of the bridge from its original location. Courtesy of McCormick Taylor. containment system necessary to blast the lead-containThe Pierceville Bridge served traffic in its original location ing paint off the bridge to protect workers and the environment until 2005, when deck, stringer, and floorbeam deterioration before disassembly. prompted the closure of the bridge to traffic. The bridge remained After removal of the steel deck stringers, disassembly of the truss closed in this condition, with no repairs made for several years, itself was completed generally from the bottom to the top, with the while a plan could be developed to determine the best course of floorbeams, lower bracing, and lower chords removed first (Figure 2). action going forward. In 2011, the Pennsylvania Department of This was followed by removal of the truss vertical and diagonal memTransportation (PennDOT) initiated a study to reopen the crossing bers, and finally separating the upper chord splices. As a pin-connected to traffic, while also addressing the need for farm equipment, emer- truss, the primary method of disassembly was to remove the pins. The gency vehicles, and school buses to cross the bridge. The bridge’s original pins were designated to be replaced to facilitate this process, maximum twenty-ton rehabilitated capacity, 9.5-foot overhead allowing for flame cutting of the pins into pieces that could then clearance, and 12.3-foot one-lane roadway were obstacles to find- be more easily driven out of the members. The cast iron pin caps ing a way for the lenticular truss to meet these needs. Because the were to be reused and therefore had to be unbolted and saved. Bolts, bridge was considered eligible for listing in the National Register turnbuckles, and threaded rods on the bridge were carefully heated of Historic Places, a Section 106 Review was required by the to expand and loosen them, enabling their removal. Once removed, FHWA to consider feasible and prudent alternatives to meet the the threads of the nuts and bolts that had remained sealed from the project purpose and need, while avoiding adverse effects (such elements for over 135 years displayed a like-new appearance. As the as demolition or alteration) of the historic bridge. During this bridge was disassembled, each part was marked with a numbered process, it was determined that the bridge could not be rehabili- tag to not only assist with reassembly but to also identify any parts tated to the project specifications. However, it was discovered that designated for repairs.

Tunkhannock Township, less than ten miles away, was interested The structural members of the truss are made of wrought iron, in relocating and preserving the historic bridge for pedestrian use confirmed through chemical and metallurgical testing. The higher in Lazy Brook Park over a flood relief channel of Tunkhannock slag (silica) content of wrought iron offers better resistance to corCreek. Through additional discussions during the Section 106 rosion than steel, and this was apparent in the condition of the review, it was agreed that the bridge was a suitable candidate for Pierceville Bridge. Most of the deterioration on the bridge was in relocation and reuse as a primarily pedestrian facility in the park. areas subject to moisture retention such as floorbeams, end posts, Additionally, the project design team agreed that an in-kind resto- and areas that had suffered mechanical damage caused by floods ration rather than rehabilitation was appropriate for this exceptional historic bridge. This meant that above-average attention to detail would be paid to maintaining the original bridge by retaining original materials whenever possible and attempting to use exact replicas for any elements requiring replacement. In particular, this included the use of hotdriven rivets matching the existing bridge rivets, instead of modern bolts typically used on projects involving riveted truss bridges. While already practiced in several other states, this in-kind restoration was a first for PennDOT. The decision to relocate and preserve the historic bridge avoided adverse effects, while also allowing for construction of a new modern bridge on SR-1029 that met the purpose and need of that location.

The $3M project included the relocation and restoration of the Pierceville Bridge and construction of the new replacement highway bridge on SR-1029. The first step of the historic bridge portion of the project was the removal and non-destructive disassembly of the historic truss bridge. After the timber deck was removed, the approximately 54,000-pound truss bridge was lifted Figure 2. Disassembly of the bridge. Courtesy of Bach Steel.

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were replaced with new riveted end posts. The Pierceville Bridge made use of cast iron in a few areas of the bridge including pin caps, end post caps, and builder plaques.

A handful of these cast iron elements on the bridge had been damaged or were missing. These were replicated in steel using surviving original castings as templates.

One minor area of deviation from the original truss design occurred with the lower lateral bracing. With the original bracing deteriorated and damaged from floods, new lower lateral bracing was required. However, the original lower lateral bracing was composed of wrought iron rods with loop-forged eyes. Wrought iron is not available commercially in the United States, and loop-forges are extremely difficult Figure 3. Completed shop repairs of vertical members. New steel is painted grey in this to fabricate using steel. As a result, the new rods were fitted photo. Courtesy of Bach Steel. with a clevis at each end in lieu of forged eyes.

Repairs to the historic truss were completed in a shop setand traffic impacts. Several lower chord eyebars were replaced out ting using new fabricated steel supplied to the repair contractor by of an abundance of caution due to possible cracks in the heads. a PennDOT-approved fabricator. Special provisions in the contract However, the vast majority of eyebars on the bridge were in out- ensured that the rivets and associated repair work were completed standing condition with minimal to no section loss, even in the by a firm with extensive prior experience. Since riveting is not part traditionally troublesome areas at the pin connections where the of typical bridge construction projects today, the standard bridge stacked eyebar heads often develop section loss while in contact prequalification classifications used by transportation departments with adjacent eyebars and members. do not consider riveting experience. Rivets for the Pierceville Bridge The built-up vertical members and portal bracing on the bridge were driven according to standard specifications including ASTM were in good condition overall. However, some sections had been A 502 and ANSI Standard B18.1.2. The rivets were heated in a gas impacted by vehicles, and one vertical member had been crudely forge and driven with hydraulic riveters and pneumatic rivet hamrepaired with welded splices in the angles. Rather than completely mers depending on the location and accessibility of the rivet being replace these damaged members, a partial replacement was under- driven (Figure 4). Care was taken to closely match the new rivet head taken. The impacted and poorly repaired areas of the verticals were size and style (dome top versus countersunk) to the original rivets. removed and replaced with new steel riveted to match the original While repairs to the truss were underway, abutment construction sections (Figure 3). Similar repairs were made to the portal bracing. was accomplished at Lazy Brook Park, the new home for the historic The bridge’s floorbeams are built-up, variable depth, “fishbelly� bridge. A noteworthy aspect of this work was that the new concrete shaped beams. Some of them had section loss on the angles and abutments were faced with some of the original stones that had been at the lateral bracing connections. salvaged from the dry-laid stone abutDeteriorated angles were removed ments at the original bridge location. and replaced with new angles riveted The original stones were transported to the beams. The lateral bracing conto the new site, cleaned, cut down to nection areas were strengthened by an approximate 6-inch thickness, and adding a small riveted plate to proattached with mortar to the freshly vide additional cross-sectional area. cast concrete abutments. This had Once the top chord splices were disasthe benefit of preserving some of the sembled, cracking between rivet holes original abutment materials and also was discovered. The original splice provided a more realistic simulation plates were replaced with slightly of a stone abutment. larger plates to develop a friction conFollowing completion of repairs, nection that would provide a secure the truss parts were shipped to a splice, to avoid having to replace these paint shop to be blasted again and members or rely on a welded repair. painted with a 3 coat inorganic zinc Two parts of the riveted truss were rich paint system. Once complete, the replaced in their entirety with repliparts were shipped to their new home. cas. The longitudinal edge bracing The reassembly of the truss at Lazy that parallels the edges of the deck Brook Park was essentially a reversal had suffered considerable damage of the disassembly process, using the from floods over the years. As a same temporary support and lifting result, the bracing was replaced with frame system. This was followed by new riveted edge bracing, matching the setting of the truss onto its new the original design of two paired abutments with the use of a smaller, angles with v-lacing. Also, the 150-ton crane due to a much shorter end posts of the truss had impact lifting radius (Figure 5). Once the truss damage as well as moisture-related Figure 4. Shop riveting a new edge bracing member was in place on the new abutments, deterioration, so all four end posts using a hydraulic riveter. Courtesy of Bach Steel. the new timber stringer, deck, and STRUCTURE magazine

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Figure 6. The completed Pierceville Bridge at Lazy Brook Park. Courtesy of McCormick Taylor.

Figure 5. The restored truss is placed on the new abutments at Lazy Brook Park. Courtesy of McCormick Taylor.

railing system were installed. The new deck is an all-timber system composed of pressure-treated Southern yellow pine; including 12- x 4-inch deck stringers, 10- x 3-inch deck planks, and pedestrian railing which provides a 10-foot walkway from curb to curb (Figure 6).

The use and detailing (size and spacing) of the structural timbers for the stringers and planks were intended to mimic what is thought to have been originally installed on the truss, in a further effort to be historically accurate.

In its new location, the bridge is a centerpiece landmark and attraction. The bridge was painted a vibrant red to catch the attention of park visitors and highlight the bridge’s distinctive lenticular truss.

Interpretive signage, which is mounted on a salvaged section of one of the original truss bridge end posts, informs park visitors of the bridge’s history and significance. The overall project represents a positive outcome for all involved, by providing SR-1029 with a new bridge meeting the needs of that roadway and preserving one of the rarest historic truss bridges in Pennsylvania.▪

Project Team McCormick Taylor was the consulting engineer for this project, providing services from the initial feasibility study through the final design of both the in-kind restoration and replacement structure. Kriger Construction of Dickson City, Pennsylvania, was the prime contractor for the project. The disassembly/reassembly of the truss, as well as the shop repairs including all hot rivet work, was subcontracted to Bach Steel of Holt, Michigan.

