STRUCTURE magazine | January 2022

STRUCTURE JANUARY 2022

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Erratum It was brought to STRUCTURE’s attention by a reader that, in the Historic Structures article, Quebec Bridge, The First Failure, 1907 (November 2021), the inset should have read: Cooper loading E30 designates that each axle has a load of 30,000#. That equates to 15,000# per wheel. It also specifies a load on the leading truck and another load on the trailing axle, plus a load for the following freight cars. At the time of the bridge’s design, a loading of E20 was common, but Cooper specified a loading of E30.

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Contents

Cover Feature

JAN UARY 2022

30 BOSTON UNIVERSITY CENTER FOR COMPUTING AND DATA SCIENCES BUILDING

By Nathan Roy, P.E., Irfan Baig, P.E., Jamie Hamelin P.Eng, and Lucy Timbers

The Center building consists of two main portions: the 19-story tower with a twostory deep basement and a 5-story podium with a one-story basement. It was determined that a mat foundation would save over $5 million. This decision had consequences: limiting tower loads by optimizing the concrete core plan size and wall thickness, and using lightweight concrete slabs on metal deck.

Features

Columns and Departments

17 STAMFORD MEDIA VILLAGE

7 Editorial

By Joe Gencarelli, P.E., and Jim DeStefano, P.E., AIA, F.SEI

Wanted: Public Service

A derelict 1920s vintage reinforced concrete factory building has been reimagined. For the newly revamped complex, Brownfield challenges, foundation issues typical of waterfront sites, concrete restoration and fortification, and a requirement for new upper floors gave rise to several innovative structural solutions.

By Brent L. White, P.E., S.E.

20 UCLA’S MARION ANDERSON HALL

By Daniel Tunick, S.E., and Nabih Youssef, S.E.

Expanding the Anderson Graduate School of Management presented a clear challenge: the lack of a viable location. The solution was to construct the new building entirely upon an existing parking structure. In an area of high seismic hazard, the new building and the existing parking structure’s retrofit followed the University of California Seismic Safety policy.

23 SHORING FACILITATING DESIGN AT 100 STOCKTON

By Robert Graff, S.E.

The 100 Stockton Street project reimagines an eight-story former department store into a multi-use building for offices, dining, events, and retail. This reimagination of the building required significant structural shoring to facilitate the design.

26 HISTORIC ALAMEDA HIGH SCHOOL RETROFIT – PART 1

By Nik Blanchette, P.E., Steve Heyne, S.E., and Chris Warner, S.E.

In 2012, the Alameda Unified School District in Alameda, California, made the difficult decision to fence off and vacate all three classroom buildings on the Historic Alameda High School campus due to seismic safety deficiencies. This article describes the long process to rehabilitate and restore these nearly century-old buildings.

45 Engineer’s Notebook The Hidden Cost of Copy and Paste – Part 2

By Jason McCool, P.E.

8 Structural Performance Shear Strength Deficiencies in Concrete Columns – Part 1

By Lawrence Burkett, et al.

46 Historic Structures Niagara’s Upper Falls Bridge Failure

By Frank Griggs, Jr., D.Eng, P.E.

14 Codes and Standards 2021 IBC Significant Structural Changes – Part 3

By Sandra Hyde, P.E., and John “Buddy” Showalter, P.E.

34 Just the FAQs FAQs on ASCE Standards

By Laura Champion, P.E., and Jennifer Goupil, P.E.

36 Structural Loads Snow and Rain Loads in ASCE 7-22 – Part 1

By Michael O’Rourke, Ph.D., P.E., and John F. Duntemann, P.E., S.E.

49 InSights Building Safety Assessments Following the Sparta Earthquake

By Colby Baker, P.E.

50 Business Practices Positioning for Continued Success

By Kacey Clagett and Tiany Galaskas

58 Spotlight Edmonton’s Stanley A. Milner Library

38 Structural Design The Long Road – Part 3

By Matthew Speicher, Ph.D., and John Harris, Ph.D.

41 Structural Analysis Two-Stage Analysis Loophole

By Steven Shepherd, S.E., and James McDonald, S.E.

In Every Issue 4 48 52 54 56

Advertiser Index R esource Guide – Anchor Updates NCSEA News SEI Update CASE in Point

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, the Publisher, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

STRUCTURE magazine is not a peer-reviewed publication. Readers are encouraged to do their due diligence through personal research on topics.

JANUARY 2022

5

EDITORIAL Wanted: Public Service

By Brent L. White, P.E., S.E.

B

y the time you are reading this, the 2021 election season is behind us. Although the recent elections did not have national office implications, offices for elected officials were likely held where the reader lives. Did you participate in any way? Did you take the opportunity to vote? Have you ever considered running for elected office?

You may be wondering why I would be writing about this in an engineering magazine directed primarily to structural engineers. It definitely is not to stir partisan debate. Nor is it to advance any political agenda. Instead, it is primarily to ask the question – “What role should engineers play in public service?” Is this something you have considered personally as an engineer, no matter the stage or your career?

How well are engineers represented in the various areas of public service? Currently, nine engineers are serving among the 535 members of Congress. Six serving state governors have degrees in engineering. Is this a representative swath of engineers relative to the population as a whole? Based on current estimates, there are approximately 800,000 P.E. registrants and 60,000 licensed surveyors in the country.

And some engineers do not have a professional license. In 2020, there were an estimated 200,000 engineering degrees awarded. Over the past 25 years, over 3 million individuals have received engineering degrees. Engineers should be represented at levels at least proportionate to the corresponding percentage of the population as a whole.

Without getting into a deep discussion regarding engineers participating in national politics, do these trends represent engineers at a local level? More importantly, why do I care, and why am I writing about this? I do not intend to suggest that the readers of this article should all have the interest or desire to enter politics on the national stage – although there may be some that do. I know that, personally, I do not have that desire. However, at a more local level, engineers and those with an engineering background have a lot to offer those around them in the public sphere, not necessarily by holding an elected office. Engineers that I know typically do not hesitate to become involved with professional engineering societies. Many engineers also participate with universities on advisory boards and student mentoring. Why not consider becoming more engaged in pursuits not related directly to engineering?

Engineers have specific characteristics that can be helpful in the area of public service and have much to offer. Engineers are problem solvers. It is undeniable that many problems need to be solved.

Problem-solving skills developed by engineers can considerably benefit the public at large beyond the civic value from the day-to-day engineering activities of our jobs. Engineers are used to and skilled at working in teams and addressing challenging problems. Thoughtful, methodical problem solving beyond engineering is a trait that can be very beneficial to the public. The opportunities to be involved and share these skills are not limited to the elected office but extend to various opportunities for local involvement.

Every community has school boards, planning commissions, neighborhood/community councils, city councils, county/township planning boards, various improvement/service districts, and more that can benefit from dedicated, thoughtful, competent service. Engineers participating in these settings can provide much-needed insight and balance to almost any topic and discussion. Prior experience in public service at any level is not necessary. Engineers often have developed skills beyond their engineering expertise to assist them in public service.

An engineering colleague in the eastern U.S. is currently serving as the mayor of his city.

Here are some of his insights into why he decided to seek public office, why being an engineer has been helpful, and advice for engineers considering public service: “As a small business owner, I was involved in local political races for years, and I saw the benefit that the good elected officials brought to their communities. When the opportunity presented itself to run for local office, I looked at the other officials in office and realized that my experiences were underrepresented. This was an opportunity to bring my voice to the people of my town.

Serving in public office is very time-consuming and requires a flexible schedule and the ability to make decisions. Because of that, the vast majority of elected officials are attorneys. However, engineers are problem solvers, and consulting engineers work in teams more than just about any other profession. That experience of working with other disciplines and commonly working towards solving a project or a problem uniquely qualifies engineers for public office.

The public is best served when their elected representatives can listen and improve their communities. The biggest fear people have about running for office is the criticism they receive… it’s always there. However, engineers are always scrutinized for our designs and ideas, but there is no greater satisfaction as an engineer than to say, ‘I designed that.’ If you have that desire to fix things that need fixing and the confidence to know when you need to listen to others, running for office is a very rewarding thing you can do for your community.

If you don’t want to run for office, every town needs more engineers to sit on local boards and committees. Those are the places that have the greatest influence on your town’s future, and engineers can always see the big picture of a project better than any other profession.”

I am not a politician. I am not a polished speaker. However, I appreciate the vast opportunities I have been afforded and have personally felt compelled to become involved in my community. If anyone reading this has felt or feels the same way, I encourage you to not only use your engineering skills to serve professionally but to use those same skills to serve the community (or state, etc.) where you live.■

“

What role should engineers play in public service?

STRUCTURE magazine

Brent L. White is President at ARW Engineers, Past President, Structural Engineers Association of Utah, and Current Chair, CASE. JANUARY 2022

7

structural PERFORMANCE Shear Strength Deficiencies in Concrete Columns Part 1

By Lawrence Burkett, Joe Maffei, S.E., Ph.D., Abby Enscoe, P.E., Marc Steyer, S.E., Mike Wesson, S.E., Ph.D., and Aniket Borwankar

C

oncrete buildings with vulnerable columns are some of the most dangerous structures when earthquakes occur. Since the 1970s, building codes have addressed the detailing of columns that are part of moment frames in high-seismic regions. Research for the Portland Cement Association [Blume et al., 1961] and subsequent studies in New Zealand established the need for close spacing of ties and a capacity design of frame members for shear strength sufficient to cause flexural yielding rather than undesirable shear failure.

While these provisions were required for moment frame columns and beams, in the U.S. it took until the 1997 Uniform Building Code (UBC) before such provisions were required for “gravity” columns, i.e., columns categorized by the structural engineer as not part of the seismic-force-resisting system.

Even in countries with a history of advanced seismic codes, concrete buildings have collapsed because of vulnerable gravity columns. This happened in the U.S. in the 1994 Northridge earthquake and New Zealand’s 2010 Christchurch earthquake.

Project Testing In a recent project at the University of California, San Francisco (UCSF), the retrofit of the seven-story Mount Zion Housing building afforded an opportunity for full-scale laboratory testing of a vulnerable concrete column. In addition to other seismic deficiencies typical of a concrete structure from the 1960s, the building has interior columns that lack a close spacing of ties over most of the column height. The columns are governed by non-ductile shear

Figure 1. Test column. The central portion represents the column between floor slabs of the building.

failure, as assessed by ASCE 41-17, Seismic Evaluation and Retrofit of Existing Buildings.

At most interior columns, fiber-reinforced polymer (FRP) wrap could be applied around the entire column perimeter (i.e., all four faces). This is a typical approach to increasing the shear strength of existing concrete columns. However, interferences at nearly 30% of the interior columns prevented access to one column face, driving the need for a three-sided option. UCSF, the structural engineer of record (SEOR), and the peer reviewers sought input from FRP designers about the potential for a three-sided solution. They agreed that the designer of a three-sided FRP wrap would have to provide testing validation of the structural effectiveness of the retrofit used.

The group envisioned FRP wrapping on three sides of the column and FRP through-anchors instead of FRP on the fourth side. A detail of this type had previously been designed and tested by Aegion/ Fyfe for pilasters.

The FRP subcontractor for the project selected Simpson Strong-Tie to design and provide the FRP, and Simpson developed a proposed detail and test program to meet the requirements established by the SEOR. The testing program included a control column (i.e., with no FRP) and columns wrapped using the three-sided FRP with FRP through-anchors. This article discusses the results of the control column test. The results of the FRP-strengthened column tests will be discussed in a subsequent article.

Column Testing Program Figure 4. Test setup for imposing lateral displacement to the column with fixed-fixed end conditions.

8 STRUCTURE magazine

The testing was carried out at Simpson Strong-Tie’s Tyrell Gilb Research Laboratory in Stockton, CA. The testing included one control column specimen, discussed in this article, and two FRP-strengthened

columns, discussed in a subsequent article. Figure 1 and Figure 2-online show the column design and construction.

Material Properties (Control Column) From the mill certificates of the reinforcement, the Grade 60 #10 longitudinal bars had fy = 69.4 ksi and fu = 98.8 ksi. The Grade 40 #3 ties had fy = 55.0 ksi and fu = 82.5 ksi during field testing. The concrete strength at the time of testing was 2,568 psi, based on the average results of six cylinder tests taken over three days (i.e., two cylinders before, two after, and two on the day of the control column test).

the column clear height. The bottom-of-column displacement measurements are taken from the same independent reference frame using a displacement transducer connected at the bottom of the column clear height. Base slip of the concrete abutment relative to the strong floor is also measured with a displacement transducer. Test results were corrected to remove the small amount of base lateral movement.

All measurements are recorded with a central data acquisition system throughout the duration of each test.

Test Setup and Procedures

Designing the Control Column

For the test program to succeed, the control (i.e., un-retrofitted) column The columns were tested under imposed lateral force and displace- needed to fail in shear; otherwise, it would not be possible to show that ment, with fixed-fixed end conditions (Figure 3-online and Figure 4 ). The fixed base is achieved by clamping the lower section of the specimen to a concrete abutment with a steel plate and threaded rods. The abutment is, in turn, anchored to the laboratory’s strong floor.

The upper section of the specimen is restrained from rotation by two fixtures that each deliver the lateral load to the specimen from a horizontal servohydraulic actuator and horizontal HSS steel tube sections that act as loading struts. The actuators are coordinated and controlled to keep a fixed condition at the top of the column, with the top block of the specimen translating but not rotating.

Each actuator is equipped with a load cell and internal displacement transducer for actuator force and displacement measurements, respectively.

Each HSS loading strut has two strain • Concrete Repair Mortars gages installed, one on each vertical face, • Corrosion Protection • Construction Grouts near the end of the strut adjacent to the • Waterproofing test specimen. These strain gages were • Sealants and Joint Fillers used to record slight differences between • Coatings and Sealers the summation of both actuator load • Epoxy Adhesives cell readings on the opposite end of the • Decorative Toppings HSS struts and the actual applied load • Cure and Seals measured via the HSS strain gages on the • Densifiers test specimen end of the HSS struts. The • Structural Strengthening Products difference in these loads is the friction of the test system (i.e., the friction in the HSS struts sliding over its supports via low friction rails). The friction was also Your single-source provider for restoration, calculated as the difference in actuator strengthening and corrosion protection load readings at displacement reversal points. Such friction was found to be less MAPEI offers a full range of products for concrete restoration, waterproofing than 1% of the applied load. and structural strengthening. Globally, MAPEI’s system solutions have been A cyclic-static history of lateral disutilized for such structures as bridges, highways, parking garages, stadiums placement was applied to the specimens. and high-rises.

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Displacement Measurement The top-of-column displacement measurements are taken from an independent reference frame using a string potentiometer connected one inch below the top of JANUARY 2022

9

Figure 5. Comparison of shear strength predictions for the control (un-retrofitted) column using various models. Column end regions at top (3-inch tie spacing); column center region at bottom (12.8-inch tie spacing).

the retrofit solution prevented shear failure. Accordingly, the authors were careful to look at the range of possible best estimates of shear strength and flexural strength to ensure that shear strength would govern for the control column.

Effect of Axial Load Although many equations for shear strength (such as Equation 22.5.5.1 of ACI 318-14, Building Code Requirements for Structural Concrete and Commentary) have neglected the effect of axial loads, presumably for simplicity, it is well recognized that axial compression increases shear strength. Similarly, axial compression increases flexural strength for

columns with an axial load below the balance point, increasing the maximum shear demand under induced displacement. (Maximum shear demand in this configuration equals twice the flexural strength divided by the column clear height, 2M/L.) Previous research on the shear strength and governing behavior of concrete columns (Kowalsky and Priestley, 2000) concluded that axial compression increases shear strength to a similar extent as it increases flexural strength. Tests of columns having different axial load levels have shown that changing only the axial load does not change the governing behavior mode from shear to flexure or vice-versa.

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10 STRUCTURE magazine

This observation helped justify testing of the column without the additional variable of superimposed axial load, since the omission of the axial load was unlikely to change the governing behavior mode.

Estimating Shear and Flexure In designing the control column to ensure shear failure, it became evident that, for this column, some methods predicted substantially higher shear strength than the ASCE 41 equations (Sezen and Moehle, 2004) (Figure 5). The method that the authors evaluated that gave the highest shear strength for the column was the modified UCSD method (Priestley et al., 2007).

The most lightly reinforced columns in the building have 4-#7 longitudinal bars. Without axial load, the columns have a flexural capacity corresponding to a shear demand of approximately 24 kips.

This just exceeds the ASCE 41 shear strength of 23 kips, making them shear governed by ASCE 41. (The 23 kips includes the Vs contribution of the ties at 12.8-inch spacing; by the letter of ASCE 41, the nonconforming tie spacing means that Vs should be neglected, resulting in an ASCE 41 shear strength of 12 kips.)

However, to have a successful test, the authors wanted to ensure shear failure for the highest predicted shear strength, 39 kips for the UCSD model. To achieve this, and considering the testing uncertainties, the flexural strength was increased to ensure shear demand well above 39 kips.

Increasing the reinforcement from 4-#7 to 4-#11 would create a demand (2M/L) of 41 kips, which was judged not high enough. This led to a choice between 8-#9 or 8-#10 longitudinal reinforcement.

8-#10 were chosen to reliably ensure shear failure, creating an expected shear demand of 56 kips (Figure 6 (online) and Figure 7).

While this heavy amount of longitudinal reinforcement did not occur in the Mount Zion Housing building, the authors have seen similar designs with heavy column bars in concrete buildings in California from the 1960s and 1970s, presumably the result of working stress gravity design of the columns coupled with a desire to limit the size of the column section. The heavy amount of reinforcement in the test column would lead to higher flexural compression strain in the concrete and earlier spalling of the cover concrete, but this was expected at deformations sufficiently larger than those at shear failure.

A concrete mix was chosen that was intended not to exceed f´c = 3,000 psi to further avoid increased shear strength. Overall, it was assumed that the column design with heavy longitudinal reinforcement, low concrete strength, and the deficiency of ties would provide a more rigorous test of the effectiveness of the FRP.

Shear Strength Predictions Figure 5 shows the predictions of three shear strength equations compared to the shear demand coming from the column design with 8-#10 longitudinal bars.

The UCSD and ASCE 41 models consider the degradation of shear strength with displacement ductility, which applies to the test column only in the end regions where the flexural yielding occurs. The bottom graph of Figure 5

Figure 7. Predictions of moment capacity based on flexural and shear strength used to evaluate test column behavior mode. Shear strength per the UCSD model.

shows that with the high shear demand coming from the high flexural strength, all the equations predict that shear failure in the center region of the column preempts any flexural yielding that would occur in the end regions.

Final Column Specimen Design

The column specimen was the same as an actual column in the following respects: • Cross-sectional dimensions (14 inches square), concrete cover to ties (1½ inches), and tie size, shape, and detailing (#3 square perimeter ties with 135-degree hooks) • Column clear height (105 inches) • Tie spacing at column ends (four spaces at 3 inches o.c.)

The column specimen differed from an actual column in the following respects: • No axial load other than specimen self-weight • Tested with fixed-fixed end conditions, eliminating the flexibility of floor structures that exists for the column in the actual structure • Actual concrete compressive strength for specimen, at time of testing, equal to 2,568 psi compared to specified strength on the existing drawings of 3,750 psi • Tie spacing over the mid-height region of the column at 12.8 inches (adjusted from the specified 12 inches to avoid adding another tie) • Longitudinal reinforcement of 8-#10 instead of 4-#7 • Continuous longitudinal bars instead of lap splices Lap splices were eliminated to avoid complicating the objective of the test because the splices would cause signifi cant congestion with the heavy Figure 8. Shear failure in diagonal tension of the control (un-retrofitted) column and subsequent spalling of cover concrete. bars used. Also, the assessment for the JANUARY 2022

11

original building was that slip or failure of the lap splices would not govern the column behavior. (The authors assessed the lap splices of the building per ASCE 41 and the recommendations of Priestley, Seible, and Calvi, 1996. The splices have ties at 3-inch spacing over most of the splice length.)

Control Column Results The control column failed in the behavior mode predicted, as desired: shear failure in diagonal tension in the mid-height region of the column, which had a tie spacing of 12.8 inches (Figure 8, page 11).

The shear failure preempted any flexural yielding of the column. The shear failure did not extend into the column’s end regions, which had a tie spacing of 3 inches. After shear failure, spalling of cover concrete occurred because of the high compression strain resulting from the heavy longitudinal reinforcement.

Degradation of strength after the shear failure was immediate and significant, as shown in Figure 9-online.

Shear Strength As shown in Table 1, the UCSD model closely predicted the shear strength, while the ASCE 41 and ACI 318 equations under-predicted the strength.

All three shear strength equations correctly predicted the actual behavior mode and its occurrence in the mid-height of the column.

This is partly because the specimen was designed to be clearly and reliably governed by shear failure.

Findings and Practical Implications The testing of the control column shows, for this case, that the UCSD model provides a good prediction of shear strength. The ASCE 41 and ACI 318 equations under-predicted the shear strength, by a factor of three in the case of ASCE 41 (12 kips vs. 40 kips).

