STRUCTURE magazine | March 2016

March 2016 Seismic

A Joint Publication of NCSEA | CASE | SEI

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STRUCTURE

42 EDITORIAL

7 Global Challenges for Structural Engineers

STRUCTURAL TESTING

36 Coring of Concrete Masonry Walls: Is it Necessary?

By Taka Kimura, P.E.

By Richard M. Bennett, Ph.D., P.E.

INFOCUS

STRUCTURAL ANALYSIS

9 Oil and Water

By Barry Arnold, P.E., S.E., SECB BUILDING BLOCKS

10 Rethinking Seismic Ductility

By Timothy W. Mays, Ph.D., P.E.

38 Challenging Issues When Conducting Nonlinear Seismic Analysis

By Monique Head, Ph.D., Rakesh Pathak, Ph.D., P.E., Susendar Muthukumar, Ph.D., P.E. and Kevin R. Mackie, Ph.D., P.E.

By Ronald O. Hamburger, S.E., SECB STRUCTURAL PERFORMANCE

18 Alternative Diaphragm Seismic Design Force Level of ASCE 7-16

HISTORIC STRUCTURES

64 Schuylkill Falls Chain Suspension Bridge (1809)

By Frank Griggs, Jr., D. Eng., P.E.

68 Advancing Technology

By Pedro Sifre, S.E.

CODE UPDATES

INSIGHTS

24 Changes in the ACI 318 Anchoring to Concrete Seismic Provisions

74 Fastener Corrosion

By Mersedeh Akhoondan, Ph.D. and Graham E.C. Bell, Ph.D., P.E.

By Richard T. Morgan, P.E.

28 Seismic Retrofit with Fiber Reinforced Polymers

By Sarah Witt and

Lincoln Square Expansion

By Cary Kopczynski, P.E., S.E. and Mark Whiteley, P.E., S.E. When complete in 2017, the Lincoln Square Expansion in Bellevue, Washington, will add two 450-foot towers, a four level retail podium, and six levels of subterranean parking.

With the use of Performance Based Design, the project is an excellent example of how innovative structural design can respond to high seismic requirements.

46

FEATURE

The Modern Temple

50

FEATURE

Innovation

SPOTLIGHT STRUCTURAL REHABILITATION

42

By Brendan Ramos, S.E. and David W. Cocke, S.E. The iconic and historic Masonic Temple in Glendale, California, appears to be a purely concrete building. Its exterior concrete walls mask the true skeleton of the structure: a steel-framed building. A strategic retrofit scheme was developed, voluntarily increasing the performance of the building.

CASE BUSINESS PRACTICES

By S. K. Ghosh, Ph.D.

March 2016 FEATURE

CODES AND STANDARDS

14 Seismic Design Value Maps

®

75 Design on Stage: Goldsmith Theater

By Jay Love, S.E. In 2007, Sutter Health brought together an Integrated Project Delivery (IPD) team to produce its new San Francisco flagship hospital, California Pacific Medical Center.

The Structural Engineer of Record was challenged to develop the highest performing seismic force resisting system that provided the greatest value.

By Samuel Mengelkoch, S.E. and David W. Cocke, S.E.

Greg Gilda, P.E., S.E. STRUCTURAL FORUM STRUCTURAL PRACTICES

32 Are You Communicating Seismic Concepts Correctly?

82 The Engineering Way of Thinking: Adaptation

By William M. Bulleit, Ph.D., P.E.

By Brent Maxfield, S.E. On the cover The new Sutter Health Hospital in San Francisco, currently under construction, incorporates viscous wall dampers into the seismic resisting system to reduce the effects of earthquakes. This is the first hospital in California to use this system, developed in Japan over 25 years ago.

Photo courtesy of Brett Drury. See feature article on page 50.

IN EVERY ISSUE 8 Advertiser Index 71 Resource Guide (Software Updates) 76 NCSEA News 78 SEI Structural Columns 80 CASE in Point STRUCTURE magazine

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57

FEATURE

2016 Looking Good for Steel-Related Companies

By Larry Kahaner The 2015 construction year was a robust one for steel-related companies. Looking ahead, 2016 could be even stronger.

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, C 3 Ink, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

March 2016

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Editorial

Global new trends, new Challenges techniques and current forindustry Structural issues Engineers

By Taka Kimura, P.E., M.ASCE, F.SEI

S

imilar to the industrial revolution of the early 19th century, we are in the midst of an information revolution. But rather than reducing the price and increasing the speed at which a cog can be manufactured, the current information revolution is reducing the price and increasing the speed at which services can be rendered. We live in a time where a variety of influences, including technological advances, growing affluence of the working class, and a changing geopolitical landscape, have created a world in which physical distance and geographic boundaries no longer pose barriers to international economic competition. Geographic distances are increasingly irrelevant when providing goods and services. One only has to call customer service for any major corporation to experience outsourcing first hand.