The author is grateful for the assistance from Bach Steel and McCormick Taylor in providing photos and details of the project and the work involved.

The Historic Bridge Foundation was organized in 1998 to advocate for the preservation of historic bridges – our cultural and engineering landmarks that are monuments to the people and communities that built them. Their loss threatens to change the face of our nation. www.historicbridgefoundation.com or info@historicbridgefoundation.com

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October 2018

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Material Difference A Texas Roadway Rehabilitation

Explores Alternative Embankment Despite textural similarities between soil and the kiln-processed lightweight clay aggregate, builders found the kitty-litter like material to be less solid than expected when paving the road, increasing the time required to deliver a smooth surface.

L

arge-scale construction requires making many decisions far in advance of the on-site construction work. Engineers and architects guide clients through a myriad of decisions while creating the drawings and specifications during the design phase of the project. Ultimately, successful projects are about making good decisions before moving the first shovel of dirt and then putting detailed plans and drawings into the hands of skilled people to do the work.

The U.S. 67 bridge over SH 174 outside of Cleburne, TX, needed rehabilitation which was delegated to Ed Bell Construction, as they were working on an adjacent section of highway. The bridge header was settling because deteriorating embankments at either end were causing the pavement to bunch up where the road connects to the bridge. TxDOT had already attempted to rehabilitate the embankments using traditional soil stabilization methods, but moisture issues and settlement continued. As the previous traditional repairs did not stabilize the embankment, a side-byside comparison was requested of two different fill materials, a lightweight aggregate and a Geofoam.”

The first task on the contract was to rehabilitate the actual bridge over SH 174, which was being diminished by failing embankments.

Once the bridge was sound, the embankments were excavated in sections roughly 6 feet deep and 120 feet long, giving builders a fresh, wide trench on either side of the bridge to fill with the alternative fill materials. After each trench was filled, it was to be paved over with the new road, reconnecting both sides to the rehabilitated bridge.

Working with two different structural fills, Ed Bell Construction’s assignment on the embankments was to plan and complete the work on each side of the bridge. They would also install electronic pressure monitors beneath each of the restructured embankments so any future settlement could be monitored independently. The project had the potential for a great learning experience and a chance to rethink common construction processes for soil embankments.

TxDOT indicated they wanted to compare a kiln-processed lightweight clay aggregate and solid, lightweight Geofoam blocks as alternative fills. On the surface, the contractor’s early expectation was that the aggregate materials would behave similarly to soil. However, regarding the geofoam block side, there were concerns.

The lightweight aggregate is supplied from a single manufacturer in North Texas and did not offer many choices. The Geofoam required additional research before the product could be selected. Two different STRUCTURE magazine

Fills and Exposes a Little More Than Expected

By Tom Huempfner

Geofoam suppliers were engaged to provide background on, and competitively secure, the materials. The product ultimately selected was Foam-Control® EPS 22 Geofoam from ACH Foam Technologies, which offers a compressive strength of 7.3 psi at 1% deformation.

Geofoam blocks manufactured at a higher density can withstand up to 18.6 psi. ACH Foam Technologies produces Geofoam in several different densities ranging from EPS 12 with a compressive resistance of 2.2 psi to EPS 46 at 18.6 psi at 1% deformation.

Furthermore, through a collaboration preconstruction process, blocks can be manufactured to size and delivered in a just-in-time delivery sequence to minimize the need for on-site material storage.

Shop drawings of the block configuration pattern were produced to validate the structural soundness of the embankment. To facilitate construction, blocks were numbered and delivered in sequence to ensure easy, precise placement according to the plan.

With traffic diverted, the bridge remediated, and the trenches dug, all that was left to do was build the embankments and document the work. Both trenches were lined with a filter fabric before the new fill materials went in. On the lightweight aggregate side, a truck simply backed up to the hole and dumped in the material, which

Lightweight, yet incredibly strong, the geofoam blocks are moved into position by laborers while engineers assure the stack configuration is built as specified.

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October 2018

the contractor compared to kitty litter in texture. The aggregate was graded flat and covered with more fabric, performing largely the same way soil would under the grader.

On the Geofoam side, first a sand leveling course was put down to produce a flat surface. The Geofoam blocks were laid in by hand with no specialized equipment or skilled labor involved. Two men easily moved blocks as large as 8 by 4 feet, weighing less than 100 pounds.

The majority of the blocks required no modification, but workers were able to quickly customize blocks to fill in around the superstructure using a hand-held hotwire cutter. With the Geofoam stacked and wrapped in fabric just like the aggregate side, there were two readyto-pave embankments made from very different materials. Special sensors had been placed beneath both types of fill to allow TxDOT to closely monitor each material’s settlement post construction.

The next step was to lay the road down. The road plan called for a crushed limestone base subgrade, covered with a hot asphalt mix and topped with 10-inch concrete paving. An initial concern that the Geofoam side would be soft under the equipment was quickly allayed, as the Geofoam provided a solid surface as the base was pushed out with a dozer. On the aggregate side, the fill was softer under the equipment and took longer to surface. The contractor decided to use a lightweight truss screed bridge paver on the aggregate side rather than a traditional concrete paver to put in the final topping, to avoid damaging the embankments with heavy construction equipment, More than five years since the original embankments were built, data suggests both alternative fills have performed adequately. On the lightweight aggregate side, the fill settled slightly more than was initially expected but was still within TxDOT’s acceptable tolerances. On the

ACH Foam Technologies works with engineers and builders to establish precisely-defined purchase orders that can be delivered to job sites in sequences that facilitate accurate and efficient placement of every piece.

Geofoam side, there was a quick initial settlement and then no further movement at all, performing better than expected. The department has been satisfied with the work and the lack of return maintenance since the rehabilitation was completed in 2012.▪ Tom Huempfner is Vice President of Sales and Marketing at ACH Foam Technologies. He has authored many educational seminars and articles for publication, and has conducted numerous educational seminars on molded polystyrene all over the U.S. (tomhuempfner@achfoam.com)

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truly striking structure, the Ritz-Carlton Residences Waikiki Beach Phase 1 project in Honolulu, Hawaii, was certainly deserving of the Post-Tensioning Institute’s (PTI) 2017 Project of the Year Award, presented to the project’s engineer, Baldridge & Associates Structural Engineering, Inc. (BASE). The architects, engineers, and contractors on this project faced several significant challenges but successfully overcame them through the creative use of post-tensioning, resulting in a stunning structure that reflects the Ritz-Carlton’s luxurious brand.

Challenges and Opportunities

Completed Phase 1 on the left and Phase 2 under construction on the right. Courtesy of AirFrame LLC.

The Ritz-Carlton Residences Waikiki Beach, Phase 1

By Steven Baldridge, P.E., S.E., LEED AP, Fernando Frontera, S.E., LEED AP, and Anantha Chittur, P.E., S.E.

Revit model of hanging penthouse slabs, PT slabs, and roof PT transfer slab.

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Gone are the days of rectangular buildings with uniform grids, perimeter beams, punched windows, and eight-foot ceilings. The current trend in residential projects is sculpted floor plans with higher ceilings and more expansive floor-to-floor glazing. This is especially true in luxury projects where spaciousness and views are of utmost importance. In most cities with height restrictions, there has not been a corresponding increase in allowed building heights to match this trend, leaving the development team with the tough decision between maximizing ceiling height or maximizing the number of floors. Caught in the middle is the floor plenum, the space needed for structure and MEP. Reducing this space becomes critical to minimize the tradeoffs between ceiling height and floor count.

Another trend in residential projects is much tighter tolerances including those for deflection, both dead and live load. The clean lines and thin joints of many high-end glazing systems do not provide much play for the many factors that can impact the final floor profile. Projects with aggressive schedules, where glazing may trail floor construction by only a few levels, do not allow for long-term slab deflections to occur prior to the installation of the glazing systems. For architecturally intricate finishes, the old standards of span-to-deflection ratios, such as L/360, may not apply as the glazing supplier, finishing sub-contractors, and architects more frequently require a smaller maximum total deflection regardless of span.

So the challenge becomes, how is structural depth reduced while still maintaining tight deflection criteria?

The old “brute force ignorance” method attempts to solve the problem by adding material. Unfortunately, this method likely results in a bulkier slab section.

Conventional reinforced concrete slab system studies done for the Ritz-Carlton project showed that the tight deflection tolerances could not be met without an extremely thick slab system. The benefits in added stiffness with increased section depth would be offset in part by the added weight and resulting deflection, making resolution of an efficient system impossible.

A more elegant and dynamic solution that utilizes the complementary materials of concrete (strong in compression) and steel reinforcing (strong in tension) is required. Take that one step further by

Load transfer through the 120-foot PT concrete truss.

Deflection plan of typical 7-inch PT slab.

varying the force and location of the steel reinforcing to best match the spans and loadings. Post-tensioned concrete is that solution as it provides the dynamic ability to change how a floor system performs by varying steel tendon drape and quantity. This solution results in thinner slabs which reduces the overall building weight. A lighter building reduces the gravity load demands on the columns, walls, and foundations as well as reducing the seismic demands.

While the basic grid remained the same through the 30 residential levels, there would end up being 11 unique floor plans. Each of these required adjustments to the post-tensioning forces and drapes to provide the required strength, and to manipulate and minimize the deflections of the floor system. The most challenging of these floors were the two-story penthouse levels which were designed with spectacular double-story atriums.