Therefore, it is worth investigating whether the conservatism in the ASCE 41 and ACI 318 shear strength equations is applicable to other situations and whether it may lead to retrofitting to prevent shear failures that are, in fact, unlikely to occur.

The latter seems to be the case for the Mount Zion Housing building columns, for which shear demand is limited by moderate column flexural strength (4-#7 longitudinal bars are typical for many columns) and by the limited capacity of the floor structure to induce column bending and shear. Figure 10 shows that the building’s interior columns (about 190 total) were retrofitted because they have insufficient shear strength by the ASCE 41 criteria. By the UCSD criteria, they have sufficient shear strength.

Figure 10. Column shear demands for the Mt Zion Housing structure, compared to shear strength criteria.

Based on the test results for this column and the application of the UCSD model to the properties of the actual building columns, as shown in Figure 10, the columns without retrofitting would have been governed by flexure, with good ductility capacity. Therefore, they would not require the FRP wrapping.

The authors would not have suspected the apparent level of conservatism in the ASCE 41 column shear strength criteria had they not had an opportunity to compare the criteria to a tested column.

It is certainly appropriate to have conservatism in shear strength requirements for concrete columns, given the potential for shearfailing columns to cause building collapse. However, a question worth investigating is whether the ASCE 41 equation is overly conservative. This includes the question of what spacing of tie reinforcement should be considered ineffective in contributing to shear strength.

Part 2 of this article will describe the retrofitting of the Mt. Zion Housing structure and report on the testing of columns retrofitted for shear strength using a three-sided FRP wrap with FRP through-anchors.■ Full references, additional graphics, project team, and an expanded Table 1 are included in the PDF version of the online article at STRUCTUREmag.org.

Table 1. Predicted versus actual shear strength (using tested f'c of 2,600 psi).

Lawrence Burkett (lawrence@maffei-structure.com) is a Senior Structural Designer, and Joe Maffei (joe@maffei-structure.com) is the Founding Principal at Maffei Structural Engineering in San Francisco, CA.

Shear strength model

Predicted or actual strength

ASCE 41-17, Equation 10-3

12 kips

ASCE 41-17, Equation 10-3, omitting tie spacing requirement

23 kips

Abby Enscoe (a.enscoe@tippingstructural.com) is an Associate, and Marc Steyer (m.steyer@tippingstructural.com) is a Principal, at Tipping Structural Engineers in Berkeley, CA.

ACI 318-19, Table 22.5.5.1, omitting tie spacing requirement per 18.14.3.1 (applicable to SDC B and higher)

30 kips

Mike Wesson is the Engineering Manager for Tyrell Gilb Research Laboratory at Simpson Strong-Tie in Stockton, CA (mwesson@strongtie.com).

UCSD

39 kips

Test result

40 kips

Aniket Borwankar is the Senior Development Manager for Composite Strengthening Systems at Simpson Strong-Tie in Pleasanton, CA (aborwankar@strongtie.com).

12 STRUCTURE magazine

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Smarter Strengthening Solutions

CODES and STANDARDS 2021 IBC Significant Structural Changes Part 3: Special Inspections (Chapter 17)

By Sandra Hyde, P.E., and John “Buddy” Showalter, P.E.

T

his five-part series (Part 1, STRUCTURE, 1704.6.2 Structural observations for November 2021, Part 2, December 2021) seismic resistance. includes discussion of significant structural 1704.6.3 Structural observations for changes to the 2021 International Building wind resistance.

Code (IBC) by the International Code Change Significance: The new descripCouncil (ICC). This installment includes an tion in Section 1704.6 is intended to provide overview of changes to Chapter 17 on special clear direction for the duties of the structural inspections and testing. Only a portion of the observer. The structural observer is expected to total number of code changes to this chapter observe, gravity and lateral force resisting sysare discussed in this article. More information tems, connection details, and gravity and lateral on the code changes discussed here can be load paths. The clarification is also intended to found in the 2021 Significant Changes to the help address a widespread perception of overlap International Building Code, available from between special inspections and structural obserICC (Figure 1). vations. Special inspections are very detailed IBC Chapter 17 provides various procedures inspections of components and materials within and criteria for inspection, testing, and labeling structural systems. Special inspections require of materials and assemblies. Its key purpose certification and specialized training, but they Figure 1. 2021 Significant Changes to the IBC. is to establish where additional inspections/ do not necessarily require an understanding of observations or testing must be provided and what submittals must how systems are designed to function as part of the overall building. be provided to the building official. The following modifications Risk Category III includes categories of buildings that represent a were approved for the 2021 IBC. Changes are shown in striketh- substantial hazard to human life in the event of failure. Given the rough/underline format with a brief description of each change’s relative risk and hazard, it is appropriate to require that a structural significance. engineer conduct site visits to verify general conformance to the design intent for these structures.

In reading the 2021 IBC code text, it may be noted that there are no Structural Observations longer separate categories for wind and seismic structural observations.

Structural observations are now required for all buildings assigned The requirements for observations were folded into the updated text to Risk Category III or IV. Additional clarification provides clear within Section 1704.6. direction for the duties of the structural observer.

1704.6 Structural observations. Where required by the proviPrecast Concrete sions of Section 1704.6.1, 1704.6.2, or 1704.6.3, the owner or the owner’s authorized agent shall employ a registered design Special inspection requirements are added to Table 1705.3 for precast professional to perform structural observations. The structural concrete diaphragms. observer shall visually observe representative locations of strucChange Significance: The American Concrete Institute’s updated tural systems, details, and load paths for general conformance standard ACI 318-19, Building Code Requirements for Structural to the design intent as defined in the approved construction documents. Structural observation does not include or waive the IBC Table 1705.3 Excerpt Required special inspections and tests of concrete construction. responsibility for the inspections in Section 110 or the special inspections in Section 1705 or other sections of this code.

Inspection Duration [unchanged text omitted for brevity] 11. For precast concrete diaphragm connections or 1704.6.1 Structural observations for structures. Structural reinforcement at joints classified as moderate observations shall be provided for those structures where one or high deformability elements (MDE or HDE) in or more of the following conditions exist: structures assigned to SDC C, D, E, or F, inspect such connections and reinforcement in the field for: 1. The structure is classified as Risk Category III or IV. Continuous a. Installation of the embedded parts 2. The structure is a high-rise building. b. Completion of the continuity of reinforcement 3. The structure is assigned to Seismic Design Category E across joints. and is greater than two stories above the grade plane. c. Completion of connections in the field.

3.4. Such observation is required by the registered design professional responsible for the structural design.

12. Inspect installation tolerances of precast concrete diaphragm connections for compliance with ACI Periodic 4.5. Such observation is specifically required by the 550.5. building official.

14 STRUCTURE magazine

Concrete and Commentary, has new provisions for designing precast concrete diaphragms in Section 18.12.11. The new ACI 318 Section 26.13.1.3 requires special inspection of panel placement and reinforcement in precast concrete diaphragms assigned to SDC C, D, E, or F using moderate or high-deformability connections. These diaphragms are also required to comply with the requirements of ACI 550.5-18, Code Requirements for the Design of Precast Concrete Diaphragms for Earthquake Motions and Commentary. ACI 550.5 has special inspection requirements for precast concrete diaphragm connections and reinforcement at joints classified as high deformability elements or moderate deformability elements.

A special inspector watches the installation of the embedded parts, checks the completion of the continuity of reinforcement across the joints, as well as completion of field-built connections when structures are assigned to Seismic Design Categories C, D, E, and F. ACI 318 Section 26.13.1.3 also requires that installation tolerances of precast concrete diaphragm connections be inspected periodically for compliance with ACI 550.5.

To match these new requirements, the 2021 IBC has added two requirements to Table 1705.3. Item 11 is added as a conservative synthesis of the two requirements from ACI 318 and ACI 550.5. A continuous special inspection is required onsite during the installation of precast concrete diaphragms for moderate and highly deformable joints with a focus on the reinforcement extension through the joint, verification of embedded part location, and full connection of the diaphragm elements to one another and the rest of the seismic force-resisting system. Item 12 mirrors the ACI 318 requirement to check diaphragm element and connection minimum and maximum distances against the tolerance requirements of ACI 550.5.

Mass Timber Installation and connection of mass timber elements in Types IV-A, IV-B, and IV-C construction requires special inspection.

1705.5.3 Mass timber construction. Special inspections of Mass Timber elements in Types IV-A, IV-B, and IV-C construction shall be in accordance with Table 1705.5.3.

Table 1705.5.3 Excerpt Required special inspections of mass timber construction.

Inspection

Duration

1. Inspection of anchorage and connections of mass timber construction to timber deep foundation systems.

Periodic

2. Inspect erection of mass timber construction.

3. Inspection of connections where installation methods are required to meet design loads.

• Verify use of proper installation equipment.

Threaded fasteners

• Verify use of pre-drilled holes where required.

Periodic

• Inspect screws, including diameter, length, head type, spacing, installation angle and depth.

Adhesive anchors installed in horizontal or upwardly inclined orientation to resist sustained tension loads.

Continuous

Adhesive anchors not defined in preceding cell.

Bolted connections.

Concealed connections.

Periodic

1705.20 Sealing of mass timber Periodic special inspections of sealants or adhesives shall be conducted where sealant or adhesive required by Section 703.7 is applied to mass timber building elements as designated in the approved construction documents.

Chapter 35 ASTM D 3498 Standard Specification for Adhesives for Field-Gluing Plywood to Lumber Framing for Floor Systems Change Significance: Special inspection provisions are added to Section 1705 for mass timber elements in Types IV-A, IV-B, and IV-C construction. The special inspections are similar to requirements for other prefabricated systems such as precast concrete and structural steel.

The specific elements requiring special inspection for construction Types IV-A, IV-B, and IV-C include: 1) Connection of mass timber elements to timber deep foundation elements. These connections are critical to transferring loads from the mass timber elements to timber piles, particularly lateral loading. Connections to concrete foundations are addressed in IBC Table 1705.3 for concrete special inspections.

2) Erection of mass timber elements. Similar to precast concrete, tall wood buildings utilizing prefabricated elements need verification that the correct elements are placed in the right location in accordance with the design drawings.

3) Specialized connections between mass timber products that utilize threaded, bolted, or concealed connections are similar to concrete connections. The strength of many connection designs is predicated on specific screw lengths and installation angles. Bolted connections require specific diameters and, for lag screws, specific lengths. Concealed connectors, many of which are proprietary, must be installed correctly for structural performance.

4) Adhesive anchorage installed in horizontal or upwardly inclined positions resisting sustained tension loads requires a continuous special inspection. This is necessary because of issues with creep in the adhesives under long-term tension loading. All other adhesive anchors need only be inspected periodically.

If, in the judgment of the building official, there are other unusual items not covered in Table 1705.5.3, the existing text in Section 1705.1.1, Special Cases, requires special inspection of these items as well. The same section also says the building official can require special inspections where a manufacturer’s installation instructions prescribe requirements not contained in the code. For example, field-glued mass timber beam or panel splices, while currently rare in North America, may become more prevalent in the future. Section 1705.1.1 would allow the building official to require special inspection for either proprietary or non-proprietary field-glued splices. Additionally, many design engineers specify the need for special inspections for unusual conditions in their structural notes in the construction documents and the statement of special inspections.

The new Section 1705.20 requires periodic special inspection of sealants or adhesives where sealant or adhesive required by IBC Section 703.7 is applied to mass timber building elements as designated in the approved construction documents. New Section 703.7 requires sealing between mass timber elements in Type IV-A, IV-B, and IV-C construction to resist the passage of air at the following locations: 1) at abutting edges and intersections of mass timber building elements required to be fire-resistance rated, and 2) at abutting intersections of mass timber building elements and building elements of other materials where both are required to be fire-resistance rated. Sealants JANUARY 2022

15

soil conditions, or inadequacy of construction procedures, a deep foundation element should be load-tested during installation to assure that there are no material defects (Figure 2). The 2018 IBC already addressed visual special inspection of deep foundations in Sections 1705.7, 1705.8, and 1705.9. In the 2021 IBC, tests are added to the new Section 1705.10 to provide a means to assess portions of deep foundation elements that cannot be visually inspected. Testing may be done by impact, thermal imaging, or ultrasonic tests. The applicable ASTM standard is used to define the procedures for the appropriate test.

Steel Storage Racks Figure 2. Verification of deep foundation element strength. Courtesy of Chaiyaporn114.

must meet the requirements of ASTM C 920, Standard Specification for Elastomeric Joint Sealants, and adhesives must meet the requirements of ASTM D 3498.

Special inspection of mass timber sealing requirements does not apply to “joints” as defined in Section 202. These joints have their own requirements for the placement and inspection of fireresistant joint systems in Section 715 and special inspections in Section 1705.17. Joints are defined as having an opening designed to accommodate building tolerances or allow independent movement. Panels and members that are connected do not meet the definition of a joint since they are rigidly connected and do not have an opening.

Lastly, some mass timber panels are manufactured under proprietary processes to ensure there are no voids at intersections. Where this proprietary process is incorporated and tested, there is no requirement for a sealant or adhesive.

Structural Integrity of Deep Foundations An engineering assessment must now be done when installed deep foundation elements appear to be understrength due to quality, location, or alignment.

1705.10 Structural Integrity of Deep Foundation Elements.

Whenever there is a reasonable doubt as to the structural integrity of a deep foundation element, an engineering assessment for structural integrity shall be required. The engineering assessment shall include tests for defects performed in accordance with ASTM D 4945, ASTM D 5882, ASTM D 6760, or ASTM D 7949 or other approved method.

Chapter 35 ASTM D 5882-16: Standard Test Method for Low Strain Impact Integrity Testing of Deep Foundations ASTM D 6760-16: Standard Test Method for Integrity Testing of Concrete Deep Foundations by Ultrasonic Crosshole Testing ASTM D 7949-14: Standard Test Methods for Thermal Integrity Profiling of Concrete Deep Foundations Change Significance: Most foundation failures are caused by inadequate soil bearing or lateral capacity. Section 1705.10 addresses a less common failure – lack of structural integrity in a deep foundation element due to material defects. Significant defects may affect the structural strength of deep foundation elements; therefore, the defects must be detected and corrected prior to the construction above ground.

When the integrity of a deep foundation element is in doubt, for example, due to issues in alignment during installation, problematic

16 STRUCTURE magazine

Steel storage rack special inspection duties have been clarified with the addition of special inspection tasks.

1705.13.7 Storage racks. Periodic special inspection is required for the anchorage of storage racks that are 8 feet (2438 mm) or greater in height in structures assigned to Seismic Design Category D, E, or F. Steel storage racks that are 8 feet in height or greater and assigned to Seismic Design Category D, E, or F shall be provided with periodic special inspection as required by Table 1705.13.7.

Table 1705.13.7 Excerpt Required inspections of storage rack systems.

Inspection

Duration

1. Verify materials used comply with one or more of the material test reports in accordance with the approved construction documents 2. Fabricated storage rack elements 3. Installation of storage rack anchorage

Periodic

4. At final inspection of the completed storage rack system, to indicate compliance with approved construction documents

Change Significance: The design of the components that go into a storage rack is based on a minimum steel thickness and minimum yield strength. It is imperative that these minimum properties are included within the design for fabrication of the components and considered in storage rack installation. Storage rack systems may have complex load paths. Installation must comply with approved drawings to create the necessary load paths. Verification must be made of material minimum quality requirements during fabrication and proper anchorage during installation.

Changes clarify that periodic special inspection is required for steel storage racks, regular or cantilevered, that are eight feet or more in height in Seismic Design Category D, E, or F locations. In Chapter 2, Definitions, a definition for cantilevered steel storage racks is added for clarity.

Conclusion Structural engineers should be aware of significant structural changes that have occurred in the 2021 IBC. Since structural engineers are often responsible for developing the statement of special inspection, awareness of these changes is essential.■ Sandra Hyde (shyde@iccsafe.org) is Managing Director, and John “Buddy” Showalter (bshowalter@iccsafe.org) is Senior Staff Engineer, both with ICC’s Product Development Group.

Figure 1. Stamford Media Village fronts on a barge canal.

STAMFORD MEDIA VILLAGE A Rhinestone in the Rough

By Joe Gencarelli, P.E., and Jim DeStefano, P.E., AIA, F.SEI

S

tamford Media Village is not just another nondescript fivestory office building (Figure 1).

What is so unique about this project? Everything!

Rewind to 2018 when Wheelhouse Properties acquired a derelict 1920s vintage reinforced concrete factory building located at the South

Figure 2. The crumbling concrete carcass before the transformation.

end of Stamford, Connecticut – a rhinestone in the rough. The site is situated along a barge canal adjacent to Long Island Sound and is surrounded by redeveloped mill buildings.

DeStefano & Chamberlain, Inc. was engaged as the structural engineer for the project. After a detailed condition assessment of the existing derelict building, the recommendation was “knock it down!” That was not an option that Wheelhouse Properties would consider. Instead, they were determined to save the structure. Their vision was to add three stories on top of this two-story crumbling carcass of a building and transform it into something spectacular (Figure 2). To do so would require innovative structural solutions.

Stamford, Connecticut, has long been known for its corporate office buildings, but today, downtown Stamford is known more for its “office space for lease” signs which can be seen everywhere. So why build another office building in a market with a glut of unleased space? Wheelhouse Properties had a different vision of how today’s office space should look.

Gone are the days of the cookie-cutter office building with cubicles, typing pools, and corner offices under acoustical ceilings and fluorescent lighting. Instead, the new wave of successful companies craves funky spaces with character. Old industrial buildings with architecturally exposed structures are now the rage.

The existing concrete structure was certainly not a gem, but it could pass for a rhinestone once polished up. Exposing the restored concrete structure, along with HVAC ductwork and sprinkler piping, resulted in precisely the kind of unique space that was desired.

It was important to Wheelhouse Properties that Stamford Media Village be a fun place to work. All work and no play JANUARY 2022

17

make for dull tenants. Office space needed to be adaptable to throwing a party as well as a corporate board meeting.

After a long 12 months of planning, designing, more planning, more designing, and another 28 months of construction, their vision turned to reality, like a scene out of a Transformers movie. The newly revamped 133,000 square foot complex has brought a fresh outlook to businesses and the work-play environment. Tenants include TV studios, a dog-friendly microbrewery/restaurant, and an organic market. In addition, Wheelhouse properties reserved the top floor for their own offices.

Brownfield Challenges A site assessment revealed that the soils below the site were contaminated with a toxic stew of hazardous and corrosive compounds, including polychlorinated biphenyls (PCBs), necessitating extensive and costly remediation.

Stamford’s south end is underlain by a stratum of coarse-grained glacial outwash deposits known as the Rippowam Aquifer. Consequently, waterborne contaminants found on any site in the area are free to migrate across the region and into the waters of Long Island Sound.

Foundations

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As is typical with waterfront sites in the area, subsoil conditions consist of uncompacted fill over a stratum of organic silt over coarse-grained glacial deposits. Test pits revealed that the existing building was supported on spread footings bearing on the fill layer above the organic silt. Groundwater was near the bottom of footing elevation at low tide.

Although the fill and organic silt are not suitable for supporting foundations, it was determined that in the 100 years since the building was built, settlement due to primary and secondary consolidation of the organic silt had run its course. Consequently, where foundations were not being subject to additional loads, no remedial work was needed. However, the foundations that would be carrying the load

18 STRUCTURE magazine

Figure 3. Custom pile driving rig.

of the additional three stories would need to be underpinned with deep foundations.

Given the limited vertical clearance, trying to install deep foundations within the building would take some ingenuity. Drilled micro-piles extending to bedrock were first considered; however, the contaminated drill spoils would have required costly remediation and disposal. The underpinning solution settled on concrete-filled steel pipe piles driven to bedrock. The piles were spliced every five feet with compression couplings that eliminated the need for field welding. The foundation contractor, Norwalk Marine Contractors, fabricated a custom rig with a pile hammer secured to the arm of an excavator that could operate inside the existing structure with restricted headroom (Figure 3).

Concrete Restoration The derelict concrete structure, which is a one-way concrete joist framing system, was in rough shape. It was reinforced with twisted square bars called “Ransome Bars” after their inventor, Ernest Ransome.

Due to the proximity to the canal, the crawl space below the building was frequently inundated with tidal water. As a result, most of the ground floor concrete structure was severely spalled and deteriorated from immersion in seawater.

This made it an easy decision to demolish the floor to facilitate removing the contaminated soil under the building and installing the driven pipe piles.

The existing structure was devoid of a lateral load resisting system. Reinforced concrete shear walls were cast around the stair and elevator shafts to resist wind and seismic loads.

The column bays were 20 by 20 feet which is somewhat close for marketable office space. The column bays for the new upper floors were made 20 by 40 feet to improve the rentability of this space – consequently, only half of the existing concrete columns needed to be fortified and underpinned.

The concrete restoration specialists from Structural Technologies were brought in to perform the column fortification. These columns, which are octagonal in shape, were jacketed with 3 inches of concrete and reinforced with vertical corner bars and ties. The concrete floor was cored in the corners of each column being fortified to install continuous vertical column bars. Selfconsolidating concrete (SCC) was pumped into the column jackets through the core holes (Figure 4).