So how will this impact our profession as structural engineers? As technological advances and social shifts continue and broaden their impact on our work, the long term consequences are yet to be seen.

One thing is clear – globalization is a trend that impacts virtually every aspect of the way we perceive, pursue, and perform our work.

Global economies are increasingly intertwined and the rapid advance of technology facilitates collaboration across national boundaries. High quality, low cost technical expertise from abroad can be viewed as a challenge for our profession. This will be offset by more opportunities for all of us to work on innovative projects around the world and to provide our expertise and services to improve conditions in areas that have previously been underserved.

Accordingly, the future structural engineer must be prepared for this rapidly changing environment in order to thrive. He or she cannot depend solely on technical prowess for success. In addition to the technical skills that have been traditionally required of structural engineers, cultural awareness, knowledge of a foreign language, team building skills, and general adaptability will be essential traits of a globally minded worker.

Globalization was one of the key issues influencing our profession that was identified by the Structural Engineering Institute (SEI) Board of Governors during their Strategic Visioning efforts in 2011.

Subsequent to that meeting, SEI established the Task Committee on the Qualifications of Future Structural Engineers to study changes impacting the structural engineering profession, and to provide background and present ideas for change, as well as recommendations for action by the SEI Board of Governors. After careful consideration of the Task Committee’s report, the SEI Board of Governors unanimously voted to establish a new Global Activities Division in November 2015.

The mission of this newly established Division is to: • Increase SEI member’s awareness of global issues that impact our profession • Advance the role of SEI and its members globally • Facilitate the development of skills that allow SEI members to thrive in the world market • Serve as the communications mechanism for Global Chapters and members to express needs and make recommendations to the SEI Board of Governors.

• Support and participate in the global activities of the profession of structural engineering.

STRUCTURE magazine

ASCE and SEI leaders meet with representatives of The Institution of Structural Engineers, (IstructE) to help further the goal of international cooperation. Left to right: Jim Rossberg – Managing Director, Engineering Programs; Tom Smith – Executive Director, ASCE; David Odeh – President, SEI; Martin Powell – Chief Executive, IstructE; Mark Woodson – President, ASCE; Darren Byrne – Director Membership and Education, IstructE.

• Act as a global voice on behalf of structural engineers.

• Promote high quality services that are global in scope.

The intent is not to have the new Division act as a separate entity within SEI, but rather to integrate itself with the rest of SEI and its Divisions and to be a seamless part of what SEI is and how it operates. These are lofty goals for the new Global Activities Division, but the experienced and diverse members of its appointed Executive Committee are prepared for this challenge and held their first in-person meeting at the Geotechnical & Structural Engineering Congress in Phoenix, Arizona in February 2016.

In a word, what globalization presents to our profession is change.

Charles Darwin once said, “It is not the strongest of the species that survive, nor the most intelligent, but the one most responsive to change.” Rather than labeling globalization as a threat or an opportunity, what is important is to recognize, understand, and prepare for it.

Change is nothing new to structural engineering. From the constant evolution of the materials used in our designs, to the incorporation of various technologies that enable us to do our work with increasing efficiency, change has always been something structural engineers have recognized, embraced, and sought. The impact of globalization on our profession should be no different.

As the inaugural Chair of the SEI Global Activities Division, I look forward to the support of SEI and the broader structural engineering community as we attempt to prepare for the challenges and opportunities that globalization presents to our profession.

If you want to learn more about the SEI Global Activities Division and become involved, please contact Taka Kimura at taka.kimura@ch2m.com.▪ Taka Kimura is a Design Manager and Lead Structural Engineer at CH2M. Taka currently serves as the Global Activities Division Chair for the Structural Engineering Institute.

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March 2016

ADVERTISER INDEX

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KPFF Consulting Engineers .................... 8 Legacy Building Solutions ..................... 41 Lindapter .............................................. 61 LNA Solutions ...................................... 37 NCSEA ................................................. 13 New Millennium Building Systems ....... 59 Powers Fasteners, Inc. ............................ 35 Professional Publications, Inc. ............... 40 QuakeWrap ........................................... 29 RISA Technologies ................................ 84 S-Frame Software, Inc. ............................ 4 SidePlate Systems, Inc. .......................... 56 Simpson Strong-Tie......................... 17, 63 Steel Deck Institute ............................... 19 Structural Technologies ......................... 69 StructurePoint ....................................... 31 Super Stud Building Products, Inc......... 54 Taylor Devices, Inc. ............................... 65 Trimble ................................................... 3