Since atrium and stair openings did not align with any of the supporting structure below, edges of the openings were supported by hanging the penthouse post-tensioned slabs with 32 steel “skyhook” columns from the roof transfer slab. The roof not only had to support the loads from the hanging columns but also the loads from heavy mechanical equipment, rooftop terrace pools, and landscaping at the roof level. This was achieved by utilizing a post-tensioned concrete slab with post-tensioned upturn concrete beams. To maximize views in the corner units, concrete wall supports were replaced with a series of small HSS 5x3 steel columns designed to be hidden in the glazing mullions to give the appearance of 40-foot-long glass walls at each side of the elevator core.

Thinning up the Floor Plenum For the Ritz-Carlton project, the residential floor plan was laid out in a long single-loaded corridor so all interior units would face the ocean, leaving the end corner units with views in three directions.

With a gentle curve, the overall floor plan stretched approximately 240 feet long and 50 feet wide. The slab had 10 spans in the long direction ranging from 13 feet to 32 feet, with the most repetitive span happening only four times at 26 feet 5 inches. Height limitations were a challenge in trying to fit a 38-story building within an envelope no taller than 350 feet.

The team’s goal was to get a 7-inch-thick slab to work to meet these requirements. While the slab system needed to be as thin as possible, it still needed to meet acceptable sound transmission, vibration, and deflection characteristics. With spans up to 32 feet, this would require some creativity. In the narrow direction of the building, standard columns were replaced with thinner, long walls creating a single span condition with cantilevers on both ends. The combination of longwall supports and cantilevers resulted in a very stiff floor system across the width of the building. A side benefit was that the use of thinner loadbearing walls also eliminated any lost sellable area due to structure.

For units with the 32-foot span, a slightly reduced clear height was agreed upon to accommodate a 7¾-inch thick slab needed for these longer spans. The area of this thickened slab, however, had to stop at the inside face of the exterior glazing so that the perimeter of the building appeared to have a consistent 7-inch slab thickness. To further reduce deflections in three of the wider units, a thicker 7-feet-wide x 9¼-inch band slab was hidden in the dropped ceiling in bathroom and kitchen areas. This required coordination with MEP services to not lose ceiling height. The final design of the post-tensioned slabs resulted in slab systems with cladding-sensitive Load transfer through the W column. deflections under 5⁄8 inch.

STRUCTURE magazine

Transfers The residential tower sits on two amenity floors with two pools, restaurants, and a spa deck; five floors of parking; one floor of back-of-house space; and an entry floor with a gracious interior porte-cochere. These diverse uses add seven more unique floors. As the optimum column and wall layouts for each use rarely matched the supporting levels below, offset and sloping columns were incorporated throughout the project to shift support locations through varying floor layouts. In this tower, more than 50 major transitions were required for the vertical elements, with no columns going to the ground in their original location and some elements shifting in plan several times throughout the height of the building.

Adding to the challenge was a lack of structural depth, which in many cases prevented the use of conventional transfer girders except at the podium floors where 17 posttensioned transfer girders in both downturn and upturn configurations were used to reposition tower columns and walls to transition through the parking levels.

The ground floor loading dock space created another significant challenge due to a number of constraints that would not allow the podium vertical elements over the loading dock to extend down to the foundation level. Site constraints included very

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The Post-Tensioning Institute’s Project Awards are presented every two years and recognize excellence in post-tensioning applications. Structures utilizing post-tensioning as a structural component and that were completed or rehabilitated in the past seven years are eligible. Entries are judged by a jury of industry professionals and on seven different

traits: creativity, innovation, ingenuity, cost-effectiveness, functionality, constructability, and aesthetics. For the 2017 PTI Project Awards, 39 projects were considered in six categories: Buildings, Bridges, Parking Structures, Slab-on-Ground, Industrial/Special Applications, and Repair, Rehabilitation, and Strengthening. (www.post-tensioning.org).

tight spaces for trucks to maneuver, city utility easements, and access requirements for an adjacent retail loading area. This meant that an area of the building with spans of 50 to 120 feet had to be columnfree. In order to support the seven floors of podium structure above, four post-tensioned transfer girders and a two-story post-tensioned concrete truss were designed to span the full 120 feet. Post-tensioning was used in the bottom chord of this truss to help control deflections.

The layout of the interior porte-cochere would require the tower columns above to shift over one-half bay. This could have been achieved with deep transfer girders but only to the detriment of a desired double-height ceiling in the area. An alternate series of four sloped columns arranged like a “W” met the layout criteria and ceiling height requirements. Post-tensioning was used in the capping beam at the top of the W to assist in the resolution of large horizontal forces created by the sloping of the tower columns.

Conclusion Construction of The Ritz-Carlton Residences Waikiki Beach, Phase 1 was completed in April 2016. Through the use of post-tensioning along with the hard work and innovation of the architect, the structural engineer, and the contractor, Waikiki now has an iconic structure worthy of both name and location.▪

Steven Baldridge, P.E., S.E., LEED AP is President at BASE and is based in its Honolulu office. (sb@baseengr.com) Fernando Frontera, S.E., LEED AP is an Associate at BASE and is based in its Honolulu office. (ffrontera@baseengr.com) Anantha Chittur, P.E., S.E., is a Senior Associate at BASE and is based in its Chicago office. (achittur@baseengr.com)

Project Team Owner: PACREP, LLC, Los Angeles, CA and Honolulu, HI Structural Engineer of Record: Baldridge & Associates Structural Engineering, Inc. (BASE), Honolulu, HI, and Chicago, IL General Contractor: Albert C. Kobayashi, Inc., Honolulu, HI Architect: Guerin Glass Architecture, New York, NY, and Honolulu, HI Post-Tensioning Supplier: Suncoast Post-Tension, Houston, TX

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Structural rehabilitation Preserving History Rehabilitating the Culver Line Viaduct with Composites

By John P. Busel

W

hen it comes to rehabilitating bridges, an increasingly attractive answer is composite materials. Composites are made from two or more distinct materials that, when combined, produce a stronger and better material offering lighter weight, corrosion resistance, versatility, and overall cost-effectiveness. These qualities are ideal for maintaining fragile historical bridges, safely installing prefabricated decking, and curbing long-term maintenance needs. Perhaps no North American project demonstrates the benefits of rehabilitating with composites more than that of the New York Subway System’s Culver Line Viaduct.

Constructed in 1933 to allow tall-mast ships to pass below it on the Gowanus Canal, the Viaduct is a crucial part of the most extensive rapid transit system in the world. It supports four tracks, two subway lines, and two stations including, at 87.5 feet, the highest aboveground station in the world, the Smith-Ninth Street stop. It spans roughly one mile and serves nearly 90,000 “straphangers” each day.

While the number of New Yorkers who rely on the Viaduct each day has only grown with time, its structural integrity has done just the opposite. The enemy of all structures, water, was able to infiltrate the Viaduct, which lacked any form of waterproofing, and progressively caused significant corrosion of the rebar and steel support members. Over time, as the rusting of the steel expanded, the surrounding concrete began to crack and spall.

By the 1990s, the concrete had deteriorated so severely that the danger posed by sporadically falling concrete forced the NYC Transit Authority to act.

The Authority opted for an initial stopgap measure, wrapping the Viaduct in a black mesh netting system in 2003 to catch and contain the crumbling concrete. This solution stayed in place for nearly a decade due to budget constraints but, in 2012, the Authority accepted that an extensive top-to-bottom approach was required to successfully address water damage, prevent future water infiltration, and conduct structural repairs, all while keeping the line safe and operable for the traveling public. Among the challenges facing the NYC Transit was how to keep this vital link open while conducting repairs below.

To inform this approach, the Authority took stock of the many challenges that any solution would need to address: • The mesh netting system hides unknown substrate conditions.

• The complex geometry of Culver Line Viaduct before. structural elements including the intersection of deep beams, joists, and arched soffits.

• An urban setting that included areas abutting privately-owned apartments, homes, and public parks below.

• The continued service of the Subway’s F and G trains, meaning that all work areas would need to be accessed from below.

• Lack of funds or time to fully rebuild or repair.

The Authority’s planners and engi- Culver Line Viaduct after. neers investigated many options before concluding that the best action was while in others the existing concrete surface to wrap the Viaduct in a unidirectional, was greater than a ¼-inch CSP-9 (concrete 27 ounce/square yard, epoxy-impregnated surface profile) per ICRI Guideline No. glass fiber reinforced polymer (GFRP) fabric, 310.2R and needed to be leveled out with thereby creating a 100% waterproof exterior repair mortars to fortify the surface. Most of membrane that would confine and protect the spalled concrete, especially in overhead all patched areas from further degradation repair areas, was patched with a polymerand significantly extend the Viaduct’s service modified, hand-applied mortar containing an life. Infrastructure specialist Sika Corporation integral corrosion inhibitor. All exposed steel was chosen for the project, not just for their and reinforcing bar was cleaned and protected expertise in FRP composite materials but also with an epoxy-modified, cementitious antifor their overall knowledge of concrete repair, corrosion coating that would provide further protection, and corrosion management. protection from carbon dioxide, chlorides, The decision to pursue a long-term solution and even acid rain. For the larger-scale and that included a flexible, fully-customizable full-depth repairs, a polymer-modified, selfmaterial was immediately validated when the consolidating concrete mixture containing a stopgap black netting was removed. Inspectors migrating corrosion inhibitor was used. All and the contractor began assessing the condi- cracks greater than 10 mils in width were tion of the concrete and the best method of pressure injected with a low-viscosity epoxy repair on a section-by-section basis. As Dave resin prior to FRP repairs.