Where the concrete was spalled and deteriorated, Culbertson Company of New York, a concrete restoration firm, was brought in to perform the restoration. High-pressure water blasting, also known as hydro blasting, was used to remove old paint, corroded reinforcing, and loose and flaky concrete. Bonding agents and patching compounds were used to restore concrete surfaces to their original shape.

Figure 4. Concrete columns were reinforced and jacketed to increase capacity.

Mass Timber Hybrid The new upper floors needed to be framed with an architecturally exposed structure with character.

Mass timber construction was the obvious choice for the 3-story vertical expansion, which crowns the building. The 20- by 40-foot column bays did not lend themselves to a pure mass timber solution since very deep glulam timbers would be needed to span 40 feet.

A hybrid of Architecturally Exposed Structural Steel (AESS) and Cross-Laminated Timber (CLT) panels was proposed. At the time that the hybrid solution was conceived in 2018, no hybrid structure of this kind had been built (Figure 5). Upon completion, Stamford Media Village became the second hybrid Mass Timber/Structural Steel building in New England.

Using a hybrid system resulted in significant advantages. It resulted in a lighter structure which in the end saved foundation underpinning costs. The combination of AESS and CLT panels resulted in a cool-looking, unique structure (Figure 6 ).

Stamford Media Village has been an enormous success that has exceeded the owner’s expectations, due in no small part to the innovative structural solutions employed to meet a host of project challenges.■

Figure 5. Cross-Laminated Timber (CLT) panels being erected over a structural steel frame.

Joe Gencarelli is an Associate, and Jim DeStefano is the President of DeStefano & Chamberlain, Inc., located in Fairfield, CT.

Project Team Owner: Wheelhouse Properties Structural Engineer: DeStefano & Chamberlain, Inc. Architect: CPG Architects

Figure 6. Architecturally exposed hybrid structure awaiting a tenant.

JANUARY 2022

19

UCLA’s

MARION ANDERSON HALL

By Daniel Tunick, S.E., and Nabih Youssef, S.E.

Figure 1. Exterior view of the completed building. Courtesy of Paul Turang.

W

hen the University of California, Los Angeles (UCLA) identified a need to expand their existing Anderson Graduate School of Management (AGSM), they were presented with a clear challenge: the lack of a viable location to place a new building on a campus with limited square footage. The solution was to construct the new building entirely upon an existing parking structure adjacent to the School of Management.

The new UCLA Marion Anderson Hall (MAH) is an $80M project that provides a LEED platinum, 64,000-square-foot, 4-story building to expand the capacity of the existing AGSM initially built in 1995. The new building supports large classroom areas, workspaces, offices, and sizeable auditorium/event spaces. MAH was shaped and cladded to blend effortlessly with the adjacent complex of the AGSM buildings (Figure 1).

The building presented the unique challenge of being constructed entirely upon an existing 6-story 430,000-square-foot reinforced concrete parking structure, which remained partially open throughout the construction of the new building above. The parking structure was originally built in 1959 but had seen multiple major structural updates over the decades. Updates included the complete removal of an adjacent hillside and associated perimeter retaining walls, voluntary

Figure 2. Longitudinal section across existing & new buildings.

20 STRUCTURE magazine

seismic retrofits, and even large floorplate sections removed to accommodate the original AGSM construction.

The existing parking structure is constructed on a sloping site, such that the top two levels are offset from the bottom 4 levels. The MAH building is constructed directly at this step between the upper and lower levels. Therefore, the new building has vertical support at two different elevations of the existing parking structure; each serves as an entrance – the 4th and the 6th floor of the parking structure. The new building also ties in horizontally to the 6th floor of the parking structure, which aligns with the first raised floor of the new building (Figure 2).

Seismic Criteria The University of California Seismic Safety policy (UCSSP) provides the governing seismic criteria for all buildings on University of California premises. The project is located in an area of high seismic hazard, and seismic design criteria for both the new building and seismic retrofit of the existing parking structure were per the UCSSP. For new buildings, UCSSP requires compliance with the current seismic provisions of the California Building Code (CBC). Because

the new MAH building contained adult education facilities with an occupant load greater than 500, the building was assigned to Risk Category III per CBC Table 1604.5. In addition, the existing parking structure completely supported the new MAH building above, and consequently, it was also treated as Risk Category III. For existing buildings, the UCSSP generally refers to the CBC, which in turn references the American Society of Civil Engineers’ ASCE 41, Seismic Evaluation and Retrofit of Existing Buildings. The ASCE 41 performance objectives for the existing building were “Damage Control” @ BSE-1N and “Limited Safety” @ BSE-2N. BSE-1N/2N are ASCE 41 hazard levels that correlate to the Design Basis Earthquake (DBE) and Risk-Targeted Maximum Considered Earthquake (MCER) from ASCE 7, Minimum Design Loads for Buildings and Other Structures.

The UCSSP also required an independent seismic peer review which was conducted for the overall project, reviewing both the new building and the retrofit of the existing building.

The overall project included a new structure upon an existing structure, vertical and horizontal combinations of lateral systems of different types and eras, horizontal connections of new floors to existing floors, and gravity and seismic demands delivered from the new structure above to the existing structure below. The seismic criteria also spanned across both ASCE 7 and ASCE 41. A customized two-tier design/analysis procedure incorporating both ASCE 7 and ASCE 41 was utilized to accurately capture the structures’ behavior to accommodate these unique circumstances while also satisfying the UCSSP requirements for both new and existing structures. The first tier consisted of a linear dynamic analysis for the new building design per ASCE 7, using the typical SMRF design coefficients of R = 8, Ω = 3, Cd = 5.5. This analysis was used to design the lateral system of the new structure, but the model included the entire existing parking structure below to capture its impact on the global behavior and demands adequately. The second tier of analysis was a full-building nonlinear time-history response analysis per ASCE 41. This analysis was utilized to evaluate and retrofit the existing parking structure to the ASCE 41 performance objectives noted above. It also provided supplemental verification of the new building’s seismic design to these same objectives when explicit nonlinear hysteretic behavior was included in both the new and existing buildings.

The New Building The structural system for the new MAH building consists of concrete-filled metal deck over steel framing, with a seismic system comprised of Steel Special Moment Resisting Frames (SMRF) with Reduced Beam Section (RBS) connections. Moment Frame beams were typically W30x, with moment frame columns typically using heavy W24x or built-up box columns. All columns of the new building were spaced at a 27-foot x 30-foot typical grid to match the existing parking structure grid and land directly upon the existing concrete columns. In addition, to increase the floor area of the new structure beyond this grid, multiple sides of the floorplate provided perimeter cantilevers typically 15 feet long.

Figure 3. Deep built-up long-span transfer beam supporting steel moment frame column above.

Multiple long-span transfer beams were required within the new building to accommodate the large column-free zones for auditorium/ event spaces. This included the transfer of Special Moment Frame Columns, resulting in large, deep transfer beams requiring installation in multiple pieces and field-spliced once installed in place (Figure 3).

The largest of these transfer beams spanned 54 feet while supporting an SMRF column at its midpoint and was a custom 60-inch-deep shape that weighed over 800 pounds per linear foot.

The layout also required multiple bi-axial SMRF hollow box columns that received two or more beams in perpendicular directions. These box columns were fabricated using electroslag procedures to accommodate the typical continuity plates required for RBS connections inside the boxes.

Retrofit of the Existing Parking Structure

The existing parking structure required a unique and extensive retrofit to upgrade to current UCSSP requirements while supporting the gravity and seismic demands imparted from the new building above. The existing structure consists of 9-inch-thick flat reinforced concrete slabs with sloping drop caps at circular columns, along with some interior and perimeter concrete shear walls.

Some of the shear walls had been seismically retrofitted over the years with Fiber Reinforced Polymer (FRP) wrap or shotcrete. However, these retrofits accounted only for the parking structure’s own seismic demand and were based on different performance objectives and a less stringent Risk Category. Additionally, walls were unevenly distributed throughout the parking structure due to past renovations.

With the new MAH building constructed above, the entire seismic base shear of the new building above would be transferred into the existing parking structure before reaching the foundation.

The existing concrete columns required strengthening to support the demands delivered from the columns of the new building above. Existing columns seeing Figure 4. Steel jacketing of the existing concrete column at compression-only loads from above were the parking structure level.

JANUARY 2022

21

Figure 5. Foundation strengthening in the form of combining multiple existing spread footings.

Figure 6. The new Marion Anderson Hall.

strengthened with either FRP or concrete jacketing. Columns seeing compression/tension loads from above were strengthened with steel jackets with customized top connection plates to receive the anchor bolts from the SMRF column above (Figure 4 , page 21).

The existing slabs received significant diaphragm strengthening, predominantly in the form of FRP on the top and/or bottom of the slab. FRP application was chosen as the typical diaphragm strengthening material to maintain head-height requirements of the parking stories, which were already short in height. This FRP strengthening was predominantly seismic and was utilized for both collectors and diaphragm shear strengthening. The slabs also received a significant connection at the new building’s second floor, essentially “stitching” the new and existing diaphragms together at this elevation. This was achieved by chipping out the concrete of the existing slab while leaving the existing rebar in place and then re-pouring a heavily reinforced connection zone with new rebar coupled onto the existing rebar.

The seismic demand was transferred from the parking structure diaphragm to multiple existing and new reinforced concrete shear walls. The new walls were constructed by chipping out a slot in the existing slab to place the new rebar and concrete while leaving the

existing slab rebar in place. During this process, the existing slab was supported with temporary shoring. Multiple existing concrete shear walls were also retrofitted with either supplemental FRP wrap or shotcrete.

The typical existing foundation consisted of square isolated spread concrete footings. The majority of footings below the new building and those below new/retrofitted shear walls required strengthening.

The foundation strengthening consisted of increasing individual footing areas and/or increasing footing depth by adding supplemental anchors, rebar, and concrete to the perimeter and/or top of footings.

Where demands were largest, or new concrete shear walls were introduced, multiple footings were connected together to create new large combined footings (Figure 5).

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NEW VERSION

INTRODUCING

Conclusion The new Marion Anderson Hall (Figure 6) provided a sizeable expansion to the Anderson School of Management. The location and exterior façade provide a continuous extension of the original buildings, resulting in the perfect blend of the desired function with a matching aesthetic. The challenges of constructing an entirely new building on an existing parking structure were numerous – but close coordination between the design team, contractor, and client resulted in a great success.■

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22 STRUCTURE magazine

Daniel Tunick is a Senior Project Engineer at Nabih Youssef Structural Engineers (dtunick@nyase.com).

Nabih Youssef is the Founder and CEO of Nabih Youssef Structural Engineers (nabih@nyase.com).

Project Team sponsored by

Based on AISC Steel Design Guide 11 and Steel Joist Institute (SJI) Technical Digest 5

Owner: University of California Structural Engineer of Record: Nabih Youssef Structural Engineers Executive Architect: Gensler Design Architect: Pei Cobb Freed & Partners General Contractor: PCL

T

he 100 Stockton Street project reimagines an eight-story former department store into a multi-use office, dining, event space, and boutique retail building. This reimagination of the building required significant structural shoring to facilitate the design.

The building in San Francisco’s historic Union Square is a 1970s concrete building consisting of 250 thousand square feet. Modifications to the building required the demolition of the roof level, demolition of over one-third of the building floor plate, removal of half of the gravity columns, shortening of the existing post-tensioned (PT) girders, and demolition of the suspended first floor and the existing perimeter shear walls (Figure 1).

These elements were reconstructed in new locations, configurations, or to new extents to Figure 2. Completed shoring columns and beams prior to column demolition. accommodate the design. Degenkolb Engineers designed the extensive shoring required to meet the needs of the and Blatteis & Schnur. The shoring was closely coordinated with the project and the vision of the architect Gensler, building Structural design team as well as the general contractor Plant Construction, the Engineer of Record (SEoR) KPFF, and developers Morgan Stanley demolition subcontractor Silverado, the shoring steel subcontractor Olson Steel, and the lifting contractor Sheedy Drayage.

SHORING Facilitating DESIGN at 100 STOCKTON

Demolition

By Robert Graff, S.E.

The roof level and all penthouses were demolished. A new roof was constructed using steel and concrete on metal deck. This rebuilt roof level allowed for a perimeter outdoor terrace for a restaurant and bar overlooking Union Square. Minimal shoring was needed for this work, but the structure was evaluated for its ability to support the necessary demolition equipment, including excavators and skid steers.

The exterior shear walls and the perimeter of the original floor plate were demolished. The floor plate was reconstructed, cantilevering out to new extents at all levels. The newly defined edge of slab accommodated a façade consisting of glass and terracotta, a significant change from the original nearly windowless exterior. Within the remaining seven stories of floor plate, there were twenty existing 24-inch square columns, of which ten columns and their foundations were demolished.

The demolished columns each carried 7 stories of floor plate consisting of PT girder, PT joists, and a 4½-inch reinforced slab totaling over 100 psf. The columns carried 800 kips each, and the combined shoring load for the ten columns approached eight million pounds.

Shoring Requirements

Figure 1. Demo plan – The red shaded area is demolished. Blue columns were demolished; other black columns remain.

Shores to support the loads from the demolished columns were designed for a maximum of 500 kips at the base of the building using two shoring posts for each building column. Fabricated structural steel was selected for the shoring at most continued on next page JANUARY 2022

23

levels due to the size of the loads. The shores were designed as 18-inch pipes at the basement level but tapered in size up the building height. At the upper two levels, adjustable steel shores were used where loads were low enough that off-the-shelf shoring systems were adequate.

The PT girders on the four main column lines at all levels were de-tensioned. This allowed the girders to be shortened by a bay at each end and was accomplished by releasing the tendon stress, chipping back the girder, cutting the existing tendons, casting new PT tendon anchorages, and finally re-tensioning the original tendons. This all was completed to allow the perimeter of the building to be reconstructed to new extents for the façade.

In their temporary de-stressed state, the PT girders could not support the joists and slab with only the mild steel reinforcement they possessed. Shoring beams placed on top of the column shores provided nearly continuous support during the girder’s Figure 3. Setting up for test lift. extremely weakened state. Shoring beam deflections were calculated while supporting the load of the slab, joists, and deSystem Deflections stressed girders. Due to a temporary condition, the deflections would become locked in when the new columns and walls supporting the slab With all the shoring elements determined, expected deflections of the were cast. Detailed checks of the steel shoring beams were completed to system were calculated to be between ¾ inch to 1 inch. This resulted control deflections within acceptable limits resulting in stiff W27×146 from a combination of column shortening, pile settlement, and beams (Figure 2, page 23). beam deflections. If allowed to occur, this deflection would become At the base of the shores, the total loads were beyond what could be permanent in the final building when columns and walls were cast. supported by cribbing which would be a typical temporary shoring Due to the capacity of the existing floor framing, topping the slab to foundation solution. The project also required the demolition of the correct such deflections was not possible. To compensate, a jacking existing foundations and excavations for new foundations. This all operation to transfer the building load was developed. Most of these occurred around the shoring system while it was supporting the build- deflections were eliminated or significantly reduced by transferring the ing. To support the large loads and load to the shoring prior to column allow for the necessary excavation, demolition.

12-inch cased micropiles were used.

The jacking operation required The micropiles could be installed in the use of hydraulic jacks at each the basement and could support the shore to transfer the loads. However, building load, while the top eleven placing upward loads on PT girders feet of each pile were exposed due to which were stressed at this phase is the foundation excavation. In addidangerous. The upward force comtion, the micropiles had a limited bined with the negative moment impact on the permanent foundainduced by the PT stress can cause a tions cast around them, making negative bending failure. Therefore, them an ideal solution. the girders were evaluated under the A steel frame or carriage (as the shoring loads and found to approach contractor named it) was designed failure in the rebar on the top side of to transfer loads from the shores to the girders. The evaluation included the micropiles. The carriage conseveral conservative assumptions sisted of 1½-inch-thick triangular about the stresses remaining in the gusset plates that were slotted into tendons after 50 plus years. The the shoring column. These gusset original tendon stresses were known, plates delivered the load of one shore but initial stress losses and losses down to a rectangular frame of wide due to long-term creep had to be flange beams. The frame, in turn, conservatively estimated. However, delivered the load to two piles for the possibility of inducing failure of each shoring column. The carriage the girders when jacking the buildserved a second purpose which was ing could not be easily disproven. to link four piles together to provide A section of the building scheduled additional stability. This stability for demolition was used to test the was exceedingly important when proposed procedure to prove the excavating around the piles for the jacking could be completed successFigure 4. Column demolition with bars buckling. foundations. fully. The shoring was designed for 24 STRUCTURE magazine

the new footings exposed around 11 feet of the previously buried cased micropiles (Figure 5). The casing provided buckling resistance to the micropile, which the surrounding soil would typically provide.

The shoring carried all 8 million pounds of load at this stage, and the shoring system was in its most vulnerable state.

One Last Challenge

Figure 5. Steel foundation forms installed and building supported on exposed piles.

1000 kips maximum per column and had four jacks to lift the load (Figure 3). The jacks were incrementally increased in load, and the building was inspected for signs of damage at each increment. A surveyor monitored the structure for movement, providing real-time feedback. Arriving at nearly 800 kips, the surveyor recorded that the building column had moved 1⁄16 of an inch upward, and the shoring system had deflected 5⁄8 inch at the shoring columns. The shores were shimmed using stacks of steel plates between the shoring column and beam to lock in the load before releasing the jacks. The building had no damage, and the full load was now supported on the shores. With the successful test, the remaining bays could be jacked to transfer the building loads from the columns to the shores, and column demo proceeded.

Column Demolition Column demo started with a text of the picture shown in Figure 4 and a concerned call from the demo contractor.

He asked if it was ok that the column bars were buckling out about 2 inches as they demolished the concrete column.

Quickly back-calculating, it was determined that around ⅛ inch of vertical deflection could cause a 2-inch buckle in a bar over the story height. A ⅛-inch vertical deflection was undoubtedly well within expectations of building movement when a column is removed. Fears of collapse were quelled, and demolition proceeded.

With columns demolished, they removed the old building foundations.

As they excavated, they also removed the abandoned brick and concrete foundations of previously demolished buildings on this site that predated the current building. Excavating for

A couple of months later, foundations were poured around the piles, and stability started to be restored.

However, the project had one last significant shoring challenge. The exterior sidewalk elevation varies by 9 feet around the perimeter of the building. The original building, designed as a single department store, had two main entrances with steps to accommodate the change in grade. The renovated building was designed for individual boutique retail. The suspended first floor over the basement was demolished and reconstructed with a stepped floor plate to allow for level entrances to each business.

The column shoring system was still supporting the building, and the shoring stopped and started under and over the first-floor girders. The concrete girders temporally remained, but the slab and joist of the first floor were demolished. By demolishing the first floor, the shores would buckle without the bracing provided by the floor. A series of steel pipe kickers were anchored to the new building foundations and up to the shoring columns to provide bracing (Figure 6 ). The American Institute of Steel Construction (AISC) provides requirements for bracing. The bracing loads are relatively small, but the stiffness of the bracing is equally important and is what drove the design.

With the bracing in place, the firstfloor demo proceeded.

Once the first floor was removed, the building as designed by the SEoR could start to be constructed up and out of the basement. With every few weeks that passed, another floor was re-supported by the new columns and walls.

Eventually, the new structure topped out. The shoring had done its job, and it was time for it to go. The demo sub returned to the job and removed the shoring, sending it off for recycling.

Shoring is often necessary to facilitate structural and architectural designs – especially those that reimagine existing buildings. Facilitating a design can be as simple as a few temporary wood shores or as complex as this project, which pushed the limits of what can be done with building shoring.■

Figure 6. Shoring braced with the first floor demolished.

Robert Graff is a Principal at Degenkolb Engineers and is active in their Construction Engineering and Education practice groups (rgraff@degenkolb.com).

JANUARY 2022

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Historic Alameda High School Retrofit Part 1: Too Valuable to Demolish, Too Expensive to Retrofit

By Nik Blanchette, P.E., Steve Heyne, S.E., and Chris Warner, S.E.

View of buildings looking down Central Avenue.

I

n 2012, the Alameda Unified School District in Alameda, California, made the difficult decision to fence off and vacate all three historic classroom buildings on the Historic Alameda High School (HAHS) campus due to seismic safety deficiencies. These buildings and the attached auditorium, lobby, gym, and locker room buildings had

stood since 1924. However, the classroom buildings lacked approval under California’s Field & Garrison Acts, putting the school district at legal risk (and any building occupants at life safety risk). The long process to rehabilitate and restore these nearly century-old buildings had entered its final chapter.

The Structures

A school destroyed by the 1933 Long Beach earthquake.

26 STRUCTURE magazine

The original campus, standing three stories tall, has an impressive presence spanning an entire block of Central Avenue adjacent to the downtown district of Alameda. The buildings are of neoclassical style with grand concrete entry columns, emulating the stone columns of ancient Rome, and elaborate detail work throughout the exterior.