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EDITORIAL BOARD Chair Barry K. Arnold, P.E., S.E., SECB ARW Engineers, Ogden, UT chair@structuremag.org John A. Dal Pino, S.E. Degenkolb Engineers, San Francisco, CA Mark W. Holmberg, P.E. Heath & Lineback Engineers, Inc., Marietta, GA Dilip Khatri, Ph.D., S.E. Khatri International Inc., Pasadena, CA Roger A. LaBoube, Ph.D., P.E. CCFSS, Rolla, MO Brian J. Leshko, P.E. HDR Engineering, Inc., Pittsburgh, PA Jessica Mandrick, P.E., LEED AP Gilsanz Murray Steficek, LLP, New York, NY Brian W. Miller Davis, CA

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Mike Mota, Ph.D., P.E. CRSI, Williamstown, NJ Evans Mountzouris, P.E. The DiSalvo Engineering Group, Ridgefield, CT Greg Schindler, P.E., S.E. KPFF Consulting Engineers, Seattle, WA Stephen P. Schneider, Ph.D., P.E., S.E. BergerABAM, Vancouver, WA John “Buddy” Showalter, P.E. American Wood Council, Leesburg, VA C3 Ink, Publishers A Division of Copper Creek Companies, Inc. 148 Vine St., Reedsburg WI 53959 Phone 608-524-1397 Fax 608-524-4432 publisher@structuremag.org

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STRUCTURE magazine

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March 2016

March 2016, Volume 23, Number 3 ISSN 1536-4283. Publications Agreement No. 40675118.

Owned by the National Council of Structural Engineers Associations and published in cooperation with CASE and SEI monthly by C3 Ink. The publication is distributed free of charge to members of NCSEA, CASE and SEI; the non-member subscription rate is $75/yr domestic; $40/yr student; $90/yr Canada; $60/yr Canadian student; $135/yr foreign; $90/yr foreign student.

For change of address or duplicate copies, contact your member organization(s) or email subscriptions@STRUCTUREmag.org. Note that if you do not notify your member organization, your address will revert back with their next database submittal. Any opinions expressed in STRUCTURE magazine are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, C3 Ink, or the STRUCTURE Editorial Board.

STRUCTURE® is a registered trademark of National Council of Structural Engineers Associations (NCSEA). Articles may not be reproduced in whole or in part without the written permission of the publisher.

InFocus

Oil new trends, and newWater techniques and current industry issues

By Barry Arnold, P.E., S.E., SECB

I

have never been concerned that three organizations represent structural engineers. So, while attending a recent engineering conference, when one of my tablemates spoke up in an agitated tone and said: “I don’t understand why there are three organizations representing structural engineers! They’re like oil and water – they don’t mix well,” I was taken aback. It is a question and a simile I have heard before, but his tone and abruptness caused me to pause and reflect.

ASCE/SEI (Structural Engineering Institute), NCSEA/SEA (National Council of Structural Engineers Associations), and ACEC/CASE (Council of American Structural Engineers), have each played an important role in my professional career, so I have never questioned why they exist. I just celebrate the fact that they do. In fact, asking me why there are three organizations representing structural engineers is like asking me to explain why each of us prefers a different style of home, why we prefer different breakfast cereals, why there is more than one fast food company, or why there are five electric hand-held drill manufacturers with sixty eight different kinds of drills on the market. (Hint: Each manufacturer and product offers unique features and benefits to the user.) Much like selecting the right drill(s) for a particular job, all three organizations have unique features and benefits that, for me, proved useful throughout my entire professional career.

I joined ASCE in 1989. The resources they had were overwhelming and exciting. I looked forward to receiving their catalog and filling my personal library with books that contained information and ideas that broadened my understanding of the profession. Because of its focus on structural engineering, when SEI was formed, I joined immediately. Although there are many benefits of membership, one I find particular meaningful now are Tara Hokes’ articles on ethics in Civil Engineering magazine, because they broaden my understanding of how to apply ethics in real-life situations.

When I accepted my first job, my employer informed me that I would become a member of SEAU (Structural Engineers Association of Utah). It was mandatory – no exceptions. Not only was I expected to be a member, but I was expected to participate by belonging to committees and attending their conferences, social events, and monthly meetings. It was not enough just to pay my dues – I was expected to contribute. My peers at SEAU provided a wealth of insights and practical knowledge and, when SEAU joined NCSEA, the opportunity to communicate with my peers from around the county provided exposure to a broad range of knowledge and perspectives that helped me improve technically and professionally.