White, P.E., Vice President of Technical Services These many tasks were made more challengfor Sika and the supplier’s chief engineer, put ing by the complex geometry of the structural it, “When we removed that netting, it was elements comprising the Viaduct, including not a pretty sight. We knew the deterioration its many intersections of deep beams, joists, would be significant, but we underestimated and arched soffits. The specified GFRP fabric what the worst areas would look like and how was flexible enough to allow the contractor to variable the conditions would be.” precisely follow the contours of the concrete In many areas, the corrosion damage was so surface while working high above the ground. severe that large sections of concrete had to Another consideration was the owner’s conbe chipped out with demolition hammers, cern about maintaining the breathability of

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the underlying concrete which, in many areas, had become quite saturated over the years due to the absence of a waterproofing membrane.

Sika, along with its partner Fox Industries, Ltd., met these challenges by custom cutting the fabrics to meet the unique geometric patterns of the elements and maintaining a gap of two to three inches between adjacent sheets of fabric to ensure adequate ventilation of the concrete and prevent full encapsulation of the structure. As White said, “there is simply no other product that would have provided us with the customizability that this highly unique and variable structure required.

Furthermore, we recognize that projects like this are not only huge investments but also extremely serious given the tens of thousands of people who rely on this kind of infrastructure. We conducted rigorous quality control testing prior to and throughout the project to ensure correct, safe, and durable installation.”

On the job, the contractor and owner’s inspector were required to verify their work performance as it took place, and each maintained daily logs to keep track of the temperature, humidity, site conditions, batch numbers, crew members, and quantities of material installed. Also, Sika required a trained field supervisor be present onsite at all times to verify surface preparation, resin mixing, application of repair materials, and protective coatings to verify the strength and condition of the substrate. Specifically, tensile bond testing was conducted on a daily basis to verify the adhesion of the repair materials to the concrete substrate. As required in the specification, all tensile pulls achieved a minimum bond strength of 200 psi (1.38 MPa) with failure in the substrate to fully develop the capacity of the repair materials.

Lastly, Sika ensured proper bond and longterm performance by taking regular surface moisture, temperature, and relative humidity readings while also preparing and sending out cured witness panel specimens to an outside

lab to verify tensile strength, modulus, and elongation properties.

All the testing in the world could not prove the strength, durability, and resiliency of GFRP as powerfully as its ability to weather Hurricane Sandy, which struck New York City midway through rehabilitation in October 2012. Deemed “the storm of the century,” Sandy brought 100-mph winds and a 14-foot storm surge to the City. Moreover, while it shut down the entire NYC Subway System from October 28-30, delaying the project for those three days due in part to workers’ inability to reach the work site, por- Culver Line Viaduct pillar before. Pillar after. tions of the Viaduct that had already undergone repair work were left unscathed.

Considering that the Department of While other structures in the city were dam- Transportation’s latest assessment reported aged significantly (like the Brooklyn Battery that infrastructure improvements require a Tunnel, Hudson River Tunnel, Hunts Point total investment of $926 billion, with transit, Market, etc.), the Culver Line Viaduct was rail, and bus systems alone requiring $26.4 up and running in a few days because the billion a year, they are deciding that the time installed GFRP materials remained intact, for stopgap measures and short-term investdespite not yet being completely repaired. ments is over. Taking to heart the words of In addition to the structural upgrades and the report’s author, “We must increase investwaterproofing of the Culver Line Viaduct, ment in public transportation nationwide which will provide many more years of service because we must take immediate action to to the aging structure, the entire superstruc- bring our transit infrastructure into a state ture was coated and protected with a brand of good repair and provide the world-class new acrylic, waterproof, anti-carbonation service that Americans deserve.” It’s time to coating. Not only does the coating offer give current and future generations of U.S. excellent resistance to UV attack, salt water, travelers the reliable infrastructure our nation and pollution, it improved the aesthetics of requires. Whether used alone or in combinathe viaduct and the neighborhood, especially tion with traditional materials, composites when compared to the black netting that com- have the advanced capabilities and long-term pletely shrouded the structure for the 10 years advantages needed to shepherd in this muchprior to repairs. needed new era of infrastructure.▪ An increasing number of federal, state, and local authorities and engineers across John P. Busel is Vice President, Composites the country are following the NYC Transit Growth Initiative, for the American Composites Authority’s lead and turning to composite Manufacturers Association (ACMA). materials to fortify and preserve essential (jbusel@acmanet.org) Learn more about using infrastructure in a smart, efficient, and costcomposites at discovercomposites.com. effective way.

Close up of Culver Line Viaduct.

Culver Line Viaduct in progress.

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Business Practices Strengthening Your Employer Brand

By Jennifer Anderson

Y

our employer brand is defined by how your current and past employees think of you. If they love you, then your brand can do well in the market; if they do not love you, then you have got some work to do. This article shares four key points about strengthening your employer brand.

1) Information 2) Leadership Understanding 3) Gap Identification and Solutions 4) Communication It is understandable that you have a lot going on in your firm. Clients, project updates, accounting, benefits open enrollment, risk management, ongoing training and education, company holiday party planning; the list goes on and on. What happens if your people leave? Who will take care of your clients? Who will do the work? If your employer brand is weak and your employees are not enjoying working at your firm, you are not only going to be challenged to keep the right employees, but you will also have a hard time recruiting the right people to join your team. Do not take the idea of your employer brand lightly.

Start with Information To strengthen your employer brand, start first by knowing where your brand stands today. The easiest way to uncover your current reputation is through surveys. Survey current employees to understand how they feel about the company. Focus on understanding their perspective about what is vital to them today; do not ask questions about the past – the past cannot be changed. Also, survey past employees, if possible, to get feedback about what they think of the company today. Again, focus on their current perspective – most, if not all, have an opinion about your firm even if they are not working there today. Finally, survey applicants to find out about their perspective of your firm. Gathering feedback from people inside and outside the company will give you a nice snapshot of what your employer brand truly is in the current marketplace.

Leadership Understanding Business leaders need to be in sync with employees on what the employer brand is today. Business owners, principals, key leaders,

and the HR team are highly encouraged to meet and review the employer brand survey findings. Be ready to come to terms with what your brand is today and then determine what you want your employer brand to be going forward. If you are not realistic about any shortcomings, you will continue to lose ground with your people. Once you have decided on what you want your brand to be going forward, share this with your current employees and get their feedback. The leadership team may need to meet a few times before nailing down the exact employer brand.

Respect the Gaps Using the survey information, identify one thing that needs to be improved with your employer brand. Do not attempt to take on all of the problems identified in the survey; instead, pay attention to the really big gaps identified from your research and focus on one first. While you are focusing on solving that first gap, make sure to include the perspective of current employees about what should be done. Choose wisely which employees you will include in this “employer brand task force” and include both those who are new to your firm and those have been around for a while.

Getting multiple points of view will help you to close the gap between the perceptions of both the employees and the leaders. Once you have a solid task force pulled together, give them one job – to strengthen the employer brand and get it completed by a reasonable date. Then get out of their way. Expect regular updates but let them do the work of solving the gaps – employees have loads of answers, so empower them to come up with a solution that will help everyone. Your firm is their employer too, so they want it to improve as much as you want it to. If they do not care, they probably will not accept the task force assignment in the first place. Respect the gaps and respect their abilities. Through this process of strengthening your employer brand,

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do not be surprised if they become your best employer brand ambassadors.

Communicate, Communicate, Communicate Once the “employer brand task force” has been assembled, use that moment to communicate with the rest of the company about how that team is working to make improvements.

When other employees know that there is an effort in place to improve the employer brand, it will help to elevate the task force team’s reputation amongst their peers and likely assist them with gathering more information in their efforts. As the task force identifies the most critical gap to close, communicate it to the whole company as well. Finally, as they develop solutions, communicate again. Your employees will appreciate that you are trying to include them in the process, which in turn will help to elevate your employer brand.

Ultimately, the employer brand is something that ties the whole company together as well as connects the company to the marketplace.

To strengthen your employer brand, start with gathering information, then the leadership team will know where to focus on building a stronger brand and what to communicate to the company and community. If you want to grow as a company, focus on strengthening your employer brand.▪ Born into a family of engineers but focusing on the people side of engineering, Jennifer Anderson (www.CareerCoachJen.com) has nearly 20 years helping companies hire and retain the right talent. (jen@careercoachjen.com)

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To The ediTor Unanticipated Stresses and the Welded Flange Plate Moment Connection

Protected zone = Sh + d Sh Shims, if required

(STRUCTURE magazine, August 2018, by Sompandh Wanant)

T

he August 2018 STRUCTURE article, Unanticipated Stresses and the Welded Flange Plate Moment Connection, presents information and opinions on welded steel moment connections and more specifically on connections in the standard ANSI/AISC 358-16, Prequalified Connections for Special and Intermediate Steel Moment Frames for Seismic Applications. Some of the information in this article may be potentially misleading, particularly regarding statements made on vertical beam shear distribution and weld shrinkage residual stresses.

These issues are addressed below.