The buildings consist of cast-in-place reinforced concrete exterior walls supported by shallow foundations. All floors and roofs are wood-framed, except the second-story corridor floors, which were concrete (and removed during the retrofit). Reviewing the buildings against either ASCE 7, Minimum Design Loads and Associated Criteria for Buildings and Other Structures, or ASCE 41, Seismic Evaluation and Retrofit of Existing Buildings, showed that the major elements of the lateral force-resisting system were significantly deficient, lacking strength, stiffness, and interconnection.

Key plan of building retrofit history.

The Field Act

and highly fenestrated exterior concrete walls. Horizontal steel truss diaphragms were specified to augment the straight sheathed wood roof diaphragms. Concrete wall-to-diaphragm anchorage was to be improved. Unfortunately, only the locker rooms, gym, and auditorium improvements were upgraded, presumably due to limited funding.

The classroom wings, auditorium lobby, and science building were not upgraded.

For decades following, students and faculty still occupied all buildings. Based on a study of available documents from the school district, this issue had sporadically been discussed among hired structural engineers and the Division of the State Architect. Finally, in 1978, students were moved to a replacement campus down the street, leaving behind the large, underutilized, historic, seismically deficient buildings. During this time, one of the wings was fully

A Retrofit Unrealized The 1933 Long Beach earthquake was a “wake-up call,” resulting in the preparation of seismic rehabilitation drawings for the HAHS campus. Reinforced concrete shear walls were specified to supplement the overstressed diaphragms

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In the early evening of March 10, 1933, less than 10 years after the construction of the HAHS campus, a magnitude 6.4 earthquake jolted Long Beach, California. Widespread damage occurred, primarily to unreinforced masonry structures. Among the wreckage were 70 destroyed schools and 120 damaged schools, totaling about 75% of the schools in the Long Beach area. Thankfully, few students and staff were present, though fatalities were not avoided.

The 1933 Long Beach earthquake resulted in creation of the Field Act for new public school construction in California, one of the early pieces of legislature incorporating seismic standards in building design. The Act prohibited unreinforced masonry construction and required consideration of a seismic design force. The Act also created the Division of the State Architect (DSA) to oversee the design and construction of public schools. DSA reviews several billions of dollars of construction every Structural design prowess year for K-12 schools and community colleges throughout California. Since meets architectural vision. the creation of the Field Act, no public school building has collapsed, nor has loss of life occurred in a public school building due to an earthquake. The Field Act was followed by the 1939 Garrison Act, which provided criteria for analysis and rehabilitation of existing public school buildings constructed prior to the Field Act.

Seattle | Tacoma | Portland | pcs-structural.com

JANUARY 2022

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vacated and remained so until the retrofit was completed in the 2010s. The other classroom wing and science building served various uses, such as adult school classrooms, a public library, and district office space.

In 1977, the community failed to pass bonds to rehabilitate the remaining non-upgraded historic buildings, so the school board voted to demolish the 50-year-old deficient buildings. However, a group of community leaders rejected the idea of losing the historic property and rallied to save the buildings. Also in 1977, perhaps not so coincidentally, the campus was recognized as a national historic monument and was placed on the National Register of Historic Places.

Finally, in 1989, the group was able to pass a bond that paid for rehabilitating the auditorium lobby and seismically isolating it from the classroom wings. This allowed students to use the lobby and auditorium, sandwiched between the deficient classroom wings.

Structural debris fence surrounding an abandoned building.

A Path to Rehabilitation In 2012, at the school district’s request, ZFA Structural Engineers completed a districtwide review of DSA project certifications. During this review, it resurfaced that prior rehabilitation work to make the subject structures compliant with the Field Act was never performed.

Based on this information and consultations with the DSA, the threestory science building and the two-story classroom wing that had been in use were abandoned entirely. A structural fence was erected around the buildings to prevent potential falling debris from harming the public. The 1977-78 scenario essentially replayed itself: the buildings presented a financial burden and liability to the school district, while the community, including the local historical society, saw the value of the buildings and wanted them preserved.

California voters approved Proposition 1D in 2006, which provided $199.5 million for critically seismic deficient public school buildings across the state. Under the Seismic Mitigation Program (SMP), eligible buildings receive matching funds (i.e., 50% cost reimbursement) from the state for seismic rehabilitation costs.

The DSA created Procedure 08-03 to outline the SMP process to stakeholders. The first phase is the Eligibility Evaluation Report (EER) that quickly screens and confirms the eligibility of buildings for SMP funding. Next, if the rehabilitation cost is shown to be less than half of the building replacement value, the building is eligible for rehabilitation funding; otherwise, if the cost is greater, it is eligible for replacement funding. The third phase for seismic rehabilitation is the Evaluation and Design Criteria Report (EDCR). The report characterizes the building, describes the structural analysis procedure, and outlines data collection requirements. The fourth phase is the creation of rehabilitation construction documents subject to review and approval by the DSA. The fifth and final phase is funding from the State, which occurs after the project is constructed in accordance with the DSA-approved plans.

Before involvement with the HAHS project, ZFA Structural Engineers was one of two firms hired by the State to help create the EER template as a modified version of ASCE 31, Seismic Evaluation of Existing Buildings, (now Tier 1 & 2 of ASCE 41) to identify which critical seismic deficiencies are most likely to trigger the collapse of the building. The EER is a uniform, straightforward approach to screen vulnerable buildings. However, some building types are excluded from eligibility due to inherent redundancy and documented performance in earthquakes, such as buildings primarily framed with wood. An example of a critical deficiency is wall 28 STRUCTURE magazine

anchorage; if the wall anchorage is deficient, the diaphragm could separate from the wall, and the building could collapse.

While the EER is a quick look at some aspects of the building, the EDCR represents a more significant review of the existing construction. There are numerous sources of uncertainty in the analysis of existing buildings. Construction standards in structures built long ago, including quality assurance and quality control, were typically less rigorous than current industry practice. The EDCR is the time for the DSA and the design professional to agree on an approach to analyzing the existing building. The report describes the existing construction, potential deficiencies, a methodology for calculations, and data collection.

Data collection is expected to substantiate the material properties of the existing building to be used in rehabilitation design. The main form of data collection is material testing in accordance with ASCE 41.

For each building material, ASCE 41 specifies the type and quantity of testing to perform, such as concrete cores, steel coupons, or visually grading lumber. DSA reviews and approves the EDCR before the submittal of the construction documents for the rehabilitation project. Ideally, the EDCR process leads to a smoother review of the rehabilitation drawings by the DSA. The SMP rehabilitation must follow an ASCE 41 Tier 3, Systematic Evaluation and Retrofit approach. Every component resisting seismic forces must be analyzed and shown to comply with current code requirements as if it were a new building. The retrofit is not limited to the deficiencies identified in the EER. DSA specifies the performance objective: a seismic hazard (i.e., demand) and a performance level (i.e., capacity). The performance objective for a rehabilitation of this nature is similar to what DSA specifies for new construction.

Accessibility and fire life safety aspects of the building must also be made to comply with current regulations.

The SMP, along with the community discussions, convinced the school board that the retrofit of the buildings was viable, and they went forward to place and pass a bond on the ballot in 2014. This project was the largest funded in the bond.

A follow-up article will detail the technical challenges and achievements of retrofitting a nearly 100-year-old structure under DSA jurisdiction while maintaining the historic significance.■ All authors are with ZFA Structural Engineers in Santa Rosa, CA. Nik Blanchette is an Engineer (nikb@zfa.com). Steve Heyne is an Associate (steveh@zfa.com). Chris Warner is a Principal (chrisw@zfa.com).

Boston University Center for Computing & Data Sciences Building

By Nathan Roy, P.E., Irfan Baig, P.E., Jamie Hamelin, P.Eng, and Lucy Timbers

Figure 1. Building rendering.

L

ocated on Commonwealth Avenue, the Center for Computing & Data Sciences rises dramatically above the central campus of Boston University. Consisting of a 19-story, 305-foot-tall tower, and 5-story podium, the building is a hub for the campus and a showcase for the departments of Mathematics & Statistics, Computer Science, the interdisciplinary faculty of Computing & Data Sciences, and the Rafik B Hariri Institute for Computing and Computational Science & Engineering. Designed by internationally renowned design firm

a)

b)

Figure 2. a) Building section; b) render of structural systems.

30 STRUCTURE magazine

KPMB Architects of Toronto, Canada, the Center for Computing & Data Sciences capitalizes on its location in the heart of campus to create an inviting meeting place. The design of the building encourages collaboration and innovation between disciplines by creating vertically stacked research “neighborhoods” with staggered green-roofed terraces, interconnecting feature stairs, and generous public spaces. The podium houses student-focused facilities and amenities, including much-needed study spaces. The dynamic ground floor level enlivens and extends the streetscape.

The key sustainability and resilience goals were fostered throughout BU’s design, resulting in one of the first large fossil-fuel-free buildings in Boston, aligning with the campus climate action plan with a target of LEED Platinum. The transparency and porosity of the building’s envelope displays the elegance and complexity of the cantilevered steel structure and acts as a beacon on the Charles River skyline (Figure 1). Teaming on the design to bring the building structure to reality are structural engineering firms Entuitive of Toronto, Canada, and LeMessurier of Boston, Massachusetts.

Common to Boston, the site is on reclaimed land from the second half of the 19th century. Originally part of the Back Bay, the site is underlaid with 6 to 15 feet of miscellaneous fills located over a 5- to 10-foot layer of organic deposits. Below the organic layer is a 7- to 16-foot glacier deposited

sand layer that is a common layer for supporting the timber piles of historic Boston buildings of the late 19th and early 20th century. Below the sand layer is 150 to 165 feet of marine deposited clay with bedrock below.

Thus, the site represents one of the deepest locations in Boston to bedrock.

The building consists of two main portions: the 19-story tower at the west side of the site with overall plan dimensions of 140 by 140 feet with a two-story deep basement, and a 5-story podium at the east side with plan dimensions of 70 by 170 feet with a one-story basement. Haley and Aldrich conducted the geotechnical exploratory program and worked with the design team and Suffolk construction to select the appropriate foundation system. The analysis focused on two primary foundation schemes: deep load-bearing slurry wall elements (LBE) extending to bedrock and a mat slab foundation. Working with the construction manager Suffolk, it was determined that a mat foundation would save over $5 million Figure 3. Tower mat slab construction. compared to the LBE foundations.

The 19-story tower height was close to the limit acceptable for a mat north-south. The total core height is 338 feet above the top of the foundation. Tower loads had to be limited by optimizing the concrete mat foundation, with a height-to-width ratio of approximately core plan size and wall thickness and using lightweight concrete slabs 11 to 1. As noted, the soil conditions required the superstructure on metal deck to realize the savings with a mat foundation. A 5-foot to be as light as possible to limit short and long-term settlements mat slab bearing 40 feet below grade on the marine clay, and thicken- and keep the subgrade soil stresses below the allowable values. ing to 6 to 9 feet under the core walls (Figure 3), was utilized at the Therefore, the core wall was limited to 14 inches thick to help tower. The building weight was reduced achieve the weight reductions required to limit bearing pressures under dead and to meet the soil pressure and settlement live load to a maximum of 6 ksf under limits. High strength, self-consolidating the core with an average of 4.5 ksf under concrete with a strength fć = 10,000 psi the tower footprint. A 3.5-foot mat slab at the base transitioning to fć = 8,000 psi bearing on the sand layer was provided at the top was used for the core. below the podium. Approximately 1 to 1½ Contrary to conventional construction inches of elastic settlement and an addiof cast-in-place framed concrete slabs tional ½ to 1 inch of long-term settlement within the core, 3¼-inch lightweight is predicted at the tower mat slab, while ¼ concrete slabs supported by a 3-inch to ½ inch of total elastic settlement and deep composite metal deck are employed up to an additional ½ inch of long-term to reduce the total weight. The 5-story settlement is calculated at the podium. podium has its own lateral force resisTo address resiliency and the Boston tance system comprised of a structural University Climate Action Plan, the steel elevator and stair “core” made up ground floor was set 1 foot above of concentric braced frames. the project design flood elevation, The lateral loads on the building are 5 feet above the Boston Planning & wind controlled, with Exposure C. The Development Agency design flood 14-inch shear wall thicknesses necessielevation. Building weight under the tated high-strength threaded #14 Grade 19-story tower was sufficient to resist 105 steel reinforcing at boundary elehydrostatic pressures from the design ments. The large diameter reinforcement flood elevation. 40-ton tension minihelped eliminate rebar congestion in the piles are provided below the 5-story core. In addition, staggered mechanical podium to resist hydrostatic pressures splices were used for the boundary elefrom the design flood elevation. ment steel along the height of the core Responding to the building programwalls, further helping reduce congestion ming requirements and compact floor and conflicts with various other building plate, the lateral forces resistance system components, including embed plates. in the tower consists of a slender conThe three slender north-south walls crete shear wall core (Figure 4 ). The contained door and MEP penetrations, core footprint is 52 by 30.5 feet with creating a challenging scenario for the two 52-foot-long walls in the east-west design of these wall and link beams. direction and three 30.5-foot-long walls Figure 4. Tower core.

Typical link beams occur in the center JANUARY 2022

31

a)

b)

c)

Figure 5. Cantilever framing; a)16 t h floor framing with overlay, b) staggered massing, c) steel cantilever framing.

of the wall and contained MEP penetrations to allow services out of the core. This placed significant constraints on the placement of rebar within the link beams. In addition, the narrow thickness of the wall and link beams precluded embedded structural steel sections in the link beams at the lower levels to help carry the shear/ flexure demands. Instead, 1¼- to 1½-inch-thick grade 50 steel plates are embedded in the link beams at lower levels to provide adequate strength and stiffness to the link beam sections. The plates are made to act compositely with the concrete with headed studs on each side of the plate. Careful coordination with architectural, mechanical, and electrical teams was required to place the penetrations through these plates. The effort and care of planning these link beams paid off during the construction phase, with few conflicts resulting from rebar placement and/or wall penetrations.

Figure 6. Truss framing at floor level.

32 STRUCTURE magazine

The tower has typical 115- by 115-foot floor plates of five 23-foot bay modules in the north-south and east-west directions. The floor plans have footprints that line up with each other vertically for no more than three floors consecutively. The center four-bay by four-bay portions of the tower supports all tower gravity loads and includes a concrete core and columns that run continuous the full height of the tower (Figure 5a).

The architectural massing of the building uses shifting, free-floating volumes to create outdoor terraces associated with research “neighborhoods” that capitalize on the spectacular views from all sides of the tower. The overall building footprint of 138- by 138-foot comprises floor plates made up of six 23-foot bay modules in the north-south and east-west directions (Figure 5a). Floor plates shift by one bay in a counter-clockwise arrangement around the core every two or three stories. This creates an offset block layout of masses, with different volumes cantilevering over the floors below (Figure 5b). Columns in this area do not extend to grade. Two-story deep trusses made of wide flange steel and located along the perimeter support these volumes. Typically, a single truss spans the full length of the building, which is, in turn, supported by a truss in the perpendicular direction that cantilevers a single bay (Figure 5c). This results in a mixture of traditionally supported and hung floors.

This careful placement of the trusses creates a load path that guides the gravity loads back towards the columns that run continuous through the height of the building. The architecturally exposed steel truss framing and connections are expressed visually and are fireproofed with intumescent paint (Figure 6 ).

The cantilever steel framing was superelevated for 80% of the predicted dead load deflections (Figure 7 ).

Suffolk Construction, Prime Steel Erecting, and their erection engineer, Simon Design Engineering, worked closely with the design team on the temporary shoring and jacking systems. Full height shoring was provided to allow for erection of the

steel. Jacking boxes as part of the shoring system were included to allow for superelevation. Hydraulic jacks were utilized to unload the shoring and uniformly load the cantilever framing.

Opening in late 2022, the Boston University Center for Computing & Data Sciences building will foster innovation and collaboration as a leader in Computing & Data Sciences. The building is set to demonstrate Boston University’s commitment to sustainability, resiliency, and social responsibility.

Part 2 will focus on sustainability, life cycle assessment, and opportunities realized to reduce the embodied carbon of the building structure.■ Nathan Roy is a Principal with LeMessurier (nroy@lemessurier.com).

Irfan Baig is a Principal with LeMessurier (ibaig@lemessurier.com).

Jamie Hamelin is a Senior Associate with Entuitive and is based in its Toronto, Canada office (jamie.hamelin@entuitive.com).

Lucy Timbers is a Senior Associate with KPMB Architects of Toronto, Canada (ltimbers@kpmbarchitects.com).

Project Team Structural Engineers: LeMessurier, Entuitive Architect: KPMB Architects Contractor: Suffolk Construction Steel Fabricator: Canatal Industries Steel Erector: Prime Steel Erecting Inc Concrete Contractor: S&F Concrete Contractors

Figure 7. Steel framing. Courtesy of John Cannon.

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just the FAQs FAQs on ASCE Standards What You Always Wanted to Ask

By Laura Champion, P.E., F.SEI, F.ASCE, and Jennifer Goupil, P.E., F.SEI, M.ASCE

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elcome to this new quarterly column for STRUCTURE magazine. These articles will address some of the questions received (along with responses) about structural standards developed by the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE), such as ASCE 7 and ASCE 41. Questions received from engineers, building officials, and other design professionals are often considered for the development of future editions.

Following are some questions received by SEI as well as responses to clarify the provisions.

ASCE/SEI 7: Minimum Design Loads and Associated Criteria for Buildings and Other Structures Seismic Design for Tanks and Vessels: Chapter 13 vs. Chapter 15 Q: Would you please clarify when to use Chapter 13 Nonstructural Components versus Chapter 15 Nonbuilding Structures? For example, which chapter requirements govern the design of the anchorage for a small tank to a concrete roof structure: Chapter 13 (and to apply Table 13.6-1, Seismic Coefficients for Mechanical and Electrical Components), or Chapter 15 (specifically Section 15.7.5, Anchorage)?

A: There is an overlap between ASCE 7-16 Chapters 13 and 15. Tanks are found in Chapter 13, Table 13.6-1, and Chapter 15, Section 15.7, Tanks and Vessels. Philosophically, Chapter 13 covers relatively small components supported above grade in a building and Chapter 15 covers nonbuilding structures (large components) supported at grade. However, there is an overlap between the two chapters. Chapter 13 provisions can be applied to components supported at grade, and Chapter 15 Section 15.3 covers nonbuilding structures supported by other structures. The Section 15.3 rules apply to relatively large nonbuilding structures.

The intent of Section 15.7.5, in conjunction with 15.7.3, Strength and Ductility, is to make the anchor the seismic fuse on tanks and vessels. Forcing the anchor to yield and stretch has proven to be the most effective way to provide ductility for a tank or vessel and minimize or eliminate damage during a seismic event. These requirements work well for moderate to large diameter tanks and vessels but do not work for relatively small tanks and vessels. The practical anchor size used to anchor small tanks and vessels precludes the anchor from stretching and yielding. This is addressed in ASCE 7-22 for small tanks supported at grade; ASCE 7-22 Chapter 15 permits the design of 34 STRUCTURE magazine

the anchorage using any of the options in the American Concrete Institute’s ACI 318-19, Building Code Requirements for Structural Concrete, Chapter 17 Anchoring to Concrete. In this case, small is defined as a tank or vessel less than or equal to 5 feet in diameter and less than or equal to 10 feet in height. The definition of a small tank and vessel was derived by determining the diameter that would limit the convective mass to 10 percent of the total liquid mass while holding the height to approximately one story.

For very small tanks supported on the roof of a building, Chapter 13 requirements are likely more applicable than the requirements of Chapter 15. This is because, as described above, it is unlikely that the anchorage provided for a small tank would yield under seismic loads, so the objectives behind the Section 15.7.5 provisions cannot be achieved.

Can a Site Class Change from F to D on a Liquefiable Site?

Q: Does the exception in ASCE 7-16 Section 20.3.1, Site Class F, allow the Site Class of a liquefiable site to change from an F to a D for buildings with fundamental periods equal to or less than 0.5 seconds?

A: Yes, the exception to Condition 1 of Section 20.3.1 does allow the site class to change from Site Class F to Site Class D if the fundamental period of the structure is less than or equal to 0.5 seconds, but ONLY if the soil 1) does not meet any of the other conditions listed under Section 20.3.1 for Site Class F, 2) does not meet the requirements for Site Class E under the exceptions to Section 20.3.1 Conditions 3 and 4, 3) is not classified as Site Class E under the requirements of Section 20.3.2, Soft Clay Site Class E, and 4) is not classified as Site Class E under the requirements of Section 20.3.3, Site Classes C, D, and E. The exception to Condition 1 of Section 20.3.1 does not take precedence over the other requirements of Sections 20.3.1, 20.3.2, and 20.3.3.

Q: As a clarification, is it the intent of this standard that the exception in Section 12.13.9, Requirements for Foundations on Liquifiable Sites, allows foundation ties to be omitted when all the parameters are met, even if the Site Class would have been an F? A: No. The exception under Section 12.13.9 does not allow foundation ties required by Section 12.13.8.2, Foundation Ties, to be omitted for spread footings founded on Site Class E and F soils. The exception under Section 12.13.9 simply avoids the ADDITIONAL requirements for ties in 12.13.9.2.1.1, Shallow Foundation Design, Foundation Ties, from being applied. Please note that the commentary Section C12.13.9.2 clarifies this requirement; “Shallow foundations are required

to be interconnected by ties, regardless of the effects of liquefaction.”