My first exposure to ACEC occurred when I was a local host during their conference in Utah. The firm I work for has been a long-time member of ACEC and an adamant supporter of CASE. Although I was technically competent, my knowledge of leadership and managerial issues was lacking and prevented advancement. Graduating from the ACEC Leadership Institute in 2005 broadened my understanding and appreciation of non-technical, business-related topics. I still listen to the CD’s I purchased from my first ACEC conference fifteen years ago. They are dated, but a great refresher.

STRUCTURE magazine

I appreciate that SEI/NCSEA/CASE have a common goal – to improve the profession; and despite having the same goal, each organization takes different paths and focuses on different areas. Is there overlap? Of course, it is unavoidable. Each organization offers publications, webinars, and holds annual conferences accented with their unique perspective. Each organization has its place, its unique purpose, its value proposition, and its committed members.

Therefore, I dispute the notion that the SEI/NCSEA/CASE do not mix well. Because they serve different parts of the structural engineering community, they have different perspectives and disputes sometimes arise. But they also work together. The leadership meets yearly to discuss what their respective organizations are working on and where help and support is needed. Strategies and tactics are discussed and, if an overlap exists, decisions are made as to which organization should work on the problem moving forward. A good example of spirited cooperation is SELC (Structural Engineering Licensure Coalition), which was formed to unite all three organizations into one body to promote SE licensure. STRUCTURE magazine is another good example of cooperation between the organizations, to inform, promote, and improve the profession.

My other tablemate chimed in with her thoughts: “I can’t imagine anyone who is serious about their professional career and creating a business of lasting value not belonging to and actively supporting all three organizations.” I will second that. Instead of bemoaning what appear to be similarities between the organizations, I encourage you to dig deeper and celebrate their unique differences.

Barry Arnold (barrya@arwengineers.com) is a Vice President at ARW Engineers in Ogden, Utah. He chairs the STRUCTURE magazine Editorial Board, is the Immediate Past President of NCSEA, and is a member of the NCSEA Structural Licensure Committee.

9

March 2016

Building Blocks updates and information on structural materials

D

esign of auger cast and prestressed concrete pile elements supporting building structures in Seismic Design Categories A through F is covered in Chapter 18 of the 2015 International Building Code (IBC). It is anticipated that subsequent additions of the IBC will no longer contain these provisions and ACI 318 is currently being revised to include provisions for these elements beginning in 2019.

The committee in charge of foundation elements (ACI 318F) is already in the process of carefully studying the history of pile foundation detailing provisions and drafting language applicable to both foundation systems. The purpose of this article is to present shortcomings associated with the current prescriptive seismic design philosophy used for both auger cast piles and prestressed piles, to contrast this design approach with that used for other structures, and to do a side-by-side comparison of issues associated with these two commonly used reinforced concrete foundation elements.

Rethinking Seismic Ductility A Critical Comparison of Auger Cast and Prestressed Piles in Areas of Moderate to High Seismicity

By Timothy W. Mays, Ph.D., P.E.

Timothy W. Mays is a Professor of Civil Engineering at The Citadel in Charleston, SC. Dr. Mays previously served as Executive Director of the Structural Engineers Associations of South Carolina and North Carolina. He currently serves as NCSEA Publications Committee Chairman. Timothy can be reached at mayst1@citadel.edu.

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

Seismic Design Philosophy

Building piles are designed for lateral forces, axial forces, and bending moments that occur at the top of the pile (e.g., fixed head piles) and below grade, and are the result of significantly reduced earthquake forces caused by the inelastic behavior of the lateral force resisting system above the foundation. Regardless of whether an equivalent lateral force, modal, or elastic/inelastic time history analysis is used for the building itself, actions at the top of the piling, as provided by the structural engineer to the geotechnical engineer, are forces that are based on an anticipated elastic response of the piling. In harmony with this elastic design approach, the 2015 IBC has an allowable lateral load limit of fifty percent of the load causing a deflection of one inch at the top of the foundation element or the ground surface.

Standard practice is that the geotechnical engineer provides the structural engineer shear and bending moment diagrams for all piles that consider, as a minimum, the following: • An appropriate subgrade modulus for each soil layer (i.e., p-y springs), • Elastic response of the pile using an effective stiffness, and • Liquefied layers modeled using reduced p-y springs (if applicable).

In areas where liquefaction is likely to occur, the geotechnical engineer usually provides a maximum moment for each pile that is the larger of the results obtained including and neglecting liquefaction, respectively. The purpose for this requirement is twofold. First, although it may be expected, liquefaction may not occur and the foundation should be designed for this scenario.