Vertical Beam Shear Distribution Mr. Wanant notes in his article that the common assumption that the beam web connection transfers the shear and the beam flange connections transfer the moment is not accurate. Mr. Wanant is completely correct in this observation. The fact that much of the beam shear is transferred by the flange connections is well recognized based on post-Northridge earthquake moment frame connection research and is well recognized in the prequalification of connections for ANSI/AISC 358.

A number of researchers have concluded, based both on experimental evidence and on finite element simulations, that substantial shear is transferred to the column through the beam flange connections. (See, for example, El-Tawil et al. 1998, Lee et al. 1997, Uang et al. 2000). Further, the distribution of shear between the beam web and flanges near the face of the column depends on the connection details, is different in the elastic and inelastic ranges of behavior, and differs depending on the degree of inelastic rotation sustained by the beam. The formulas suggested by Mr. Wanant to predict the distribution of shear between the beam web and flanges (Eqn. 1 and 2 in the article) are based on elastic analysis and therefore are not valid in the inelastic range of behavior. Connection behavior in the inelastic range is of greatest interest in seismic design and the prequalification of connections for ANSI/ AISC 358. Further, even in the elastic range of behavior, Eqn. 1 and 2 do not correctly predict how much shear is carried by the beam flanges at the beam-to-column connection.

It is well understood from research that the beam flanges transfer a large portion of the

Continuity and doubler plates as required

• • • • •

Single-plate web connection

beam shear at welded beam-to-column connections. It is also understood that the shear in the beam flanges not only adds shear stress to the beam flanges but also generates secondary bending Shims, if required stresses in the beam flanges. Additional complexities occur in the state of stress in the beam flanges near the face of the Figure 2. Reprint of Fig. 7.1 (AISC 358-16) bolted column due to column flange bending flange plate moment connection. and due to the influence of weld access holes. These issues have also been extensively movement a short distance from the weld, repstudied in research. Nonetheless, despite resenting a more severe condition of restraint the complexities in the state of stress in the than found in actual construction. Both cases, beam flanges near the face of the column, with or without external restraint, showed very research and testing have also shown that high, yield level residual stresses in both the welded beam-to-column connections can longitudinal and transverse directions, along develop very large cyclic ductility, if prop- with significant triaxiality. These results suggest erly designed and detailed. ANSI/AISC 358 that the external restraint is not needed to generprovides the needed guidance to design and ate high welding residual stresses. Rather, local construct beam-to-connections capable of effects, which are well represented in moment providing satisfactory seismic performance. It connection test specimens, provide restraint would be incorrect to infer that issues such as leading to high residual stresses. For example, shear in the beam flanges have somehow been in the multi-pass welds used to connect a beam overlooked by the research community or flange to a column, shrinkage of each pass is have been overlooked in the development of restrained by the previous passes. Further, difprequalified connections in ANSI/AISC 358. ferential cooling of a weld pass across the width of the beam flange will generate large residual Weld Shrinkage Residual Stresses stresses even in the absence of external restraint Mr. Wanant notes in his article: “Although to the beam, since the portions of the weld that both the RBS and WUF-W are prequalified cool first will restrain the portions of the weld by laboratory testing, the validity of the test- that cool later, for the same weld pass. ing is questionable since it did not represent Studies on the impact of welding residual the actual building frame construction. In stresses on the behavior of welded moment conpreparing the test specimens, there was no nections were conducted by Chi et al. (2000). external restraint in the welding of beam-to- In their study, they used the residual stresses column specimens or the cyclic loading tests.” computed by Zhang and Dong (2000) for the Most moment frame connections are tested case in which the beam flange was completely using specimens constructed with an approxi- restrained. Their research showed these residual mately 10- to 20-foot segment of a beam stresses increase the fracture toughness demand connected to a column. Mr. Wanant is correct on the beam flange welds and can reduce the that no external restraint is normally applied fracture strength of the connection in the presat the free end of the beam during welding. ence of low-toughness welds, as was the case However, it is incorrect to assume this invalidates for pre-Northridge connections. However, these tests. Zhang and Dong (2000) conducted the high level of fracture toughness required a computational study of welding residual in post-Northridge welded moment connecstresses in welded moment frame connections. tions, combined with improved detailing and In their study, they computed residual stresses inspection requirements, provide substantially in a simple butt-welded joint where the plates improved protection against premature brittle had no external restraint. They also computed fracture in beam flange welds and make these residual stresses in a weld connecting a beam welds highly tolerant of large residual stresses. flange to a column, where the beam flange In summary, it is misleading to suggest that was completely restrained against horizontal tests used to prequalify moment connections

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3 PLATE O(," OR 1/,, THICKER THAN BEAM FLANGE) WF BEAM

COLUMN--++-�-!

FLANGE

NOTE: TIGHTEN BOLTS TO THE REQ'D TENSION PRIOR TO WELDING.

SLIP-CRITICAL BOLTED SHEAR PLATE O(,• OR 1/,," THICKER THAN BEAM WEB)

BOTTOM FLANGE PLATE "DIMENSIONS CONFORM TO THE FEMA 350 WELD ACCESS HOLE DETAIL

Seismic Design Criteria for New Steel MomentFrame Buildings (FEMA 2000). Note that a number of details of the WFP connection in FEMA 350 differ from those shown in Figure 3 in Mr. Wanant’s article. Designers with interest in the WFP connection are referred to FEMA 350 for suggested design and detailing rules.

Note, however, that the WFP connection has not yet been prequalified in ANSI/AISC 358.

The AISC Connection Prequalification Review Panel (CPRP) is responsible for prequalifying connections for inclusion in ANSI/AISC 358.

The CPRP, in turn, follows the requirements for prequalification specified in Chapter K of ANSI/ AISC 341-16, Seismic Provisions for Structural Steel Buildings. The AISC CPRP welcomes new connections for possible prequalification and has prioritized connections based on the strength of available prequalification data and based on industry demands. As described above, a limited number of tests were conducted on the WFP connection by Kim et al. (2000), and it is unclear if this test data is sufficient to support prequalification. However, the CPRP may consider the WFP connection for prequalification in the future.

Figure 3. Welded flange plate moment connection.

Closing Remarks

for ANSI/AISC 358 are not valid on the basis of welding residual stresses. Further, as described above, current requirements for welding, detailing, and inspecting moment frame connections provide robust welds with a high tolerance for residual stresses.

In his article, Mr. Wanant also states: “The Bolted Flange Plate (BFP) moment connection (Figure 2) employs the flange plates as connecting elements, and the plates are free to move when welded to the column; therefore, only minor residual stress exists at the welded joint and may be considered negligible.” Based on the research by Zhang and Dong (2000) described above, very high residual stresses would be expected in the welds that connect the flange plates to the column in the Bolted Flange Plate Connection, and certainly should not be considered negligible. Consequently, designers should not relax requirements for welding, detailing, and inspection for the Bolted Flange Plate Connection based on a belief that welding residual stresses are negligible in this connection.

The prequalified moment connections in ANSI/AISC 358 have undergone extensive testing supported by computational and

In his article, Mr. Wanant indicates that the Welded Flange Plate (WFP) connection should be considered for prequalification, but that has not yet been tested for prequalification. The WFP connection was tested as part of the FEMA/SAC project, and is included in FEMA 350, Recommended

The online version of this article contains references. Please visit www.STRUCTUREmag.org.

Submitted by: James O. Malley, P.E., S.E., SECB, Senior Principal, Degenkolb Engineers and Chair, AISC Committee on Specifications.

Lawrence F. Kruth, P.E., Vice President of Engineering & Research, AISC. Michael D. Engelhardt, Professor, the University of Texas at Austin and Chair, AISC Connection Prequalification Review Panel.

TriMet Lafayette Pedestrian Bridge Portland, Oregon

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Welded Flange Plate Connection

analytical studies. The development and prequalification of connections for ANSI/ AISC 358 is informed by the extensive body of research on moment connections conducted after the 1994 Northridge Earthquake, including research on shear in beam flanges, welding residual stresses, and host of other issues. While there are certainly needs for additional research, the existing body of knowledge on moment connection behavior for seismic applications is very strong, and designers can have confidence that the connections in ANSI/AISC 358 will provide robust behavior under seismic loading.■

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Product: Scorpion™ Yielding Connectors Description: Modular, replaceable, standardized hysteretic fuses that provide enhanced ductility and improved performance in the retrofit of seismically deficient structures or for use in the Seismic Force Resisting System of new structures.

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Spotlight Portland’s Sellwood Bridge

By Eric Rau, P.E., and David Goodyear, P.E., S.E., P.Eng. T.Y. Lin International was an Outstanding Award winner for the Sellwood Bridge Replacement project in the 2017 Annual Excellence in Structural Engineering Awards Program in the Category – New Bridges or Transportation Structures.

T

he Sellwood Bridge is the busiest two-lane bridge in the state of Oregon. Located near downtown Portland, the bridge serves as a vital east-west link across the Willamette River. Its predecessor, a four-span, continuous 1,091-foot-long steel Warren Truss was built in 1925 to replace a ferry line that had serviced the community since 1903.

Constructed on a restricted budget, the truss bridge was only 32-feet-wide, with two traffic lanes and a single 4-foot sidewalk. It was not designed for seismic loads or relatively large streetcar vehicles, as other Portland bridges of its era had been. The crossing was also located at the narrowest section of the river, later determined to be the result of an active landslide at the western landing.

In 1960, a 3-foot section of approach girder was removed to mitigate for landslide distress.