When are “Openings” Open?

Q: Can a building with large overhead doors on one side be designed as enclosed instead of partially enclosed. Their assertion is that the overhead doors are not openings because they are designed to be closed during a design wind event.

A: In ASCE 7-10, Chapter 26, Wind Loads: General Requirements, defines enclosed, open, and partially enclosed. Overhead doors can provide the degree of enclosure required to meet the definition of an enclosed building, provided that these doors are designed for the design wind pressures without excessive deflection. However, there are exceptions to this general situation. For example, doorways must be considered openings for a fire station because of the requirement that the doors be opened during the wind event to respond to emergencies. The same situation would be for ambulance garages or emergency room entrances. The open area around the doors should be considered when determining the enclosure classification of the building. In ASCE 7-16, the definition of enclosed was clarified by specifying the total area of the openings, Ao, permitted, and the definition of partially open was also added.

Wind Loads on Solar Arrays Q: In ASCE 7-16 Section 29.4.4, Rooftop Solar Panels Parallel to the Roof Surface on Buildings of All Heights and Roof Slopes, there are two factors, γE and γa, that are confusing. The Array Edge Factor, γE, definition describes the location of the array on the roof and in relation to other arrays. When does the γE = 1.5 factor not apply?

Also, Figure 29.4-8 includes the Solar Panel Pressure Equalization factor, γa, which is based on the effective wind area. Is this Effective Wind Area based on each connection, as determined in Chapter 30, Wind Loads: Components and Cladding, or does it refer to the total area of the solar array?

A: The Array Edge Factor, γE, is used to determine how “exposed” the panel is and, thus, how susceptible the panel is to wind uplift.

Therefore, if the panel is greater than 0.5 × the mean roof height, h, away from the edge of the roof, but the distance to the building edge (or to adjacent array), d1, is greater than 4 feet or the distance between the rows, d2, is greater than 4 feet, then the γE = 1.5 factor applies. However, if the distances are not greater than 4 feet, a value of γE = 1.0 can be used instead of γE = 1.5.

To determine the Effective Wind Area, A, refer to the definition contained in Chapter 26, Section 26.2: “For rooftop solar arrays, the effective wind area in Fig. 29.4-7 is equal to the tributary area for the structural element being considered, except that the width of the effective wind area need not be Less Than One-Third Its Length.”

Wind Tributary Area for Components and Cladding Q: Regarding components and cladding uplift, Section 30.2.3 Tributary Areas Greater than 700 ft 2(65 m 2), of ASCE 7-16 states, “C&C elements with tributary areas greater than 700 ft2 (65 m2) shall be permitted to be designed using the provisions for main wind force resisting systems (MWFRS).” When using “the provisions for MWFRS,” does this also mean that the member should be designed for the interaction of two directions of wind load? When designing

strictly for C&C, the load being designed for is not directional, so it is unclear if this comes into effect for “the provisions for MWFRS” design.

A: Section 30.2.3 describes the situation where the component or cladding element has a large tributary area instead of the small, typical effective wind area. For an element with such a large tributary area, the high localized wind pressures associated with the loading of a component and cladding element are not present. Thus, the smaller design pressures used for the design of the MWFRS may be used in the design of this component or cladding element. Further, C&C loading may be bi-directional; consider a corner window system, for example. Wind pressures in both directions should be applied to the corner window simultaneously.

ASCE 41: Seismic Evaluation and Retrofit of Existing Buildings Clarification for Tier 1 Q: Neither ASCE 41-13 nor ASCE 41-17 has a Tier 1 Immediate Occupancy checklist or evaluation requirements for Building Type C1: Concrete Moment Frames for High Seismicity Level. I am referring to the following sections in ASCE 41-13 and ASCE 41-17: • ASCE 41-13: Section 16.9IO Immediate Occupancy Structural Checklist For Building Type C1: Concrete Moment Frames • ASCE 41-17: section 17.11 Structural Checklist For Building Type C1: Concrete Moment Frames Why does ASCE 41 not specify an Immediate Occupancy checklist for Concrete Moment Frames for High Seismicity Level?

A: There is no separate Tier I checklist for Building Type C1 for High Seismicity Level because the checklist for moderate level is also applicable for high level.■ If you have a question to be considered for a future issue, send it to sei@asce.org with FAQ in the subject line.

Visit asce.org/sei to learn more about ASCE/SEI Standards.

This article’s information is provided for general informational purposes only and is not intended in any fashion to be a substitute for professional consultation. Information provided does not constitute a formal interpretation of the standard. Under no circumstances does ASCE/SEI, its affiliates, officers, directors, employees, or volunteers warrant the completeness, accuracy, or relevancy of any information or advice provided herein, or its usefulness for any particular purpose.

ASCE/SEI, its affiliates, officers, directors, employees, and volunteers expressly disclaim any and all responsibility for any liability, loss, or damage that you may cause or incur in reliance on any information or advice provided herein.

Laura Champion is a Managing Director of the Structural Engineering Institute and Global Partnerships at the American Society of Civil Engineers.

Jennifer Goupil is Senior Manager of Codes and Standards and Technical Activities at the Structural Engineering Institute at the American Society of Civil Engineers.

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structural LOADS Snow and Rain Loads in ASCE 7-22 Part 1

By Michael O’Rourke, Ph.D., P.E., and John F. Duntemann, P.E., S.E.

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he American Society of Civil Engineer’s ASCE 7-22 load standard, Minimum Design Loads for Buildings and Other Structures, is now available. Substantive changes have been made to the snow and rain provisions within the standard. In particular, the ground snow loads have been revised to reflect more recent snow load data and reliability-targeted values.

In addition, the method for estimating drifts has been revised to include a wind parameter, and the procedure for determining design rain loads has been revised to explicitly consider a ponding head. Some of the more substantive changes are discussed, along with the reasons for these changes. This article is Part 1 of a two-part series and reviews the new ground snow loads and a new winter wind parameter. Part 2 will include the other more substantive changes to the snow load provisions and the new rain load provisions.

A very small fraction of the locations defined in the Geodatabase indicate that a case study must be completed to determine the ground snow load. These case-study regions are now limited and apply only to locations higher than any locally available snow measurement locations. Database ground snow load values are still provided to the user, with a warning that the estimated value lies outside the range of elevations of surrounding measurement locations.

Information from local experts, from reports by Bean et al. (2021) or Buska et al. (2020), can be used to determine values at these locations.

ASCE 7-22 also includes GSL maps for each Risk Category. Each of these maps (and associated datasets) is based on reliability calculations that target the reliability objectives of Chapter 1 of ASCE 7-22. A copy of the GSL map for Risk Category II for the conterminous United States is reproduced in Figure 1. (Due to scale, Figure 1 is included in the online article at STRUCTUREmag.org.)

Ground Snow Loads The adoption of reliability-targeted design The previous editions of ASCE 7 included ground snow loads represents a significant mapped values for ground snow load (GSL) change from ASCE/SEI 7-16 and prior editions, based on a statistical analysis using National which previously used ground snow loads with Weather Service snowfall data from 1952 to a 50-year mean recurrence interval (MRI). Due 1992. This map was first included in the 1992 Figure 2. Box plot of the ratio of proposed to climatic differences, reliability-targeted loads edition of ASCE 7 and was updated with addi- factored loads to previous factored loads. are adopted to address the nonuniform reliabiltional information for the 1995 edition. It has Average ratio:1.12. Additional data is included ity of roofs designed according to the 50-year remained essentially as it was in 1995 for each in the online article at STRUCTUREmag.org. snow load in different parts of the country. For subsequent edition through 2016. Additionally, example, in some parts of the country, designat the time that map was generated, the authors (researchers at the ing for the 1.6 load factor times the 50-year value does not meet the Cold Regions Research and Engineering Laboratory [CRREL] of reliability targets of the standard (and, in some of these places, failures the US Army Corps of Engineers) marked as Case Study, or CS, due to an underestimated ground snow load have been observed). In several significant regions, encompassing large parts of eighteen states, other places, designing for the 1.6 load factor times the 50-year value where significant ground elevation related changes to GSLs resulted is unnecessarily conservative. in unreadable maps. The CS regions placed a significant burden on Figure 2 presents a box plot of the ratio of the new factored flat roof structural engineers to perform snow load hazard analysis, and very load to the factored ASCE 7-16 uniform loads for 65 locations in the little guidance had been provided on how to conduct such studies.

United States. This plot indicates that while some locations changed The new GSL in ASCE 7-22 is an updated national GSL dataset in drastically, the majority of structures have a roof load ratio (new/ electronic and map form. The new snow loads are also based on nearly current) of 0.91 to 1.30, with an average of 1.12.

30 years of additional snow load data since the previous study and With the change to reliability-targeted values, the load factor on updated procedures for estimating snow loads from depth-only mea- snow loads has also been revised from 1.6 to 1.0 to represent the relisurements. The loads account for site-specific variability throughout ability basis of the values appropriately. Snow importance factors have the United States in both the magnitude and variation of the annual also been eliminated because values now are provided for each Risk ground snow loads. Additionally, this approach incorporates advanced Category. The 0.7 factor is intended to provide roughly equivalent spatial mapping that significantly reduces the number and size of case strength when design follows Allowable Stress Design (ASD) procestudy regions in mountainous areas and eliminates discontinuities in dures. For some materials, the ratio between design strength given design values across state boundaries (Bean et al., 2021).

by Load Resistance Factor Design (LRFD) procedures and design

36 STRUCTURE magazine

strength given by ASD procedure is 1.5. For some other materials, the ratio varies depending on the limit state being checked. The inverse of 1.5 was rounded to 0.7 for this purpose.

The same database and mapping scheme was used to prepare a 20-year mean-recurrence interval map and values for use when evaluating serviceability. The new value, 100 percent of the 20-year MRI load, is based upon the judgment of the ASCE 7 Snow and Rain Load Subcommittee and represents an increase for ASCE 7, but is less than the load specified in the International Building Code (IBC) 2021 (100 percent of the 50-year MRI load).

Figure 3. Map of winter wind parameter W 2.

Winter Wind Parameter Since the late 1980s, snow drift loads in ASCE 7 have been a function of the size of the snow source area as characterized by the ground snow load, Pg, and the upwind fetch length of the snow source area, lu. Recent research has shown that the drift load is also a function of the winter wind speeds. The addition of a winter wind parameter is intuitively appealing since one expects, with all other things being equal, that locations with relatively calm wind in winter would have smaller drifts than locations with strong winter winds.

The new relation for the drift height, hd, in ASCE 7-22 is hd = 1.5

√

74 70 1.7 Pg. lu. W2 Equation 7.6-1 γ

where γ, as before, is the snow density. The new parameter W2 is defined as the percentage of time during the winter (October through April) when the wind speed is greater than or equal to 10 mph, the nominal threshold for wind-induced snow drifting. Figure 3 presents the winter wind parameter for the lower 48 states.

Note that West of the Rockies and in the Southeast, W2 is comparatively small (typically 0.25 to 0.45), while in the Midwest and Northeast, W2 is comparatively large (typically 0.45 to 0.65). As such, the new winter wind parameter has about as strong an influence on drift surcharge load (proportional to the square of the drift height) as the ground snow and upwind fetch parameters. That is, the new ground snow load, Pg, for the lower 48 states varies from nominally 12 psf to 120 psf. Hence, its influence is nominally a factor of (120⁄12).74 or about 5.5. The upwind fetch typically varies from 100 to 1000 feet, and its influence is nominally a factor of (1000⁄100).70 or about 5. The winter wind parameter, W2, varies from 0.25 to 0.65 and hence is nominally a factor of (0.65⁄0.25)1.7 or about 5.1.

An advantage of the functional form of the relation is that there is no need for a “lower bound” drift size. That is, with the old drift relation, hd = 0.43 (lu)33 (Pg + 10).25 – 1.5, one calculates a negative drift height for low values of Pg and lu. With the new functional form, the drift height is positive for all possible combinations of the input parameters.

The most frequently asked question about the new drift approach is whether the drift loads, in general, will increase or decrease. For

locations with a low W2 of 0.25, the new drift heights are typically 50% to 70% of the old ASCE 7-16 drift heights, on average about a 40% decrease. For locations with a high W2 of 0.65, the new drift height is typically 100% to 150% of the old ASCE 7-16 height, on average about a 25% increase. For locations with an average W2 of 0.45, the new drift height is typically 75% to 110% of the ASCE 7-16 height, on average about a 10% decrease. As such, one could argue that the snow drifts from ASCE 7-22 are, on average, a bit less conservative than those in ASCE 7-16.

Summary This article summarizes some of the more substantive changes to the snow provisions of ASCE 7-22. The changes to the ASCE 7-22 ground snow loads are based upon 30 years of additional data, represent a shift away from uniform hazard to uniform risk, and significantly reduce the Case Study regions. The addition of a winter wind parameter accounts for the variability in winter wind speeds on drift loads.

Part 2, in an upcoming STRUCTURE issue, will review other revisions to the snow loads, including a more accurate estimation of the horizontal extent of windward drifts, revised thermal factors Ct to account for the current trends in roof insulation and venting, and guidance on the design loads for snow capture walls. Part 2 will also discuss a significant change to Chapter 8, including adding an explicit ponding head to the rain load and a simple relation for calculation of the ponding head.■ Full references, Figure 1, and additional information are included in the PDF version of the online article at STRUCTUREmag.org.

Michael O’Rourke has been a Professor in the Civil Engineering Department at Rensselaer Polytechnic Institute since 1974. He served as the Chair of the ASCE 7 Snow and Rain Subcommittee from 1997-2017 and currently serves as the Vice-Chair and a Fellow of the Structural Engineering Institute (SEI) (orourm@rpi.edu).

John F. Duntemann is a Senior Principal at Wiss, Janney, Elstner Associates in Northbrook, Illinois. He is the current Chair of the ASCE 7 Snow and Rain Subcommittee and a Fellow of the Structural Engineering Institute (SEI) (jduntemann@wje.com).

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structural DESIGN The Long Road

Advancing First-Generation Performance-Based Seismic Design for Steel Buildings Part 3: Future Efforts for All Structure Types

By Matthew Speicher, Ph.D., and John Harris, Ph.D.

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apabilities to conduct a performance-based seismic design (PBSD) of retrofitted existing buildings and new buildings have advanced exponentially over the past 25 years. This progress has augmented our knowledge of building behavior given an earthquake intensity. Still, we must be cautious Figure 1a. Theoretical range of building performance and relative placement of safety-based and recovery-based goals. of considering a PBSD as an exact Figure taken from NIST (2021a). answer; instead, a PBSD gives us information to support decision-making. There is still much work As an example, assume that an assessment is being conducted for the needed to support PBSD capabilities, and this depends on the type collapse prevention (CP) structural performance level (SPL) at the of assessment being conducted. At the same time, a vision for the risk-targeted maximum considered earthquake (MCER) prescribed not-so-distant future must also be established. in ASCE 7, Minimum Design Loads for Buildings and Other Structures The previous two parts of this series on advancing first-generation (2010 edition and later). Furthermore, suppose the collapse risk is PBSD principles and provisions for steel buildings (STRUCTURE, taken according to ASCE 7. In this case, the question becomes what October and November 2021) discussed the history of PBSD. They percentage of components needs to fail the CP SPL to achieve a 10 also outlined a project initiated at the National Institute of Standards percent probability of collapse given MCER shaking? Since there is and Technology (NIST) that evaluated what advancements could be no mechanism to assess risk based on the analysis results, exposure made. That project started by benchmarking ASCE 41 to ASCE 7 to risk cannot be communicated to shareholders and stakeholders. to develop a baseline.

A mechanism is needed to relate failure (for any performance level) This third and final article highlights several concepts that could of components based on consequences posed to the building owner, advance current PBSD capabilities. This article goes beyond steel occupants, service users, etc. An example of such an approach could buildings and takes a heuristic view of needs for all types of building be that a building poses more than a 10 % probability of not satisfying construction, non-building structures, and lifeline infrastructure a performance target (based on collapse, economics, loss of function, (generically referred to here as a system). These concepts can apply etc.) if either of the following occurs: to both first- and next-generation PBSD and include the following: • more than some percentage of the total structural components in • intrinsic risk assessment; one direction do not satisfy the target performance level; and • procedures and metrics to evaluate functional recovery time; • more than some percentage of the structural components • multi-system coordination; and resisting seismic force or deformation in one story in one direc• resilience-based seismic design. tion do not satisfy a target performance level.

These concepts may be initiated by NIST or by any partner agency The challenge would be defining the percentages in a codifiable in the National Earthquake Hazards Reduction Program (NEHRP). manner for policymakers and easily understood by the public. NIST Realization of these concepts will be dependent upon available GCR 12-917-20: Tentative Framework for Development of Advanced resources.

Seismic Design Criteria for New Buildings (NIST 2012) started evaluating risk targets for new buildings for adoption by ASCE 7. The same process can advance ASCE 41 using the methodology given in the Intrinsically Evaluating Risk Exposure Federal Emergency Management Agency’s (FEMA) P-58: Seismic First-generation PBSD principles contained in the latest edition Performance Assessment of Buildings (FEMA 2015), which can explicitly of ASCE 41: Seismic Evaluation and Retrofit of Existing Buildings evaluate seismic risk in a probabilistic sense.

(ASCE 2017) fundamentally result in a component-level binary pass or fail evaluation. A consequence of this process is that component Functional Recovery Time performance and the potential need to retrofit or replace is based upon an analysis output rather than the effect that the component A common consequence of an earthquake is interruption of building performance has on the system's overall performance. Therefore, an functions and operations and community support services (e.g., power engineer cannot effectively use the assessment results to explicitly or water distribution). These downtimes can range from a few hours evaluate risk, which is highly dependent upon the degree of redun- to years. Recovery time for a building is impacted by the following, dancy built into the building. to list a few factors:

38 STRUCTURE magazine

• the extent of damage to building systems (structural or nonstructural) and the operative contents used for services; • availability of financial and damage assessment resources; • demolition and repair processes, including mobilization time; • construction material availability; and • recovery of lifelines supporting building function and operation.

Recovery time may also be tied to community-level indirect losses such as loss of employment, displacement, and interruptions to education, childcare, and community services. This is discussed later in this article. Anecdotally, after the M7.1 Anchorage, AK, earthquake in November 2018, a building housing a notable coffee service sustained ceiling damage (non-structural). The business continued the next day by removing the ceiling to avoid such losses.

There is a need to develop and implement enhanced performance levels and PBSD guidance that address post-earthquake re-occupancy and functional recovery time. Figure 1-a shows an example of what the performance continuum may look like within this performance objective. Comparing Figure 1-a to Figure 1-b (also included in Part 1 of this series, October 2021) indicates that there may be cases when functional recovery governs performance and other cases when collapse prevention governs, but in no instance shall collapse prevention be overlooked. Within this effort, a set of distinguishable terminologies must be developed as well as a clear understanding of the result. For example, function and operation may be interpreted differently within the same organization. Is continuity of operations by providing services elsewhere (or in their parking lot) deemed to satisfy recovery requirements? Does the case satisfy re-occupancy requirements when the only elevator in a building is down, and one stair system is operational when the building may serve users who cannot use stairs? These are just a few of the multitude of inquires that need to be addressed by architects and engineers to define the needed metrics. Additional topics can be found in NIST SP 1269: NIST-FEMA Post-Earthquake Functional Recovery Workshop Report (NIST 2021b), which summarizes the feedback received by workshop participants on functional recovery concepts and options.

There is a need to enhance PBSD to support integration across systems and sectors beyond prioritization by risk categories.

In the latest reauthorization of NEHRP (Public Law 115-307, December 2018), NIST and FEMA were tasked to report on recommendations for improving the built environment and critical infrastructure to reflect performance goals stated in terms of postearthquake re-occupancy and functional recovery time. Their report, FEMA P-2090 / NIST SP-1254: Recommended Options for Improving the Built Environment for Post-Earthquake Reoccupancy and Functional Recovery Time (NIST 2021a), identified seven recommendations for design or retrofit of buildings and lifeline infrastructure that would culminate in a framework to address re-occupancy and functional recovery time.

Multi-System Coordination The results from a current PBSD, either using ASCE 41 or FEMA P-58, tend to focus on assessing the design or retrofit of a single building. This approach caters to a single isolated building or even an organization where operations within multiple buildings are mutually exclusive.

However, it does not address interactions among multiple associated buildings, among multiple lifeline infrastructure sectors, or between a combination of the two. For example, how does the performance of one school across town affect another school with regards to consequences to the school district? Assume that a school is closed due to earthquake damage (Figure 2, page 40). In this example, the students are required to go to another school; however, that school is too small to handle the increased student population.

Consequently, the school splits classes and holds some on Saturdays. Further, how does the school district coordinate with the local Department of Transportation to address adjusting school operations based on changes in traffic pattern demands?