Twenty years later, the entire west approach was reconstructed. Continued movement and cracking required Multnomah County, the owner of the bridge, to post a 10-ton load restriction in 2003. Since the bridge was deemed functionally obsolete and structurally deficient, the County initiated a National Environmental Policy Act (NEPA) planning process in 2006 to evaluate options.

Numerous structure types, vehicular traffic and pedestrian configurations, and roadway and interchange alignments were evaluated.

The County also formed a Community Advisory Committee (CAC) to represent public interests throughout the evaluation process. Composed of local residents, Willamette River users, and business owners, the CAC recommendation of a steel deck arch structure on the same alignment as the original bridge was reviewed and approved by the County’s Board of Commissioners in 2011.

The 1,275-foot-long, three-span steel deck arch structure is the signature component of the new, 1,976.5-foot-long Sellwood Bridge.

The arch is formed by two parabolic arch ribs, which are 6-foot-deep welded box sections.

Four Vierendeel braces are provided between the arch ribs in each span, which coincide with the location of spandrel columns and floor beams at the deck level. Longitudinal girders frame between the floor beams, both

of which are composite with the concrete deck. Over 5,000 tons of structural weathering steel was used to construct the steel deck arch structure.

The roadway varies from 63 to 90 feet, carrying two 12-foot vehicular lanes, two 6.5-foot bike lanes/emergency shoulders, and two 12-foot shared-use sidewalks, and is designed to accommodate future streetcar service. Twice as wide as the original structure, the new bridge provides the same vehicular lane configuration as recommended by the CAC to serve the Sellwood community. Traffic mobility is substantially improved with a modern intersection connection to Oregon Highway 43, accomplished through additional turn lanes on the arch bridge paired with 3,600 feet of ramp and retaining wall structures.

The design team developed an innovative anchored shear pile slope stabilization system to mitigate chronic landslide movement. Displacement of the 800-foot-long, 500-foot-wide, 50-foot-deep landslide mass is resisted by 40 vertical six-foot-diameter drilled shafts and 70 diagonal prestressed anchors. The arch foundation is also considered in the 3D analysis, which eliminates displacement at service levels with a 4-inch limit at extreme seismic events.

Seismic design for both the bridge and the landslide mitigation considered multiple earthquake events, including a moment magnitude scale (MMS) 9.0 Cascadia Subduction Zone earthquake. The new bridge was designed to be operational with minor repairable damage through a 500-year earthquake and to withstand a 1000-year event without collapse.

The four foundations of the steel deck arch structure consist of 22 ten-foot-diameter drilled shafts, with tips anchored in bedrock up to 155 feet below the riverbed. Large, 2,500-cubic-yard concrete footings with multiple rows of drilled shafts were provided at the landside piers to resist the thrust of the deck arch. Concrete placement occurred over a 30-hour period, with an active cooling system that piped river water through the footings, used to meet mass concrete requirements.

At the two in-water piers, shafts extend into

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the pier walls in a “perched foundation” configuration. This eliminated the need for large conventional footings located at the river bottom and minimized risks and costs associated with cofferdam construction.

Concrete Y-arms extend from the substructure, with the steel arch ribs located above the 100-year flood stage. The springing connection was hinged through construction and fixed after deck construction to control demand levels on the arch ribs. Ten 4-inchdiameter rods are anchored 15 feet into the Y-arms to provide fixity, post-tensioned to service level demands, and grouted for corrosion protection.

A key aspect of the project was to minimize traffic disruption during construction for the 33,000 daily users of the bridge. The project team developed an innovative solution to this challenge, translating the existing 1,091-foot long truss via hydraulic jacks to temporary foundations constructed on a detoured alignment to the north. The bridge move was accomplished over a 12-hour period and eliminated the need for staged construction of the new bridge.

The project utilized the Construction Manager/General Contractor (CM/GC) method for project delivery. Multnomah County selected Slayden/Sundt Joint Venture as the CM/GC and a design team led by T.Y. Lin International (TYLI), with key subconsultants CH2M and Cornforth Consultants. As the Prime Consultant and Project Manager, TYLI oversaw all engineering disciplines during final design and served as Engineer of Record for the steel deck arch structure. The Sellwood Bridge opened to traffic on February 29, 2016.▪ Eric Rau is a Senior Bridge Engineer for T.Y. Lin International and was the Engineer of Record for the steel deck arch component of the Sellwood Bridge Replacement Project. (eric.rau@tylin.com) David Goodyear is the Chief Bridge Engineer for T.Y. Lin International and was the Chief Engineer responsible for the design of the Sellwood Bridge Replacement Project. (david.goodyear@tylin.com)

2018 NCSEA Special Awards Honorees News form the National Council of Structural Engineers Associations

The Special Awards honor NCSEA members who have provided outstanding service and commitment to the association and to the structural engineering field. Congratulations to this year’s recipients!

NCSEA Service Award This award is presented to an individual who has worked for the betterment of NCSEA to a degree that is beyond the norm of volunteerism.

It is given to someone who has made a clear and indisputable contribution to the organization and the profession.

Barry Arnold P.E., S.E., co-owner and Vice President of ARW Engineers, has been a practicing structural engineer for 29 years. He served on the Board of the Structural Engineers Association of Utah (SEAU) for 5 years and was president in 2004. He also served on the Board of Directors of NCSEA for 7 years and was the 2014-2015 president. In 2007, Barry was selected as the Engineer of the Year by the Utah Engineers Council.

Mr. Arnold has supported the structural engineering profession by serving on several SEAU Committees, including: Wind Committee, Seismic Committee, and the Structural Licensing Committee (acting as Co-chair). He currently serves on NCSEA’s Structural Licensing Committee, and has also served as chair of NCSEA’s 2006 Annual Conference (Summit) Committee, and as SEAU’s Delegate to NCSEA since 2006. Barry is the current Editorial Chair for STRUCTURE magazine and has contributed numerous articles over the years. He has presented sessions on ethics, structural licensure, base plate design, and temperature effects on steel structures at dozens of conferences nationwide.

Robert Cornforth Award This award is presented to an individual for exceptional dedication and exemplary service to an NCSEA Member Organization as well as to the structural engineering profession.

Ryan A. Kersting, S.E., is an Associate Principal at Buehler with over 20 years of experience specializing in performance-based seismic design and evaluation of various types of buildings.

He earned his Bachelor’s degree from Cal Poly, San Luis Obispo and his Master’s degree from University of California, Davis.

Throughout his career, Ryan has consistently served our profession through activities with SEAOC, ATC, NCSEA, ASCE, and other organizations. He is highly respected for his technical expertise, thoughtful leadership, and commitment to the profession. Ryan has served as SEAOC President (2014-15); was on the SEAOC Seismology Committee for many years, including serving as Chair (2010-11); was SEAOC Convention Committee Chair (2007); and, is currently Chair of SEAOC’s Legislative Committee (2013+). In this most recent role, Ryan’s active advocacy and tireless effort has extended the influence of SEAOC and our profession beyond building codes and design projects into the realm of public policy.

NCSEA News

Susan M. Frey NCSEA Educator Award This award, established to honor the memory of Sue Frey, one of NCSEA’s finest educators, is presented to an individual who has a genuine interest in, and extraordinary talent for, effective instruction for practicing structural engineers.

Ronald O. Hamburger, S.E., Senior Principal with Simpson Gumpertz & Heger, has nearly 45 years experience in structural design, construction, education, and failure investigation. He is a past President of NCSEA, SEAOC, and SEAONC, past chair of SECB, and current SEI Governor. He is internationally recognized for his work in failure investigation, building code and standards development, and performance-based design approaches. He served as a member of the select FEMA/ASCE panel that investigated the 9/11/2001 failures of New York’s World Trade Center towers, and was co-Project Director for the FEMA/SAC project on welded steel moment frames.

A past chair of NCSEA’s Code Advisory Committee, he currently chairs the ASCE 7 Standards Committee, is a member of the ASCE 41 Committee, the AISC Committee on Specifications, the Connection Prequalification Review Panel, and the BSSC Provisions Update Committee. He is a recipient of the ASCE Norman Medal and Walter P. Moore award, AISC’s Higgins award, NCSEA’s Delahay award, is a fellow of SEAOC, SEAONC and SEI, and is a member of IStructE. Mr. Hamburger has presented more than 200 lectures on diverse areas of structural engineering practice ranging from earthquake performance and resistance, to blast effects, progressive collapse, and performance-based design. In 2015, he was elected to the National Academy of Engineering.

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This award is presented at the recommendation of the NCSEA Code Advisory Committee to recognize outstanding individual contributions towards the development of building codes and standards. It is given in the spirit of its namesake, a person who made a long and lasting contribution to the code development process.

Jonathan (Jon) C. Siu, P.E., S.E., serves as the Principal Engineer/Building Official for the City of Seattle’s Department of Construction and Inspections, a position he has held for the last 18 of the over 34 years he has been with the City. He has served on many national code and standard development committees since 1995, including several stints on ICC’s IBC Structural and IEBC Committees. He also served on the ASCE 7-05 Seismic Subcommittee, and the Main Committee, and General Requirements and Seismic Subcommittees, for ASCE 7-10 and -16.

Jon has served as a director with the Seattle Chapter of the Structural Engineers Association of Washington and as president of the Washington Association of Building Officials. In his spare time, Jon teaches ATC-20 classes and is helping to organize a statewide disaster response program for volunteer engineers, architects, and building officials. Jon received his B.S and M.S. in engineering from UCLA.

NCSEA News

James M. Delahay Award

These awards will be presented at the Awards Banquet on October 26th during the 2018 NCSEA Structural Engineering Summit in Chicago, IL. Visit www.ncsea.com to learn more about these awards and their recipients.

The California Office of Emergency Services (CalOES) Safety Assessment Program (SAP), hosted by NCSEA, is highly regarded as a standard to train emergency second responders.

This SAP training course provides engineers, architects, and code-enforcement professionals with the basic skills required to perform safety assessments of structures following disasters.

Licensed design professionals and certified building officials will be eligible for SAP Evaluator certification and credentials following completion of this program and submission of required documentation.

The fee for this course has been lowered by more than 40%, making it more accessible to more people. Register for the November 7th course by visiting www.ncsea.com.

NCSEA Webinars

NCSEA’s Structural Engineering Engagement and Equity (SE3) Committee is happy to announce the completion of its new Resource Guide. The SE3 Committee’s mission is to study and promote engagement and equity in the structural engineering profession. Additionally, the NCSEA SE3 Committee facilitates and assists in the creation of local SE3 groups for SEA member organizations to foster the mission of SE3 in their communities. This guide is intended to be a resource for anyone interested in starting an SE3 committee within their Member Organization.

The new resource guide can be found on the SE3 Committee’s page on www.ncsea.com.

Register by visiting www.ncsea.com.

October 11, 2018 BIM and Large Scale Multi-Family Wood Frame Structures – When is it a Fit? Brian Larrabure and Stepan Tresl November 1, 2018 Taking Wood to the Next Level – Application and Design of NLT and CLT Michelle Kam-Biron, P.E., S.E., SECB November 13, 2018 Engineering Structural Glass; an Introduction to the Engineering Structural Glass Design Guide Marcin March, P.E., C.Eng, MIStructE November 29, 2018 Changes to the 2016 TMS 402/602 Building Code for Masonry Structures Richard Bennett, Ph.D, P.E. Courses award 1.5 hours of continuing education after the completion of a quiz. Diamond Review approved in all 50 States.

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October 2018

News from the National Council of Structural Engineers Associations

Next Month’s Training to Become an New SE3 Resource Guide Released Emergency Second Responder

Learning / Networking

SEI Update

The News of the Structural Engineering Institute of ASCE

Captain Scott D. Tingle, NASA Astronaut, recently returned from the International Space Station, delivers the opening address on Engineering a New Frontier – The Next Age of Space Exploration.

View full program and special events, and for the best rate register by October 17 at www.etsconference.org.

Save the date for dynamic learning and networking, collaboration with partners, and fun social events. www.structurescongress.org #Structures19

Join or Renew SEI/ASCE

For innovative solutions and learning, to connect with leaders and colleagues, and enjoy member benefits such as SEI Member Update monthly e-news with exclusive opportunities and resources visit www.asce.org/myprofile or call ASCE Customer Service at 800-548-ASCE (2723).

Advancing the Profession

Don’t Miss – By November 1: O.H. Ammann Research Fellowship for creating new knowledge in structural design and construction – Application Due November 1 – www.asce.org/SEI-Students Advance to SEI Fellow The SEI Fellow (F.SEI) grade distinguishes members as leaders and mentors in the profession. Review criteria and complete your application package by November 1 to be recognized at Structures Congress in Orlando www.asce.org/SEIFellows Nominate by November 1 for SEI/ASCE Awards at https://bit.ly/2NtDYql.

Students/Young Professionals

Students and Young Professionals (35 and younger): Invest in Your Future Join SEI/ASCE for technical, professional, and leadership experience to advance your career and the profession. Apply by January 4 for a scholarship to participate and get involved in SEI at Structures Congress. www.asce.org/SEI “ Thank you! Structures Congress is the best opportunity I have had for career networking, and the first major step in developing my professional network. I was exposed to an abundance of knowledge and learning in a very fun atmosphere. My passion for Structural Engineering has increased dramatically.”

– Matthew Tyler, S.M.ASCE, Virginia Tech, 2018 Student Scholarship Recipient

2018 Student and Young Professional Scholarship Recipients

SEI Futures Fund in Collaboration with the ASCE Foundation Your partner to: • Invest in the future of structural engineering • Support student and young professional involvement in SEI • Provide opportunities for professional development Learn more and give at www.asce.org/SEIFuturesFund.

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October 2018

Announcing the 2018-2019 SEI Board of Governors

Thank you to SEI members for voting in the recent online election for new SEI Board members Randy Bernhardt representing SEI Business and Professional Activities, and Don Scott representing SEI Codes and Standards Activities. Congratulations to Randy and Don, and also to Aimee Corn on her recent appointment

by ASCE President-Elect Robin Kemper as ASCE representative to the SEI Board.

2018-2019 SEI Board of Governors, effective October 1 David W. Cocke, S.E., F.SEI, F.ASCE, SEI President Glenn R. Bell, P.E., S.E., SECB, F.SEI, F.ASCE, SEI President-Elect Laura E. Champion, P.E., M.ASCE, SEI Secretary Robert E. Nickerson, P.E., F.SEI, M.ASCE, SEI Treasurer Cheng Lok Caleb Hing, Ph.D., P.E., F.SEI, F.ASCE, SEI Past President Randall P. Bernhardt , P.E., S.E., F.SEI, F.ASCE Aimee Corn, EIT, A.M.ASCE

Aimee Corn

Joseph G. DiPompeo, P.E., F.SEI, M.ASCE Satyendra K. Ghosh, Ph.D., F.SEI, F.ASCE Ronald O. Hamburger, P.E., F.SEI Satish Nagarajaiah, Ph.D., F.SEI, M.ASCE Donald R. Scott , P.E., S.E., F.SEI, F.ASCE Sivaji Senapathi, P.E., F.SEI, F.ASCE Victor E. Van Santen, P.E., S.E., F.SEI, M.ASCE

Leadership and the Power of Determination “Great deeds start with an idea or a thought. If the idea and/or thought is of great significance to you, take on the challenge, be patient, be determined, and you may surprise yourself with what you can achieve. If you are not a member of an SEI committee, find one that interests you and join it. If you are already a member of a committee, seek a leadership role to make a further positive impact. Get involved! You will be part of shaping the future of our profession.” – Sam A. Rihani, P.E., F.SEI, F.ASCE Check out the latest, full news items at www.asce.org/SEI.

SEI Online

Advocating for Performance-Based Design One of the primary recommendations of the SEI Case for Change: Vision for the Future of SE to “establish a standing committee to champion performance-based building and bridge design codes and standards and the reduction of unnecessary constraint on design” was approved by the SEI Board at their April 21 meeting. Review the report by the task committee chaired by Don Dusenberry on Advocating for Performance-Based Design. https://bit.ly/2CK55sP

REACH SEI MEMBERS WITH SEI SUSTAINING ORGANIZATION MEMBERSHIP

THANK YOU TO SEI SUSTAINING ORGANIZATION MEMBERS

SEI ELITE SUSTAINING ORGANIZATION MEMBERS:

Increase your exposure to more than 30,000 SEI members through www.asce.org/SEI, SEI Update e-newsletter, STRUCTURE magazine, and at SEI conferences year round.

SEI SUSTAINING ORGANIZATION MEMBERS:

Follow SEI on Twitter @ASCE_SEI

Visit www.asce.org/SEIStandards to: • View ASCE 7-22 Committee Meeting schedule and archive • Submit proposals to revise ASCE 7

Errata

SEI Standards Supplements and Errata including ASCE 7.

See www.asce.org/SEI-Errata.

If you would like to submit errata, contact Jon Esslinger at jesslinger@asce.org.

Become an SEI Sustaining Organization Member at www.asce.org/SEI-Sustaining-Org-Membership

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October 2018

The News of the Structural Engineering Institute of ASCE

A special Thank You, on behalf of SEI/ASCE, for those Board Members that finished terms September 30 – for your service and leadership on the SEI Board: John L. Carrato, P.E., S.E., F.SEI, F.ASCE, for service 2014-2018 Andrew W. Herrmann, P.E., SECB, F.SEI, Pres.12.ASCE, for service 2013-2018 Chris D. Poland, P.E., S.E., NAE, F.SEI, M.ASCE, for service 2014-2018

SEI Update

Membership

The News of the Council of American Structural Engineers

CASE in Point

CASE Winter Planning Meeting – SAVE THE DATE!!

The 2019 CASE Winter Planning Meeting is scheduled for February 7-8, 2019, in Tampa, FL. If you are interested in attending the meeting or have any suggested topics/ideas from a firm perspective for the committees to pursue, please contact Heather Talbert at htalbert@acec.org. An agenda will be published in early December!

Did you know?

CASE has tools and practice guidelines to help firms deal with a wide variety of business scenarios that structural engineering firms face daily. Whether your firm needs to establish a new Quality Assurance Program, update its risk management program, or keep track of the skills young engineers are learning at each level of experience, CASE has the tools you need!