There is a need to enhance PBSD to support integration across systems and sectors beyond prioritization by risk categories. For example, allowing multiple systems to provide feedback to other potentially impacted systems during the design or assessment process enhances risk assessments and associated decision metrics.

Resilience-Based Seismic Design Figure 1b. Illustration of building performance when subjected to increased earthquake intensities.

(Part 1, October 2021)

The conceptual difference between PBSD and resilience-based seismic design (RBSD) is that the latter JANUARY 2022

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evaluates the performance of a system with regards to its impact on the performance of a more extensive network of building and lifeline infrastructure systems and sectors. RBSD is thus a potential mechanism to evaluate impacts on community resilience. With that said, there may be conditions when the performance of a system does not impact community resilience, and the prioritization of network systems must be made at the community level. For example, should a building housing a large construction material retailer be designed Figure 2. Example depiction of consequence decision-making based on PBSD interaction between multiple as an ordinary building if downtime can associated buildings within an organization. hinder the recovery progress of the residential market it serves? economic impacts, to list a few. However, an aspect of resilience that PBSD can be used to estimate whether a design is highly likely to be is more challenging to quantify is its impact on society. functional after an earthquake (e.g., sustains less damage). In so doing, The engineering community must be able to communicate risk expothe system will contribute to the community’s resilience. PBSD can sure to shareholders and stakeholders. The performance of a building be augmented to include the impact of utilities and services needed must be able to address the welfare of its occupants or the public that for the system to regain function and operations, but it is still focused use the services provided to progress. It is straightforward with PBSD on a system. RBSD can employ PBSD and incorporate prioritized to address physical damage and downtime of the physical structure as community resilience concepts such as addressing impacts to the a primary indicator of performance. The losses from consequences on transportation network and, in turn, how that system may impact society such as mental and physiological health, displacement, interother systems and services. Essentially, RBSD can be envisioned as a ruptions to education, work, childcare, and community services play series of nodal enhanced PBSDs within a network communicating a key factor in estimating the holistic performance needed in RBSD. with each other.

RBSD must also address compounding consequences from coinciConclusion dental hazards and/or sequential hazards and the societal responses to them. In this context, coincidental hazards are one or more hazards This article discussed several future concepts to advance PBSD. These unrelated to the earthquake hazard that may occur within the same concepts can be somewhat aspirational but nonetheless outline the response and recovery period. Sequential hazards are one or more needs for progress that, when integrated, build upon each other. secondary hazards that directly result from the earthquake that may PBSD is not a tool strictly used to circumvent prescriptive building occur within the same response and recovery period. code provisions or save upfront construction costs, though this has A metric for earthquake resilience is challenging to define, beyond been a result. Instead, PBSD provides a rational estimate of design qualitative characteristics – having the ability to withstand, respond performance in a future earthquake. It must also be used to understand to, and recover from an earthquake and its consequences, and not the associated risks that such a design may pose to the community it just one earthquake. Moreover, quantitative assessment of resilience serves. Unfortunately, decision-makers generally only see part of the can only be measured after the impact on a network of systems from picture. Absent appropriate financial incentives, public and private an earthquake is known because response and recovery are time- organizations tend to invest in measures that they believe protect dependent functions without pre-defined timelines. Therefore, the their economic welfare, not necessarily those that augment the comsubsequent resilience score (for the next earthquake) is a function munity’s wellbeing. of the measurable change in resilience based on mitigation efforts, With the current trend towards defining and implementrepair or improvements, availability of construction resources, and ing resilience measures and guidance, it is difficult to continue along the path where new and existing buildings can be treated differently. In the eyes of the public, there is no difference in the function and operations of either. It could be possible within the context of RBSD to envision ASCE 41 and ASCE 7 as one standard. Seismic safety is a choice based on risk, and the developed tools need to address these risks so that users may augment them as needed and set priorities to benefit the community they serve.■ Full references are included in the online PDF version of the article at STRUCTUREmag.org.

Matthew Speicher is a Research Structural Engineer in the Earthquake Engineering Group at NIST. John Harris is the Acting Deputy Director of NEHRP and a Research Structural Engineer in the Earthquake Engineering Group at NIST.

structural ANALYSIS Two-Stage Analysis Loophole

By Steven Shepherd, S.E., and James McDonald, S.E.

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ngineering projects and building code provisions can often seem like Rorschach tests where two people looking at the same thing can draw sharply different conclusions. This article reviews the twostage analysis procedure in ASCE 7-16, Minimum Design Loads for Buildings and Other Structures, to consider if the provisions are an innocent inkblot or possibly may be interpreted differently by some.

Engineers simplify what is too complex to solve and what is too complex to solve efficiently in practice. The simplification imposes the obligation to validate that it does not result in a solution that works for the simplified model but is invalid for the complex realities.

Because simplification is often an imperfect step away from reality, simplification incurs the obligation of conservatism. The authors’ purpose is to explain how the two-stage analysis simplification can be applied inappropriately to allow for designs that do not provide the level of safety intended by the code. This article also offers remedies to prevent future misuse of this procedure.

A bold and useful simplification in the code’s seismic provisions is the equivalent lateral force (ELF) procedure, predicated on the assumption of approximately equal deformation distribution in one dimension over the structure’s height. But an efficient and ubiquitous building type like podium construction with several stories of light framing perched on one or more levels of concrete (or concrete with concrete masonry) framing is not consistent with the assumptions inherent in the ELF procedure. Using ELF for podium construction can result in mass from the heavier base being applied as inertial loads to the flexible upper portion. Rather than subject podium construction to the rigors and expense of a dynamic analysis, code authors opted for another simplification to keep the ELF procedure on the table for the design of podium construction by adopting the two-stage analysis provisions first introduced in the 1988 Uniform Building Code (UBC).

Conceptually, the two-stage analysis introduces a reasonable simplification to reflect the physical phenomenon of a rigid base not amplifying ground motions to more flexible stories perched above.

Further, it appropriately builds in conservatism because analyzing a single building as two separate shorter buildings results in shorter periods for individual building portions, and therefore, equal or greater base shear coefficients for each portion of the structure.

However, as demonstrated by the example building described later, this conservatism can be insufficient compensation if the two-stage analysis technique masks the deleterious effects of a base with a torsional irregularity. The two-stage analysis allows the flexible upper portion to be designed as a separate structure fixed at its base using the ELF or modal response spectrum procedure. The reactions from the upper portion are transferred to the rigid base (lower portion), amplified, not reduced, as appropriate for relative seismic response modification factor (R) and redundancy factor (ρ) values. Consistent with the procedure’s imposition of a static force at the top of the rigid base, the lower portion is designed using the ELF procedure.

Figure 1. Example building's finite element model.

Code Provisions The two-stage analysis procedure provisions from ASCE 7-16, Section 12.2.3.2 are listed below: a. The stiffness of the lower portion must be at least 10 times the stiffness of the upper portion. b. The period of the entire structure shall not be greater than 1.1 times the period of the upper portion considered as a separate structure fixed at the base. c. The flexible upper portion shall be designed as a separate structure using the appropriate values of R and ρ. d. The rigid lower portion shall be designed as a separate structure using the appropriate values of R and ρ. The reactions from the upper portion shall be those determined from the analysis of the upper portion amplified by the ratio of the R/ρ of the upper portion over R/ρ of the lower portion. This ratio shall not be less than 1.0. e. The upper portion is analyzed with the equivalent lateral force or modal response spectrum procedure, and the lower portion is analyzed with the equivalent lateral force procedure.

To the main point, these provisions strain the obligation to validate the simplification. Only (a) and (b) provide restrictions to apply the procedure; the other items specify how the procedure is used. While item (a) imposes a relative stiffness requirement, the stiffness parameter is undefined. More importantly, the stiffness obligation does not mandate that the lower portion have properties that provide support equivalent to “fixed at the base.” While item (b) imposes a requirement to compare dynamic properties of the two portions, the period comparison does not ensure the lower portion responds rigidly as the procedure allows the engineer to assume. In the authors’ opinion, although compliance with item (b) is not uniformly adhered to, the provision at least helps diligent practitioners and code enforcement officials keep designs closer to the code intent of a fixed base. continued on next page JANUARY 2022

41

Changes to the two-stage analysis procedure proposed for ASCE 7-22 clarify the application of the ASCE 7-16 provisions but do not result in significant changes to the procedure or the criteria to qualify for its use.

While it is necessary to stipulate quantification of the relative stiffness and period, leaving these as the only criteria to qualify for the simplifying procedure expands eligibility beyond the original intent to simplify the design of podium construction. In this case, the engineering quest to quantify the dynamic nature of the podium obscures the requirement rather than clarifying it.

The loosely defined relative stiffness requirement between the upper and lower portions and the absence of a requirement specifying the lower portion to provide a fixed base to the upper portion throw open the gates for misapplication of the two-stage provisions. The following sections demonstrate some examples of potential misapplications.

Although this article elucidates a flaw in the provisions, a call to action requires the specter of significant consequences – remedied with the analysis of an example building.

applies limitations, penalties, and other requirements to structures with torsional and extreme torsional irregularities, but the two-stage analysis allows the upper portion of a building to potentially avoid these requirements even if it is significantly affected by an extreme torsional irregularity in the base. The code requirements associated with a torsional irregularity include the following: • A 25% increase in demands for collectors, collector connections, and connections of diaphragms to vertical elements of the seismic force-resisting system for buildings in SDC D through F • The structure must be analyzed using a 3-D model • The effects of accidental torsion must be amplified per Section 12.8.4.3 • Structures in SDC D through F exceeding 2 stories must be analyzed using a dynamic analysis • In addition to the above requirements, an extreme torsional irregularity also requires a 30% increase in the horizontal seismic forces through a redundancy factor, ρ, of 1.3 for buildings in SDC D through F

Height Limit Loophole

Example Building

One of the proposed updates for ASCE 7-22 clarifies that the height limits for a given seismic design category (SDC) and building type can be applied as measured from the base of the upper portion. So, if a special reinforced concrete shear wall is being designed in SDC D and is limited to 160 feet per Table 12.2-1, the designer can design a 181-foot-tall building using the two-stage analysis if the bottom 21 feet are at least 10 times as stiff as the upper 160 feet. Based on parametric studies, this is most likely to be the case for a building of similar construction and where the upper portion is more than seven times the height of the lower portion. At that aspect ratio for such a building, the bending flexibility contributes enough to the response to meet the required stiffness ratio. The period of the 181-foot-tall structure would be within the 1.1 limit of item (b) using the approximate period equation of ASCE 7. Note that buildings this tall can usually qualify for both requirements (a) and (b) by virtue of their height, even if the base of the structure is not significantly different than the upper stories.

The authors developed a 3-D finite element model of a 13-story concrete shear wall building with a large podium level having an extreme torsional irregularity at the bottom story (Figure 1, page 41) to investigate the effects that a torsionally irregular base can have on the upper portion of a building. The example building is set in a location of high seismicity, such as near a significant fault in coastal California classified as SDC D with seismic parameters SDS and SD1 of 1.57g and 0.65g, respectively.

The building has rigid concrete diaphragms (assumed) and four fullheight core shear walls at the center of the tower plan area. Relative to the tower, the bottom story has twice the area, a larger unit weight, and two additional perimeter shear walls at two orthogonal building edges. These conditions are typical of a tower and base configuration, plaza construction, and construction on a sloping site.

To compare the seismic demands obtained from a single, coupled building analysis to those obtained using the two-stage analysis procedure, the authors also created separate models of the 12-story upper portion and one-story base, as shown in Figures 2 and 3, respectively.

Torsional Irregularity Loophole The building qualifies for the two-stage analysis procedure with a base The two-stage analysis provisions do not address the effect that a approximately 160 times stiffer than the upper portion and a combined torsional irregularity in the stiffer model period 1.1 times that of the upper lower portion can have on the flexible portion. With accidental torsion effects upper portion. Torsional irregularities included, the largest ratio of maximum are common in podium construction story drift to average story drift at any where entry-level architectural features level in the upper portion building or sloping sites can necessitate an eccenmodel is 1.39, giving that model a tric layout of seismic shear-resisting torsional irregularity per ASCE 7-16, elements.

Table 12.3-1, but not an extreme torA torsional response in the base of the sional irregularity. On the other hand, structure results in a rotational accelerathe bottom level of the full structure tion being input into the upper portion. has an extreme torsional irregularity.

Moreover, if the center of rigidity at the Because of the torsional irregularities, base is eccentric to the center of mass in both the separate upper portion model both horizontal directions, the response and the coupled full building model are in the two horizontal directions will be required to be analyzed using a dynamic coupled for both the lower and upper analysis per ASCE 7-16, Table 12.6-1, portions. Neither of these effects is capand accidental torsion effects need to be tured in a two-stage analysis, where the included for both models. Both buildupper portion is analyzed as a separate ings were analyzed using a Response structure. Furthermore, ASCE 7-16 Figure 2. Example building's upper portion.

Spectrum Analysis (RSA) with forces 42 STRUCTURE magazine

Figure 3. Example building's base.

scaled per ASCE 7-16, Section 12.9.1.4, to match the base shear obtained from an ELF analysis. Effects of accidental torsion are captured per Section 12.9.1.5 by modeling a center of mass eccentricity in the dynamic analysis equal to 5% of the diaphragm length.

The shear force in one of the second story core shear walls is examined to compare the response of the upper portion using a two-stage analysis to the response from a combined building model, with the following observations: • The upper portion building response computed using the twostage analysis is not affected by horizontal directional coupling (HDC) and has a redundancy factor, ρ, of 1.0.

• The response from the coupled analysis, omitting effects of HDC and ρ, is 23% greater than that obtained from the two-stage analysis. This increases to 34% when adding in HDC effects using the 100-30 combination rule and increases to 74% when also including the ρ of 1.3 required for the coupled analysis.

• Drifts and displacements throughout the structure increased similarly to the shear wall shear demand.

• For this example, ASCE 7-16, Table 12.6-1 requires an RSA for the upper portion due to the torsional irregularity.

Performing a two-stage analysis for this building does not significantly affect the effort required to analyze the building.

Discussion The coupled analysis produced significantly greater responses than the two-stage analysis for this building. Prima facie, the authors accept the coupled analysis as being more reflective of the code intent. If such a building were designed using the two-stage analysis and the framing designed near the code limit states, the limit states would be exceeded based on an analysis of the same building subjected to the coupled analysis. Hence, the building, possibly code-compliant with the two-stage analysis, would not provide the level of safety intended by the code.

Conclusions and Recommendations The two-stage analysis procedure was developed to simplify the design of light-frame residential buildings on top of one or two-story concrete or masonry podiums. This simplification was needed because a coupled dynamic analysis of this building type has historically been impractical and, due to the significantly heavier base level, a coupled ELF analysis can significantly overestimate the story shears in the upper levels. In most cases, the two-stage analysis procedure produces reasonable, potentially conservative results for this building type.

However, the procedure was written in a way that can be applied to almost any building type if the building is tall enough.

The two-stage analysis approach is unnecessary for most structures without light framing as part of the primary lateral load path. This is predicated on the fact that full-building finite-element modeling of these building types is already common practice and is not made significantly more difficult by the presence of a podium. However, this procedure can be used for such buildings to reduce seismic demands from what the code intends.

Buildings with extreme torsional irregularities at their base induce a torsional response in the upper portion of the structure, even if the upper portion of the structure does not have an extreme torsional irregularity. As currently presented in ASCE 7-16 and proposed for ASCE 7-22, the two-stage analysis allows the designer to ignore torsional effects from the base when designing the upper portion of the structure. It also allows the designer to bypass the limitations, penalties, and other requirements associated with this irregularity when designing the upper portion of the structure.

One of the proposed changes to the two-stage analysis procedure for ASCE 7-22 clarifies how height limits are to be interpreted for the procedure but also opens the door for misuse of the procedure to increase height limits for certain building types.

The authors recommend modifying the provisions for the two-stage analysis procedure to accomplish the following: • Limit the procedure so that it can only be applied to lightframe structures over concrete or masonry bases.

• Impose a maximum period on the base model with the mass of the upper portion lumped at the top of the lower portion. The period should be short enough to approximate a rigid dynamic response and should not be a relative requirement based on the period of the upper portion.

• Require the designer to account for the effects of torsional response in the base when designing the upper portion of the structure, including rotational accelerations, horizontal response coupling, and other code requirements associated with torsional and extreme torsional irregularities.

• If the recommendation to limit the procedure to light frame upper portions is not implemented, require the height limits of Table 12.2-1 to be measured to the base of the full structure rather than the base of the upper portion.

• Revise the code commentary to express the intent of a twostage analysis.

Closing The building code is a minimum standard for safety and should not leave room for interpretations that fail to achieve the code-intended level of safety. Hoping that conventional interpretations, or what some would consider reasonable judgment, covers engineering flaws in the code is professional abdication. Or, returning to the Rorschach analogy, why would we create an inkblot that one person could interpret vastly different than another?■ References are included in the PDF version of the online article at STRUCTUREmag.org Steven Shepherd is a Senior Consulting Engineer at Simpson Gumpertz & Heger Inc. in Newport Beach (srshepherd@sgh.com).

James McDonald in a Principal at Simpson Gumpertz & Heger Inc. in Newport Beach (jamcdonald@sgh.com).

JANUARY 2022

43

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engineer's NOTEBOOK The Hidden Cost of Copy and Paste Part 2

By Jason McCool, P.E.

I

n the first installment of this discussion (STRUCTURE, December 2021), I looked at 3 examples of bad habits in contract documents that can cause problems later for fabricators. This month, I offer an additional four examples.

“Just get something on paper!” Beware of reusing details to meet deadlines. Like it or not, your preliminary A typical overly-broad connection design note. drawings may get used at some point for pricing and – unfortunately – for fabrication or delegated design. I have questioned details so complex and formula-intensive that only another engineer fabricators about odd framing, only to find out that there had been can interpret them. In fact, if you have complex formulas in your several revisions put out by the structural engineer or architect since details (and I have seen “details” that looked more like Mathcad® the set of drawings they last shared with me. One project did not calcs), you should probably dial it back a bit. Under Option 1 get to what I would call a “complete” set of structural drawings until for connection design in the AISC COSP, they are unnecessary; Addendum #12. A lot changes as drawings develop, but making the under Option 2, even an experienced steel detailer often will not drawings look more complete than they really are causes problems know what to do with all of that “engineerese” (as experienced as people like me start to base their downstream work on what turns steel detailers have told me); and, under Option 3, the connection out to be mere eyewash. Saying “We will fix it in the addendum” is designer should already know how to design the connection. Most like saying you don’t have time to do it right. contract documents I have received in some way prohibit the fabIncomplete connection information. Delegated design is not what ricator from simply copying the structural drawings to use as their you do when you have run out of time or budget on a project; it shop drawings; do not do the same with the steel manual or other takes a fair bit of documentation to convey everything needed for engineering references. another engineer to do a complete connection design. If you are Misuse of AISC tables. I understand that referencing AISC’s delegating connection design to the fabricator, remember that, even Uniform Design Load (UDL) tables (e.g., Table 3-6) is a simple as fellow engineers, we are not mind readers. How much informa- one-line statement to add to one’s drawings or specs that “covers tion is enough? If you need more information than what you have you,” but does it really? AISC has been discouraging engineers from shown to solve the applicable connection design equations, so will doing this since at least 1995, but I still see it all the time. However, a the fabricator’s engineer. Consider the example of transfer forces at capacity table is not a load table. Short beams have the most capacity braced frames. I see a lot of braced frame elevations with brace forces and (typically) the least load for a given beam size. It does not make shown, but very few with transfer forces noted. Yet that is required sense to require a 50-kip connection capacity for a W10 infill beam information for delegated design as listed in the American Institute that is 4 feet long and has less than 1 kip of actual load while only of Steel Construction’s (AISC) Code of Standard Practice (COSP). requiring a 51-kip capacity on a W21 girder with a 50-kip reaction.

It is also information not possible for the connection designer to Besides leading to wild variations in safety factors, the practice can determine from a typical “envelope” of member forces without being also underspecify beam reactions of composite beams or beams with extremely conservative, possibly rendering a connection infeasible. large loads near the beam end. Hence AISC’s description of it as an AISC’s excellent Design Guide 29 on vertical bracing connections “inappropriate” practice. Another example is when a short, shallow (in Appendix D) lays out both the need to communicate transfer beam around a stairwell or elevator core is used in a low-seismic, forces and the difficulties in doing so. Passing design tasks to another wind-controlled braced frame. I might suspect there is only a 10-kip engineer requires passing on a lot of information. Consider the time beam reaction, at the most, to combine with the vertical component commitment for adequate documentation when deciding how – or of the brace force, but a 50% UDL directive from the project EOR whether – to delegate connection design. can easily require me to design that connection for a 40- or 50-kip Overly broad details and notes. Specificity takes time but is reaction plus the brace force vertical component, thus eliminating often essential in communicating your intent. Did the Engineer of options that are more than adequate and more straightforward to Record (EOR) who slapped the note “All steel connections to have fabricate and erect. Most engineers presented with extreme examples full moment capacity” on the drawings (see Figure) really mean all? from their drawings readily acknowledge this was not their intent, Even assuming they only meant the actual moment connections, but let’s break the cycle here and now before it becomes a question would a short cantilever beam supporting a 2-foot wide strip of on your next project. floor slab really need to develop the full beam moment capacity?

We should always be looking for ways to improve our While definitely erring on the side of caution, the best practice drawings, and next time I will wrap up with three final is to put the large safety factors into the connections where they ways to do that.■ are warranted, rather than just by blind application.

Jason McCool is a Project Engineer with Robbins Engineering Consultants There is another way to be overly broad. In the zeal to cover every in Little Rock, Arkansas (jmccool@robbins-engineering.com). conceivable variation in typical details, please do not make your

STRUCTURE magazine

JANUARY 2022

45

historic STRUCTURES Niagara’s Upper Falls Bridge Failure

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng, P.E., P.L.S.

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effert Lefferts Buck (STRUCTURE, December 2010) had long experiences with bridges across the Niagara River. First, he replaced wires and added anchorages to Roebling’s 1855 suspension bridge. He followed this with replacing the wooden trussing with iron and, still later, he replaced the stone towers with iron, all of these without stopping traffic for any extended period. These modifications took place between 1877 and 1886. With his associate R. S. Buck, he replaced Edward W. Serrell’s Lewiston/Queenston Suspension Bridge in 1889. He widened and strengthened Samuel Keefer’s 1868 Honeymoon Suspension Bridge in 1888. Unfortunately, the deck collapsed in a fierce windstorm in 1889, shortly after it opened. He rebuilt it within two months. In 1895, the owners wanted to widen the bridge and have it support trolley traffic. Buck was chosen to design and build an arch bridge under and around his Bridge complete. bridge in 1897. He chose a braced arch, open-spandrel span of 840 feet that made it the longest arch span in the world when it and long-continued southwest gale on the lakes. At about 4 PM, the opened on June 20, 1898. river channel under the ice became choked, and within a few minutes, In addition to fire, decay, and floods, ice jams had taken out many the water near the head of the ice-field rose to a height of 25 feet or of the early wooden bridges built in the Northeastern United States. 30 feet above its level at the bridge, three-quarters of a mile below, History had shown that ice jams were frequent in this part of the and began to pour over the pack for some distance below the Falls.

Niagara River below the falls. Buck placed the abutments on the The pressure caused the whole pack to move downstream. Not only canyon walls at an elevation he believed to be above any recorded ice was the ice piled up over the masonry abutments, but it was swept jams. Six months after the bridge opened, large quantities of ice came against the steel-work of the arch as high as the upper chord (panelover the falls and formed a large ice pack. Usually, this ice was confined point 2) on each side of the river. As it struck the rib chords, it was to the center third of the river due to strong eddies moving upstream shaved off as with a knife and deposited in large masses upon the on the sides. This ice pack reached thicknesses of over 100 feet, with upstream truss members and the lower laterals. The bridge quivered up to 80 feet being below the surface and 25 feet above the surface. from end to end as the ice ground against it but did not sway. The Buck wrote, the “high water level and the movement of the ice were pack moved 250 feet in about 10 minutes, after which the channel investigated before fixing the span of the arch and the elevation of underneath the ice cleared and the water subsided as rapidly as it had the abutments; according to the worst conditions previously known.” risen. On each side of the river, one main lateral and one sub-lateral, He described the events of January 22, 1899, as follows, “an ice-jam four members in all, were badly bent. The abutments were uninjured, took place which exceeded all past experience. The ice field, firmly and no other damage was done.” One account has it that men were anchored to both shores, then caught a heavy run of ice coming put to work with dynamite to keep the ice away from the steelwork down the river, the water of which was greatly augmented by a stiff as much as possible.

Plan and profile of the bridge.

46 STRUCTURE magazine

However, the fates were on his side this time, and he straightened the bent struts immediately and replaced them in the spring. He wrote, “in order to guard against similar trouble in the future, heavy concrete walls were built around the abutments, extending as far out in front as possible, and the first two panels of laterals in the plane of the lower chord on each side of the river were changed from latticed struts to plate-webbed struts.”

The bridge was not threatened again until January 25, 1938, forty years later, when a 5-day January thaw and major windstorm off Lake Erie again jammed the river gorge below the falls with ice. The ice rose to over 50 feet above the river’s mean water level and started to move at glacial speed down the river. Despite Buck’s 1899 reinforcement, the ice crumpled the main arch members of the bridge near the skewbacks resulting in the bridge being closed at 9:15 on the 28th. The bridge, surprisingly, held together for another day, and, at 4:20 PM on the 29th, the bridge collapsed into the Niagara River. A reporter for the Evening Review wrote of the failure in the following way, “Belch may seem a strange term to attach to the death of the long, slim span, yet that was the sound that thundered up to my ears. It was as though the great spidery giant was ridding itself of the terrible pain which had crept up into its bowels, bowels that were struts, spars, rivets, and all those structural bits that went into its creation. The sound was too robust for a sigh, more startling than a cry. And death flicked his grim harp in the echo which fled frantically up and down the cliffs, finally smothered in the spume of flying snow that jerked into the air as the shaking sections crashed into the ice.

Those two sections at the ends slowly sloped downward in a shower of powdery snow, dirt, and tumbling rock. Then the two sections inside these began to rise as though in a frantic effort to escape that mangling death below, attempting to soar skyward. It was a last futile gesture. Slowly – it seemed minutes though it was but split seconds Bridge after the collapse.

– the giant folded into the masses of ice. The result was a giant “W” for Winter resting on the river ice.”

Pieces of the wooden deck were shot into the air from the center of the Fortunately, the early warning kept people off the bridge, and there wreck as the end came. Loud cracking noises were heard as the bridge were no fatalities. Since there was no loss of life, there was no Coroner’s twisted and turned with the ice breaking up beneath it. Soon there Inquest to look into the failure. With that, was nothing left but floating fragments of over forty years after its opening and thirty the flooring, which were carried down the years after Buck died, the longest arch span river. It was a spectacular funeral, in keeping in the world at the time of its construction with the dramatic way in which the bridge In short, he, like the engineers disappeared. The bridge had served its funccame to its end.” tion well but had fallen prey to the severe The bridge was replaced in 1941 with of today, designed for what he weather typical around the Great Lakes. another steel arch bridge designed by considered a worst-case scenario.

The owners of the bridge immediately Waddell and Hardesty. It is located about blasted it into three parts as it lay on the 400 feet downstream from Buck’s Bridge Unfortunately, however, a unique set ice. After attempts to salvage the steel, and is a hingeless steel box arch with a span of weather events created an even they found that they could not remove of 950 feet. It, like its predecessor, is called worst-case scenario, and the bridge any portion of it. So it sat there until, on the Honeymoon or Rainbow Bridge.

April 12 at 7:10 AM, the ice let loose and It may be suggested that Buck should have failed after a life of 40 years. the end span dropped into the water to be known the ice could rise this high in the followed by the sinking of half of the arch area of his arch abutments. He, however, on the American side. On the following relied upon historical information to place day, at 3:25 PM, the remaining section them above any danger of the ice taking of the bridge began to move down the out the bridge. In short, he, like the engiriver on the ice. As it moved, the “huge ice floe turned pointing the neers of today, designed for what he considered a worst-case scenario. bridge section like the prow of a freighter as it sailed down the river Unfortunately, however, a unique set of weather events crewith the current.” It was described as follows: ated an even worst-case scenario, and the bridge failed after “It was a most unusual funeral procession. The spectators followed a life of 40 years.■ the progress of the bridge, running along the River Road to keep up with the floating wreck. It continued down river on its icy bier for what Dr. Frank Griggs, Jr. specializes in the restoration of historic bridges, having seemed an impossible distance for such a heavy weight until it reached restored many 19 t h Century cast and wrought iron bridges. He is now an a point just opposite the foot of Otter Street, almost a mile from the Independent Consulting Engineer (fgriggsjr@twc.com). starting point. Here it sank at 4:05 PM. JANUARY 2022

47

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48 STRUCTURE magazine

Resource Guide forms are now available on our website.

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INSIGHTS Building Safety Assessments Following the Sparta Earthquake

By Colby Baker, P.E.

A

magnitude 5.1 earthquake struck near 1886, a magnitude 5.8 earthquake near Sparta, North Carolina, on August 9, Washington, D.C., in 2011, and a mag2020, at 8:07 a.m. local time. The event was nitude 5.2 earthquake in Buncombe the strongest North Carolina earthquake County, North Carolina, in 1916. since 1916, producing “very strong” shaking Additionally, earthquakes on the east and over 100,000 Did You Feel It? reports coast tend to affect larger areas than throughout the southern, midwestern, and earthquakes of similar magnitude along northeastern United States. More than 500 the west coast. For instance, the 2011 residential and commercial structures were earthquake in Washington, D.C., was damaged during the quake. As a result, North felt up to 600 miles from its epicenter.

Carolina Emergency Management deployed In contrast, the magnitude 6.0 eartheight post-disaster building safety evaluaquake in Napa Valley, California, in tors, including four engineers, to evaluate 2014 was felt just 250 miles from its the damaged structures for safe occupancy. epicenter. These differences are most Unsurprisingly, the most visible damage likely due to regional variations in the was non-structural and included collapsed composition and geologic history of the ceiling finishes, toppled chimneys, bowed underlying tectonic plates. or collapsed brick veneer, and displaced Furthermore, many of the larger earthcontents. Structural damage was generquakes in North Carolina occurred ally limited to those systems known to be when the state was more rural. The seismically hazardous such as plain (i.e., existence of denser population centers unreinforced) brick, concrete block, and only magnifies the potential for damage stone masonry. While most single-family Displacement of unreinforced masonry at an industrial building during future events. in Sparta, NC, following the August 9, 2020, earthquake. residences were wood-framed construcWhat can structural engineers do to tion, their foundations were often composed of plain concrete block promote seismic resilience in their communities? A first step might masonry. In general, short unreinforced masonry stem walls sur- be to become familiar with and promote the use of FEMA P-154 vived with minor cracking; however, deep crawl spaces and walkout Rapid Visual Screening of Buildings for Potential Seismic Hazards: A basements suffered considerably, developing wide cracks and lateral Handbook. The methodology described in this document enables differential displacement within the mortar joints. building owners to identify whether their structures are vulnerable In total, 17 buildings were red-tagged (i.e., deemed unsafe for re-entry), to earthquake damage. Once potentially vulnerable buildings are including 10 single-family dwellings, one school, and a handful of identified, the owners can prioritize future earthquake risk reduction commercial and mixed-use buildings. At least 5 of these buildings were and mitigation efforts. later razed as a result of the damage. Moreover, 260 additional buildings Additionally, engineers may choose to volunteer their expertise in were yellow-tagged (i.e., deemed restricted use) due to the potential for the aftermath of an earthquake by providing post-disaster building falling debris and/or further displacement during aftershocks. safety assessments. These assessments are vital to the expeditious reThe construction of each of the red-tagged structures pre-dated habitation of affected but otherwise safe structures and the long-term 1978 and consisted of plain masonry foundations and/or exterior recovery of the community at large. walls. North Carolina building codes of that era generally allowed To become a volunteer post-disaster building safety evaluator, contact unreinforced masonry in foundation walls up to 7 feet in height, as your Structural Engineering Emergency Response (SEER) Committee permitted by the building official.

Chair. A complete listing of the committee chairs can be found at To date, little consideration has been given to the seismic retrofit of www.ncsea.com/committees/seercommittee. The requirements vary buildings along the Eastern Seaboard, particularly residential dwell- by state, but, in general, qualified volunteers must complete a disaster ings and rural communities. This oversight is almost certainly due to responder training course such as the Safety Assessment Program the high economic investment required for such retrofits combined (SAP) offered by the California Office of Emergency Services with a perceived immunity to seismic risk. However, the seismic risk (CalOES) or the “When Disaster Strikes…” Institute offered may be higher than realized. by the International Code Council.■ The eastern United States has a history of strong earthquakes. Several Colby Baker is a Forensic Structural Engineer with U.S. Forensic and a faults along the east coast were created 250 million years ago with structures specialist with North Carolina Emergency Management. He the formation of the Appalachian Mountains. These faults caused co-chairs the NCSEA SEER committee (colby.baker@usforensic.com). the magnitude 6.7 earthquake in Charleston, South Carolina, in

STRUCTURE magazine

JANUARY 2022

49

business PRACTICES Positioning for Continued Success

By Kacey Clagett, LEED AP BD+C, and Tiany Galaskas

J

anuary typically prompts business planning for a new year.

However, since early 2020, the Covid-19 pandemic has made business planning and operations much more volatile. While we all have learned to be more resilient to thrive, it is difficult to gauge when companies will experience more stable economic conditions. This article offers some insight into economic conditions and how to position your company to continue to be successful in the coming months.

Industry-Specific Economic Conditions

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Even in regions where there have been fewer governmental pandemic restrictions, the economic effects of Covid outbreaks have been broad, affecting entire market sectors, the availability of design and construction labor, and supply chains. In 2021, office vacancy rates in cities like New York and San Francisco approached 20%, putting further pressure on construction financing and stalling new building starts. As Barry Sternlicht, CEO of the real estate investor Starwood Capital Group, said at the time, “It [is] very hard to underwrite these cities.” Mary Corley of Rosen Consulting Group, real estate economists, observed, “Covid has been an accelerant of trends we were already seeing before March 2020, such as the growth of suburbs and secondary markets, and we expect to see continued growth there. We don’t know if it’s to the long-term disadvantage of gateway markets. Gateways have a lot of existing infrastructure, which is not something to walk away from.” With such uncertainty, buyers will continue to look for ways to reduce risk in budgets and schedules.

At the same time, construction has continued robustly in such markets as residential, industrial, science, and technology. And while the pandemic has temporarily disrupted the office, retail, hospitality, higher ed, and government sectors, these will bounce back, likely with some fundamental changes. Many expect that economic problems will abate soon. In July 2021, Appleseed Strategy surveyed a national focus group of 39 design and construction companies.

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L O V E L A N D

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Across the spectrum, expect opportunities to retrofit existing buildings to new uses.

Asked about their predictions moving to the immediate future, respondents were largely positive: • Civic/government, healthcare, higher ed, industrial, mixed-use, and multifamily were ranked as good opportunities. Science and technology were seen as particularly strong.

• Corporate/workplace, cultural/nonprofit, and hospitality were perceived to have headwinds but still offered opportunities. Of the sectors queried, only retail was seen as struggling significantly in the near term.

• Geographically, all major U.S. regions were seen as presenting good-to-strong prospects. Those who worked internationally felt immediate prospects were good.

New Opportunities Across the spectrum, expect opportunities to retrofit existing buildings to new uses. Owners who can find the funding will reposition buildings in anticipation of a future boom. Underused shopping malls might be remade into logistics centers and mixeduse communities, and offices and retail eyed for multifamily.

Some building owners will try to convert to science and technology tenants. Temporary outdoor dining parklets will become permanent. Distressed properties will be bought up and reused.

A real boon to design and construction is the recently enacted $1.2 trillion federal infrastructure bill that ushers new funding for infrastructure, sustainability, and resilience. The bill provides $65 billion in clean energy; $50 billion in infrastructure resilience; $66 billion for Amtrak; $25 billion for airports; and $100 billion for roads, bridges and other projects. We can expect to see more federal projects as standalone commissions or as part of indefinitequantity, indefinite-definite delivery contracts. However, because

federal spending is proposed to be funded by increasing real estate taxes, critics argue that the infrastructure plans will depress commercial real estate. This snag will likely be ironed out. Much federal funding will trickle down as state and local government projects. This presents excellent opportunities for minority, disabled, and woman-owned businesses and companies who make meaningful actions toward diversity, equity, and inclusion in staffing and teaming.

Sustainability and resiliency design will continue to rise due to public policy, pressure from insurers, and severe weather events. For owners using institutional capital or those who are publicly traded, the adoption of ESG policies – environment, social, and governance standards – will increasingly demand that engineering companies help owners reduce their carbon footprint, use less water and energy, and avoid red-listed materials. Mary Corley of Rosen Consulting Group underscored the increasing influence of ESG policies on the real estate market.

As with retail, technology is accelerating significant changes across market sectors. Corley noted that remote work has become a significant force in the real estate strategies of many companies, causing them to rethink how much physical space they really need. Now, a change management outlook and approach comes before any real estate decision. Can a company thrive by relying on technology? Smart companies are focusing first on achieving the kinds of behaviors at work that will reach their desired outcomes. For many, a physical presence will be an option, not a given.

Questions to Ask The design and construction market will reach equilibrium as the pandemic subsides, but any disruptive event at this scale leaves lasting marks, bringing new market demands and ways of working. Make your business planning more resilient by taking advantage of the positive aspects of these disruptions. As you develop your plans and budgets, probe whether you are adequately assessing the inherent risks and opportunities in the markets and regions you work in: • What fundamental changes have occurred?

• What are the new, positive changes that we should leverage?

• How have these changes affected our clientele, and how can we help them position to succeed?

• Who should be our new target clients moving forward?

• How can we leverage technology not just to streamline operations but to advance the discipline of engineering and make us more desirable?

Keep the focus on higher return, lower risk ventures.

Build from your core strengths.

• Invest in technologies that advance the disciplines you practice.

Think about what will be automated ten years from now, so you can position yourself as a market leader. Be aware that these will be areas where nontraditional competitors may enter, yet this also presents partnership opportunities.

• Likewise, get ahead of other industry trends like building repurposing, net-zero energy and water, design/build, modularization, and ESG/CSR policies. Set the standard, and you may be rewarded with better fees and less competition. Remember that creativity never gets replaced.

• If a service or market sector you have relied on will not return to previous levels, scale back or eliminate it. As hard as it can be, do not let sunk costs influence your decisions.

• Develop future leaders through mentoring and increasing accountability for company performance.

• Always practice financial resilience. Create a strong safety net, streamline operations and expenses, forecast realistically, and build a business development culture across your company.■ Kacey Clagett (kacey.clagett@appleseedstrategy.com) and Tiany Galaskas (tiany.galaskas@appleseedstrategy.com) are Principals of Appleseed Strategy.

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We recommend that engineering companies pay increasing attention to several areas: • Keep the focus on higher return, lower risk ventures. Build from your core strengths. If you are heavily concentrated in market sectors or struggling regions, think strategically about how to reposition. In many situations, you can repurpose what you do with relatively modest efforts.

• Survey your clients. Besides the typical check-in questions, ask them about the opportunities and challenges they face with their client base. It may give you broad insights into how to position your company and improve customer service. There will always be demand for companies with excellent customer service.

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Do This Now

JANUARY 2022

51

NCSEA

NCSEA News

National Council of Structural Engineers Associations

Congratulations to the 2021 NCSEA Special Awards Honorees

The NCSEA Special Awards are bestowed on individuals who exemplify outstanding service and commitment to the association and to the structural engineering field. These dedicated servants, committed volunteers, and industry leaders will be recognized and applauded at the Awards Celebration on Wednesday, February 16, at the NCSEA Structural Engineering Summit. Join us in celebrating this year’s honorees!

Susan M. Frey NCSEA Educator Award

James Delahay Award

This award, established to honor the memory of Sue Frey, one of NCSEA's finest educators, is presented to an individual who has a genuine interest in, and extraordinary talent for, effective instruction of practicing structural engineers.

This award is presented at the recommendation of the NCSEA Code Advisory Committee, to recognize outstanding individual contributions towards the development of building codes and standards. It is given in the spirit of its namesake, a person who made a long and lasting contribution to the code development process.

Michelle Kam-Biron, P.E., S.E., is a California licensed SE and has extensively provided education nationally and internationally on conventional wood and mass timber construction. Ms. Kam-Biron brings a unique skill set to her presentations with a background consisting of over 25 years as a practicing structural engineer and plan check reviewer, an influencer and educator with WoodWorks and American Wood Council, and her current position as a Mass Timber Specialist at Structurlam Mass Timber Corporation.

She is a Past-President of SEAOSC and was recently inducted into the SEAOC College of Fellows. She serves on several committees such as NCSEA Basic Education and ICC Professional Development Council Education, and she is Past-Chair of ASCE-SEI Wood Education and SEAOSC Foundation and Women in Structural Engineering.

Robert Cornforth Award This award is presented to an individual for exceptional dedication and exemplary service to a member organization and to the profession. The award is named for Robert Cornforth, a founding member of NCSEA and treasurer on its first Board of Directors, and a member of OSEA. Robert H. Durfee, P.E., SECB, is a Structural Engineer, Vice President, and Chief Bridge Engineer at DuBois & King in the firm’s Gilford, NH office. His practice focuses on the design of bridge rehabilitation and replacement. His specialty is the restoration of historic covered bridges.

In 1994, Mr. Durfee helped found Structural Engineers of New Hampshire (SENH) and served on its Board of Directors as Secretary (1994-1997) and as President (1998-2000). He has served on several committees, including: Nominations, Structural Engineering Awards, Scholarship Award, NCSEA NH Annual Conference, and Public Relations. He has served as the SENH Delegate to the Annual Conference/Annual Summit since 1998.