The following documents/templates are recommended to review/use if your firm needs to update its current Quality Assurance Program, or incorporate a new program into the firm culture: 962: National Practice Guidelines for the Structural Engineer of Record 962-B: National Practice Guideline for Specialty Structural Engineers 962-C: Guidelines for International Building Code Mandated Special Inspections and Tests and Quality Assurance 962-D: Guideline addressing Coordination and Completeness of Structural Construction Documents Tool 1-2: Developing a Culture of Quality

Tool 2-1: Tool 2-4: Tool 4-1: Tool 4-2: Tool 4-3: Tool 4-4: Tool 4-5: Tool 9-2: Tool 10-1: Tool 10-2:

Risk Evaluation Checklist Project Risk Management Plan Status Report Template Project Kick-off Meeting Agenda Sample Correspondence Letters Phone Conversation Log Project Communication Matrix Quality Assurance Plan Site Visit Cards Construction Administration Log

CASE Member Recommendations The following documents are ones that CASE members use most in their businesses: Sample Contact Documents #1: An Agreement for the Provision of Limited Professional Services #2: An Agreement Between Client and Structural Engineer of Record for Professional Services #8: An Agreement Between Client and Specialty Structural Engineer for Professional Services

Practice Guidelines 962: National Practice Guidelines for the Structural Engineer of Record (SER) 962-B: National Practice Guidelines for Specialty Structural Engineers 962-D: A Guideline Addressing Coordination and Completeness of Structural Construction Documents Sample Toolkits 1-2: Developing a Culture of Quality 5-2: Milestone Checklist for Young Engineers 5-5: Project Management Training

You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

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October 2018

Additional Risk Management Strategies for Bottom-Line Results at ACEC Fall Conference In addition to the CASE sponsored sessions, the ACEC Fall Conference will feature more than 30 advanced business programs, including the following sessions focused on managing a firm’s liability and risk:

Design-Build and Consulting Engineer Risk: The Better Path Forward David J. Hatem, Donovan Hatem, LLP and a panel of experts Ethics in Engineering – An Overview Randy Lewis, XL Catlin How to Deal with Terms and Conditions from Prime Design Firms Stephanie H. Burton, Gibbes Burton, LLC Extreme Weather Events and Resilient Design: Risk Management Considerations for Engineers Patricia B. Gary, Donovan Hatem, LLP; and a panel of experts The Conference also features: • Peter Sheehan will share his expertise on how firms can grow and thrive in today’s challenging market.

• Mick Ebeling will share his inspirational philosophy of creating Technology for the Sake of Humanity.

• Jon Meacham, Presidential Historian, will utilize his knowledge and understanding of historical context to comment on current affairs.

For more information and to register, go to www.acec.org/conferences/fall-conference-2018.

Manual for New Consulting Engineers An HR Favorite for New Hires

ACEC’s best-seller, “Can I Borrow Your Watch?” A Beginner’s Guide to Succeeding in a Professional Consulting Organization offers new engineers a head start in the business of professional consulting.

This essential guide is tailored to the unique needs of engineering firms, and the skills and experiences rookie consultants need to be successful in a large organization, including: • Proposal Preparation • Financial Management

• Client Relationships • Project Management • Staff Management With over 140 pages of consulting expertise, this resource is the perfect addition to any new staffer’s welcome pack or in-house orientation. It can even be a useful resource for more seasoned engineers looking to refine their skills.

To order this book, go to www.acec.org/bookstore. Bulk ordering is available; for more information contact Maureen Brown (mbrown@acec.org).

Follow ACEC Coalitions on Twitter – @ACECCoalitions.

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CASE is a part of the American Council of Engineering Companies

October 28-31, ACEC is holding its Fall Conference at The Bellagio, Las Vegas, NV. CASE will be holding their convocation on Monday, October 29. Sessions include: 10:45 am Understanding Professional Liability Insurance Speaker: Brandon Perry, Victor O. Schinnerer & Company 2:00 pm Professional Liability Claims – From Inception to Resolution Speaker: Charlie Geer, Greyling 3:45 pm From Scope Creep to Profitability: Managing Change, Risk, and Client Expectations Speaker: Andrew D. Mendelson, Berkley Design Professional The Conference also features: • CEO roundtables; • Exclusive CFO, CIO, Architect tracks; • Numerous ACEC coalition, council, and forum events; and • Earn up to 21.75PDHs

CASE in Point

ACEC Fall Conference Features CASE Risk Management Convocation and More!

Structural Forum (When) Will Robots Replace Us?

By Eytan Solomon P.E., LEED AP

K

urt Vonnegut, whose father and grandfather were architects, imagined a computer program called “Palladio” in his novel Timequake (1997). Palladio, the story went, could perform a complete building design according to basic user inputs about the intended use, size, location, architectural style, and even the aesthetic of surrounding buildings. The program output full working drawings – down to “plans for the wiring and plumbing” (presumably structural as well) – in less than half an hour. An architect hears of the program from a friend’s teenage daughter, and he goes to his local computer store to see for himself, bringing a wild test for Palladio: to design “a three-story parking garage in the manner of Thomas Jefferson.” True enough, Palladio quickly and successfully designs the building and additionally offers “alternative plans in the manner of Michael Graves or I.M. Pei.” The architect subsequently commits suicide at the “blow to his self-respect.”

Palladio is not possible yet, but there are signs of its approach through increasing computer capabilities and other ways that technology may dramatically change the profession of structural engineering in the next 5, 10, or 30 years.

Site Visit Drones Camera-mounted drones (or UAVs – unmanned aerial vehicles) are already providing visual access for locations too high, far, deep, or hazardous for humans to view in person, which greatly assists documenting construction progress and existing conditions.

Examples include facades, tower spires, inside partially collapsed structures, across debrisstrewn fields, and down shafts. Photo and video are already available; sample-taking and resistance-testing (for example, hammer sounding) may be possible soon. Drones controlled from a remote office also save travel time to and from the site and provide views that could be overlaid directly with the design drawings for comparison.

3D Coordination BIM programs have been steadily progressing toward “seamless coordination” and “clash detection,” and we should expect ever more useful, rapid, and clever tools for coordination

among design team collaborators. Virtual reality (VR) and augmented reality (AR) will enhance and speed up communication among the design team. Information about architectural finishes and MEP infrastructure will live in the model with embedded structural loading properties. We should expect that structural modeling programs and analysis programs will get better at sharing information, and perhaps move toward one model for all analysis and drawing production.

Automated Tasks Much as a good design spreadsheet permits unlimited and rapid calculation iterations, engineers will continue to outsource many automatable tasks to computers whose processing speeds will only increase. Ideally, this will free up the engineer for more big-picture or deeply-critical thinking. Many standard and custom programs already iterate and optimize for structural member design and load derivations based on project location and governing code, when enough information has been input by the engineer. Perhaps the computer will soon gather and interpret more of the input automatically; for example, deriving dead and live loads directly from the architect and MEP drawings. Comparing old and new versions of drawings, contracts and reports could also be automated so that only important changes are highlighted for the engineer, like when comparing construction submittals to design drawings.

Machine Learning The technology industry is developing ever more sophisticated programs for pattern recognition and translation, sometimes called machine learning or deep learning: not just rote, repetitive tasks, but distilling vast amounts of data in recursive learning loops so that the program is advancing its own insights and capabilities. Currently, computers are competing with (or already “superior” to) humans at driving cars, diagnosing cancer from X-rays, identifying faces, and translating foreign languages. Is the day far off when the computer could not just size the beams and reinforcing bars, but lay out the framing plan, choose the structural system, and negotiate the project contract?

Prefabrication Balfour Beatty, a major construction company in the U.K. and U.S., published a thought-provoking manifesto, Innovation 2050 – A Digital Future for the Infrastructure Industry (2017). One of Balfour Beatty’s visions for coming revolutions in construction is the expanded use, on a very large scale, of prefabrication and preassembly. They see this as necessary to reconcile humanity’s mass urbanization with a shortage of skilled labor and to put tighter control on the often-messy on-site construction process. As the technology and finance industries run away with profits on “scalable” technologies, mass prefabrication could be a similar opportunity for the construction industry. Alternatively, even on the small and custom level, prefabrication can limit time and risk on the construction site and facilitate the increased use of robot “workers” both in the shop and on site.

3D Printing In short, 3D printing employs technology either in material science (weld deposits, laser-zapped powders, and the like) or computer guidance of conventional materials (for example, the CNC-cut plywood creations of Wikihouse). It also reduces the need for skilled human labor on site – sometimes down to zero. As these types of construction methods gain traction and their associated architectures become more ambitious, structural engineers will need to train themselves (and their computer programs) in these materials and design methods.

Do you see these or other trends? The world is rapidly changing and we, as engineers, will need to adapt to the changes… before the robots replace us.▪ A similar article was published in the December 2017 Cross Sections (SEAoNY).

This article is reprinted with permission.

The online version of this article includes a list of suggestions for further reading.

Please visit www.STRUCTUREmag.org.

Eytan Solomon is a Structural Engineer at Silman in New York. (solomon@silman.com)

Structural Forum is intended to stimulate thoughtful dialogue and debate among structural engineers and other participants in the design and construction process. Any opinions expressed in Structural Forum are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, the Publisher, or the STRUCTURE® magazine Editorial Board.

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October 2018

MAPEI provides a world of Concrete Restoration Systems • Concrete Repair Mortars • Corrosion Protection • Construction Grouts • Waterproofing • Sealants and Joint Fillers • Coatings and Sealers

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Project: Center for Naval Aviation Technical Training Complex | Engineer: SMR-ISD Consulting Structural Engineers Photo by Pablo Mason, courtesy of Harper Construction Company

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