Mr. Durfee is a founder of the Northeast Coalition of Structural Engineers Associations (NECSEA), serving as the SENH representative from 2001-2020. He serves on the NH Engineer/Young Engineer of the Year Awards Committee Jury.

52 STRUCTURE magazine

Ramon Gilsanz, P.E., S.E., F.SEI, F.ASCE, Hon. AIA NYS, is a founding Partner of Gilsanz Murray Steficek. In his 40-year career as a structural engineer, he has worked on a wide range of new construction and renovation projects, and volunteered across the world to help communities in need. Ramon has participated in six post-earthquake investigative teams, hurricane Sandy recovery efforts, and led the WTC7 collapse analysis on the national ASCE-FEMA building performance assessment team.

Additionally, Ramon actively contributes to the industry through several professional societies and committees including ASCE 7-16 and ASCE 7-22, American Concrete Institute (ACI), and the AISC committee on specifications. Ramon served on the NYC Department of Buildings’ Structural Technical Committee in 2008 and was Chair in 2014 and 2021, revising the NYC Building Code. He is also the Chair for the NYC Existing Buildings Code Development Structural Technical committee, whose building code is forthcoming.

NCSEA Service Award This award is presented to an individual who has worked for the betterment of NCSEA, member organizations, and the profession, to a degree that is beyond the norm of volunteerism.

Thomas DiBlasi, P.E., SECB, is the President of DiBlasi Associates, P.C., in Monroe, CT. He is a Past-President of NCSEA where he currently chairs the Code Advisory Committee and also serves on the Wind Engineering Subcommittee.

He also served on the NCSEA Continuing Education Committee as well as the ad-hoc committee that developed the NCSEA Model Code of Ethics. He has been a director of NCSEA Media since its inception. He is Past-President of the Structural Engineers Association of Connecticut (SEAConn) where he chairs the Code Advisory Committee and the Peer Review Committee. He also serves as the structural engineering representative to the Codes & Standards Committee of the State of Connecticut.

News from the National Council of Structural Engineers Associations Susan Ann “Susie” Jorgensen Presidential Leadership Award – INAUGURAL YEAR The Susan Ann “Susie” Jorgensen Presidential Leadership Award is presented to an individual who has demonstrated exceptional leadership potential through their activities within NCSEA and/or their SEA (even if they did not serve in a formal leadership role). The award is to be bestowed on candidates who embody Susan’s passion, vision, and legacy of leadership, and it is intended to celebrate increased participation of emerging leaders and encourage recipients to engage (or continue to engage) in formal leadership.

Katharine (Katie) A. Courtright, P.E., is a project engineer for JVA, Inc. in Denver, Colorado. She joined the Colorado SEA in 2012 as a student member when she was selected to receive a SEAC Scholarship. During college, Katie served as the SEAC Young Member Group Collegiate Liaison for the Colorado School of Mines. Upon graduation in 2014, she joined the Young Member Committee, becoming co-chair in 2016 and chair in 2017. Her service on the YMG Committee has included coordination of the annual SE/PE Study Group Kick-offs and AASHTO Review Sessions, as well as technical presentations, mentoring, and outreach events. Katie has been instrumental in encouraging student involvement in SEAC’s annual Gingerbread Bridge Competition. She is currently serving on the newly formed SEAC SE3 Committee.

THANK YOU TO OUR SPONSORS

Structural Engineering Summit – Feel the Love this February The NCSEA Structural Engineering Summit will be in New York City February 14-17 and online January 31-Feburary 24. The Summit offers unrivaled educational opportunities, an industry-leading trade show, and unique and fun networking opportunities.

Register for the conference, learn more, and book your hotel room ($219/night at the Hilton Midtown!) at www.ncsea.com/events/annualconference.

NCSEA Webinars

Visit www.ncsea.com/education for the latest news on upcoming webinars and other virtual events.

January 6, 13, 20 and 27; February 3 and 10 SEAOC and NCSEA Seismic Connections Design Series January 11 Business Development Moving Forward January 18 The Structural Engineer's Role in Getting to Net Zero February 8 Significant Structural Changes to the 2021 International Building Code February 22 Structural Engineering Considerations for Mid-Rise, Light Wood Frame Buildings Start your year off right with an NCSEA webinar subscription!

Subscribers receive access to a full year’s worth of live NCSEA education webinars (25+) and a recorded library of past webinars (170+) – all developed by leading experts; available whenever, wherever you need them!

Courses award 1.5 hours of Diamond Review-approved continuing education after the completion a quiz.

follow @NCSEA on social media for the latest news & events!

JANUARY 2022

53

SEI Update Learning / Networking

Happy New Year from the SEI Board and Staff!

Looking forward to all that’s new in 2022 SEI launches a new quarterly “FAQ on ASCE Standards: What you always wanted to ask” in this issue (see page 34). In addition, look for a new peer-to-peer forum on standards starting soon on ASCE Collaborate and a new virtual experience on ASCE’s Future World Vision during Engineers Week in February.

Join us for a new virtual 5-part SEI Standards series on ASCE 7-22 that kicks off with the first session (free) February 10 on the Overview & Changes for ASCE 7-22. The SEI SE2050 database and commitment program continues to expand and provide educational information to firms to achieve projects’ globally stated goal of net-zero carbon by 2050. Collaborative Reporting for Safer Structures – CROSS-US – will be presented at three major conferences. We look forward to getting back to in-person conferences and seeing many of you at Structures Congress April 20-23 in Atlanta and ETS October 2-6 in Orlando!

ASCE Diversity, Equity, Inclusion Best Practices Check out new resources at www.asce.org/diversity-equity-and-inclusion

SEI Online

SEI Events

www.asce.org/SEIEvents The 2022 SEI Standards Series will preview ASCE 7-22 as a 5-part series that reviews the changes from ASCE 7-16. This unique program includes a dialogue between the leaders and experts who develop ASCE 7 and a detailed technical presentation on the specific changes and three main hazards - Seismic, Wind & Tornado, and Snow/Rain. In addition, information will be provided on the ASCE 7 Digital Products/ Hazard Tool. Attendees are encouraged to join the discussion for the extensive live Q&A portion of the session.

• February 10, 2022: ASCE 7-22 Overview & Changes (FREE) • May 12, 2022: ASCE 7-22 Seismic • June 9, 2022: ASCE 7-22 Wind & Tornado • July 14, 2022: ASCE 7-22 Snow/Rain • September 8, 2022: How & Why to Use ASCE 7-22 in Your Practice Learn more and register https://collaborate.asce.org/integratedstructures/sei-standards Structures Congress – April 20-23, 2022, in Atlanta View the program and register at www.structurescongress.org.

Electrical Transmission and Substation Structures Conference – October 2-6, 2022, in Orlando Apply for a student scholarship to participate at www.etsconference.org.

Membership

NEW – Access your SEI Member Certificate Online

At www.asce.org, log in at the upper right of the page and select Manage My Account. You can renew, update your contact info, professional interests, preferences, bio, education/license details, and download your self-service SEI member certificate now available.

Follow SEI on Social Media: 54 STRUCTURE magazine

News of the Structural Engineering Institute of ASCE Advancing the Profession

Newly Updated Minimum Design Loads and Associated Criteria for Buildings and Other Structures, ASCE/SEI 7-22 Standard Now Available 2022 Edition of ASCE’s Most Widely Used Standard Revises All Environmental Loads, Including New Chapter for Tornado Provisions and Provides Digital Data for All Hazards

The newly updated ASCE/SEI 7-22 Minimum Design Loads and Associated Criteria for Buildings and Other Structures is now available. This national standard is the American Society of Civil Engineers’ (ASCE) most widely used standard and is an integral part of building codes in the United States and around the globe. Structural engineers, architects, and those preparing and administering local building codes will find the 2022 edition of the structural load requirements essential to their practice.

“Civil engineers are responsible for the design of the buildings and structures we work, live, and play in every day, and we must ensure those structures are safe for the public,” said Tom Smith, ASCE Executive Director. “With weather hazards becoming more extreme, this updated standard is essential to improve the resilience of our communities.”

“For more than 20 years, U.S. building codes have relied on the ASCE 7 standard as the authoritative source for specification of loads and related criteria used by engineers to design safe, economical, and reliable structures. Every six years, hundreds of volunteer professional civil and structural engineers, researchers, building officials, and construction professionals collaborate to update the standard, acknowledging new engineering research, evolving construction techniques, and society’s changing expectations and concerns,” said Ronald Hamburger, P.E., S.E., F.SEI, Senior Principal with Simpson Gumpertz & Heger, Inc., and chair of the ASCE 7-22 committee. “The 2022 edition includes first-ever criteria for tornado-resistant design and substantial improvements to the design criteria for atmospheric icing, earthquake, tsunami, rain, snow, and wind.”

The 2022 edition of ASCE 7, which supersedes ASCE/SEI 7-16, provides the most up-to-date and coordinated loading provisions for general structural design. Informed by past events, including Hurricane Michael in 2018 and the Joplin Tornado in 2011, this standard prescribes design loads for all hazards, including soil, flood, tsunami, snow, rain, atmospheric ice, seismic, wind, and fire, as well as how to evaluate load combinations.

ASCE/SEI 7-22 is different from past versions because, for the first time, the digital data is available via open access from the ASCE 7 Hazard Tool so that anyone can view the hazards that are relevant to their local community.

Environmental hazards used for building design were all updated, specifically new wind speeds along the hurricane coastline, improved tsunami run-up for highly populated west coast locations, increased accuracy of seismic design criteria, new national snow design data, risk-specific atmospheric ice criteria, and an entirely new chapter for tornado loads. In addition, ASCE 7-22 modernizes design requirements for cutting edge mass timber systems and composite concrete and steel systems and big box stores/warehouses, ground-mounted solar facilities, and elevated buildings.

“The addition of tornado loads represents a nearly decade-long collaboration between NIST and ASCE to significantly advance safety and resilience for buildings at risk of tornado impacts,” said Marc Levitan, Ph.D., A.M.ASCE, of the National Institute of Standards and Technology (NIST), who chaired the ASCE 7-22 Task Committee that developed the new tornado provisions.

In addition to the print version of ASCE 7-22 – available as a two-volume paperback set or as a PDF – ASCE 7 Online is a subscription service that provides digital access to ASCE/SEI 7-22, as well as to the previous 2016 and 2010 editions, with enhanced features that make it faster and easier to work in the Standard. Functionality exclusive to ASCE 7 Online includes a side-by-side display of the provisions and commentary; redlining to track changes between editions; real-time updates of supplements and errata; two-level corporate vs. personal annotations; and toggling between Customary and SI unit measurements. Corporate subscriptions are available.

For more information, contact asce7tools@asce.org.

To purchase the print or PDF version of ASCE/SEI 7-22, visit www.asce.org/asce-7.

To subscribe to the ASCE 7 online digital platform, visit https://asce7.online.

Errata

SEI Standards Supplements and Errata including ASCE 7. See www.asce.org/SEI. To submit errata, contact sei@asce.org.

JANUARY 2022

55

CASE in Point CASE Tools and Resources What's New in CASE Publications CASE has a variety of publications and tools to help firms deal with a wide variety of business scenarios. Whether your firm needs to establish new procedures or simply update established programs, CASE has the tools you need!

This month, you can find updates to the following documents, including a Force Majeure clause developed due to the recent pandemic.

CASE Agreement #1 CASE Agreement #2 CASE Agreement #3 CASE Agreement #4 CASE Agreement #5 CASE Agreement #6 CASE Agreement #7 CASE Agreement #8 CASE Agreement #9 CASE Agreement #10 CASE Agreement #11

An Agreement for the Provision of Limited Professional Services © An Agreement Between Client and Structural Engineer of Record for Professional Services © An Agreement between Owner and Structural Engineer as Prime Design Professional © An Agreement between Client and Structural Engineer for Special Inspection Services © An Agreement Between Client and Specialty Structural Engineer for Professional Services © An Agreement Between Client and Structural Engineer for a Structural Condition Assessment © An Agreement for Structural Peer Review Services© An Agreement Between Client and Structural Engineer for Forensic Engineering (Expert) Services© An Agreement Between Structural Engineer of Record and Design Professional for Services© An Agreement Between Structural Engineer of Record and Geotechnical Engineer of Record © An Agreement Between Structural Engineer of Record and Testing Laboratory© You can purchase these and other publications at www.acec.org/bookstore.

Wanted: Engineers to Lead, Direct, Engage with CASE Committees!

If you are looking for ways to expand and strengthen your business skill set, look no further than serving on one (or more!) CASE Committees.

Join us to sharpen your leadership skills and promote your talent and expertise to help guide CASE programs, services, and publications.

We currently have openings on all CASE Committees: Contracts – The Committee is responsible for developing and maintaining contracts to assist practicing engineers with risk management.

Guidelines – The Committee is responsible for developing and maintaining national practice guidelines for structural engineers.

Programs – The Committee is responsible for developing program themes for conferences and sessions that enhance and highlight the structural engineering profession.

Toolkit – The Committee is responsible for developing and maintaining the tools related to CASE’s Ten Foundations of Risk Management program.

To apply, your firm should: • Be a current member of ACEC • Be a member of the Coalition of American Structural Engineers (CASE); or be willing to join the Coalition • Be able to attend the groups’ usual face-to-face meetings each year: August, February (hotel, travel partially reimbursable) • Be available to engage with the committees via email and video/conference call • Have some specific experience and/or expertise to contribute to the group Please submit the following information to (mkroeger@acec.org), subject line CASE Committee: • Letter of interest indicating which committee • Brief bio (no more than a page) Thank you for your interest in contributing to advancing the structural engineering profession!

Follow ACEC Coalitions on Twitter – @ACECCoalitions.

56 STRUCTURE magazine

News of the Coalition of American Structural Engineers UPCOMING EVENT

2022 ACEC Coalitions Winter Meeting – San Diego, CA, February 10-11, 2022 The Winter Meeting is open to all CASE members and is an excellent opportunity to network with your peers and engage in meaningful dialog about the state of the industry.

Directly following the Coalitions Winter Meeting is the Small Firm Coalition (SFC) Workshop, Small Firms and Human Resources – Developing the Workforce for the Future; a workshop designed to address the HR Challenges of the small firm as we design the workforce of tomorrow.

AGENDA Thursday – February 10 1:30 pm – 3:00 pm CASE ExCom Meeting 3:30 pm – 5:00 pm CASE and CAMEE (Coalition of Mechanical and Electrical Engineers) Roundtable, moderated by the Chairs of CASE and CAMEE 5:00 pm – 6:00 pm Coalitions Reception

Friday – February 11 8:30 am – 12:00 pm

Educations Sessions – (PDH’s offered) • Remote Monitoring using Today’s Technologies – How new remote sensing technologies are changing the way civil engineers work.

• Navigating the Challenges of Distance Work – Managing legal, financial, and human resource activities for out-of-state employees.

12:00 pm – 1:15 pm Lunch 1:30 pm – 5:00 pm

CASE Committee Meetings Contracts – develops and maintains contracts to assist practicing engineers with risk management.

Guidelines – develops and maintains national guidelines of practice for structural engineers.

Programs – develops program themes for conferences and sessions that enhance and highlight the structural engineering profession.

Toolkit – develops and maintains the tools related to CASE’s Ten Foundations of Risk Management program.

1:30 pm – 5:00 pm

Small Firm Coalition Workshop ($$ Paid Event) – (PDH’s offered) • Small Firms and Human Resources – Developing the Workforce for the Future.

(This workshop continues Saturday morning, February 12, 2021.)

Registration is now open. To register, go to http://bit.do/coalition-winter-22.

Questions? Contact Michelle Kroeger at mkroeger@acec.org.

It's Time to Give Back…

At ACEC’s 2021 Fall Conference, the Coalition of American Structural Engineers (CASE) awarded Celeste Carmignani a $2,500 CASE Scholarship. Ms. Carmignani is working on a bachelor’s degree in structural engineering from the Colorado School of Mines.

CASE is currently seeking contributions for the structural engineering scholarship for the upcoming 2022-2023 school year. The CASE scholarship, administered by the ACEC College of Fellows, is awarded to a student seeking a Bachelor’s degree, at minimum, in an ABETaccredited engineering program. Since 2009, the CASE Scholarship program has given $37,000 to help engineering students pave their way to a bright future in structural engineering.

We have all witnessed stiff competition from other disciplines and professions eager to obtain the best and brightest young talent from a dwindling pool of engineering graduates. One way to enhance students’ ability to pursue their dreams to become professional engineers is to offer incentives through educational support.

Your monetary support is vital in helping CASE and ACEC increase scholarships to those students who are the future of our industry.

In addition, all donations toward the program may be eligible for a tax deduction, and you do not have to be an ACEC member to donate!

Do you know a deserving student who would like to apply for the CASE Scholarship? They can find out more by visiting www.acec.org/awards-programs/acec-scholarships-program.

JANUARY 2022

57

SPOTLIGHT Edmonton’s Stanley A. Milner Library Creativity in Design and Execution

A

landmark of its time, the Stanley A. Milner Library has lived in the civic heart of downtown Edmonton since 1967.

For the expansion and renovation of the monolithic concrete library, the design team was challenged to modernize its Brutaliststyle for the 1.2 million visitors it receives every year. Transforming the original concrete façade into a floating, streamlined structure would be no small task. Fast + Epp responded by completely replacing the existing façade with a geometrically complex structure at the north face of the building and incorporating a new lateral system for the building.

The geometrically intricate expansion pushed the design team to the limits of structural inventiveness. Building upgrades were further constrained by existing conditions and financial pressures, emphasized by the challenge to design around delicate areas like an existing theatre – while cantilevering two new floor expansions over 40 feet from the base of the building. To accomplish this feat, the team responded with an interwoven array of three steel trusses. For scale, two of these trusses were almost the length of a city block, while the other truss was over five stories in height.

Analyzing the existing structure, Fast + Epp determined that upgrades would be minimal if trusses were tied back to the building columns – one of the few structural elements with additional capacity. Several iterations were required to fine-tune the design, but solutions emerged through this process to address each constraint with efficiency and elegance.

To enhance visibility and brighten up spaces throughout the library, large openings were cut into the floors of the existing structure. This proved to be a distinctively intricate challenge, compounded by missing original structural drawings of several critical floors. Nevertheless, ingenuity between the limited information available, complex site investigations, and creative use of the existing structure made these openings possible.

An architectural feature unique to the building was the visually lightweight reading ramp.

This structure underwent several design iterations to achieve the architectural slenderness desired while also meeting serviceability requirements for vibration – allowing a space 58 STRUCTURE magazine

that simultaneously accommodated people’s movements and the need for undisturbed stillness for reading. To dampen the ramp’s vibration, Fast + Epp engaged all aspects of the new and existing structure, utilizing the truss, two floors of structure, and hangers from the new concrete beams integrated with the existing concrete.

The existing building’s lateral system, built in the 1960s, was grossly deficient by modern code standards. Fast + Epp introduced a completely new lateral system to the existing concrete structure to address this deficiency. The solution was deceptively simple: a collared connection. This detail made use of readily available materials, such as steel pipe and threaded rod, and was iterated to its final constructible design through a collaborative process with the design teams, City engineers, and the Contractor’s experience.

The collared connection enabled Fast + Epp to harness the additional capacity of the existing concrete columns and allowed for steel braces to keep spaces open and sightlines clear within the floors. A similar idea was used to anchor the steel trusses back to the existing structure. This detail also provided flexibility as the columns were heavily reinforced; anchoring into the existing column was simplified. The anchors could be drilled through the existing spiral ties with confinement provided by the collar.

A cornerstone of Fast + Epp’s design intent was minimizing existing structure and foundation upgrades to reduce construction costs.

This resulted in using steel as the material of choice, which was optimized to accommodate the high capacities needed while minimizing structural weight. This also meant that many structural elements served multiple functions.

For instance, the steel trusses supported the roof, two floors of concrete, the ramp, ribbons of glazing, and secondary components such as drainage systems.

Flexible solutions and constructability were prioritized to minimize labor. Steel connections were slotted or designed for field modification. Concrete connections

had collars or were detailed to minimize investigative work and conflicts with the existing reinforcing. With incomplete drawings, several of these connections became custom details determined concurrently with construction.

While connecting to the existing concrete characterized much of the design and construction strategies, the most significant challenge was engineering the installation of the steel trusses. This required close coordination with the steel suppliers and their engineers. Several truss members were cambered for final loading conditions, which were significantly heavier than the self-weight construction condition.

The steel installers, therefore, had to install curved beams to tight tolerances. Preloading the steel beams and regular construction monitoring was used to track the observed site conditions against modeled values and guide the installation process.

The revitalized library celebrated its grand opening on September 17, 2020. Serving as the main branch of the Edmonton Public Library, the Stanley A. Milner Library receives over 1.2 million visits each year. It is a multipurpose community hub, encompassing expanded reading and public spaces, a Makerspace, a high-tech theatre, a daycare, and more.■ Fast + Epp was an Award Winner for the Stanley A. Milner Library Renewal project in the 2020 Annual Excellence in Structural Engineering Awards Program in the Category – Forensic/Renovation/Retrofit/ Rehabilitation Structures over $20M.

JANUARY 2022

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