Secondly, it is well known that liquefaction does

10 March 2016

not occur during the onset of seismic motion.

Rather, liquefaction may take place well after the maximum soil movements have taken place.

Special provisions do apply for Seismic Design Category (SDC) D and higher when the Site Class has been classified as E or F (IBC Section 1810.2.4.1). The pile systems must be designed to resist maximum earthquake induced pile curvatures resulting from the structure above (i.e., pile head loading) and free-field soil movements/ soil-structure interaction. Although this sounds cumbersome, standard practice regarding liquefied layers is already addressed in typical geotechnical reports and other concerns, such as neglecting certain soil support conditions (due to settlement at the top of the pile) and accounting for larger soil movements for soft soils, are easily addressed by the geotechnical engineer as part of the design and recommendation process. In lieu of this slightly more detailed analysis, prestressed piles and auger cast piles can be prescriptively detailed as an “assumed to meet measure” via an exception statement contained in IBC Section 1810.2.4.1.

It is the author’s opinion that the devil, as always, is in the details. The IBC philosophy of designing for an elastic response of the piling (see above) is discarded when it comes to pile detailing. Although the engineer is required to conservatively determine demands that suggest an elastic response, prescriptive detailing approaches presented elsewhere in Chapter 18 of the 2015 IBC are based on significant inelastic behavior of the pile system. It is as if the engineer is being told, “make sure the pile only moves 0.5 inches at the top but detail it to move 6 inches just in case.” Granted, the previous statement is purposely facetious but it illustrates the point. It is more important to recognize two facts: 1) The IBC approach of designing for an elastic response but detailing pile foundations to handle extreme inelastic behavior is not a surprise, nor unjustified. It is the downfall of prescriptive design. Given uncertainties in the geotechnical assumptions, the IBC committees have always been concerned that even though an elastic response may be expected, pile ductility might still be required. Uncertain how much ductility to require and where it might occur, the committees have continually required seismic ductility levels similar to those used for columns used as special reinforced concrete moment frames.

2) Pile designers must be made aware that the stringent detailing for pile design has never been required in the first place (i.e., performance based design has always been allowed under the current code provisions). Designers are permitted to base the reinforcement required on more advanced pile analysis procedures as used

in the design of other structures such as bridges, piers, and wharves, but these approaches have yet to catch on in the building industry.

Design procedures for prestressed concrete piling in areas of moderate to high seismicity vary significantly for bridge and building structures. Bridge foundations and substructures are usually detailed such that global earthquake forces are reduced by column, bent, and/or pile energy dissipation via the formation of plastic hinges in these elements. Design procedures for bridge foundations (Caltrans, SCDOT) and those for pier and wharf type structures (MOTEMS) are based on a performance based design procedure that allows the design professional to detail piles based on their anticipated level of inelastic behavior during the design earthquake. More specifically, the design professional provides an appropriate amount of spiral reinforcing to ensure that plastic hinges that develop in the pile are capable of adequate rotation, as required by a pushover or time history analysis.

on prestressed piles has actually justified the reduction when stiff soil is present and the fact that the displaced soil is under increased confining pressure (Budek et al., 1997). It is the author’s opinion that neither foundation type should get this reduction. Recent research, and common sense, show clearly that cyclic loads cause separation between the pile face and the supported soil in the ductile region of the pile (i.e., this confinement is not guaranteed when using prescriptive design).

2) The 2015 IBC does not mandate or recommend specific resistance factors for different

18˝ PCP

18˝ AC w/0.55 18˝ AC w/0.45

40 ft

55 ft

24˝ PCP

65 ft

30 ft

A comparison of typical required foundation sizes for the same demand based on example resistance factors. Note that AC stands for auger cast and PCP stands for prestressed concrete pile.

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March 2016

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Neglecting the fact that auger cast piles are not typically permitted by AASHTO (nor local DOT provisions) for support of bridge structures, the previous discussion regarding pile reinforcement detailing and seismic design philosophies is applicable to both auger cast and prestressed piles (i.e., drilled shafts and other cast in place piles are permitted by AASHTO and are somewhat similar to auger cast piles in their final form). However, there are many 2015 IBC provisions regarding the design of auger cast and prestressed piles that suggests that the two systems are being designed for very different analysis, design, detailing, and construction criteria. Some of the most critical points are summarized below: 1) Auger cast pile designers are permitted to reduce confinement steel 50% in Site Classes A through D for SDCs A, B, and C. This reduction is in addition to an already relaxed requirement as compared to prescriptive requirements for prestressed piles. The allowance for less confinement is based on the committee opinion that the soil surrounding the auger cast pile (in response to the form like grout pressure) is firm enough to act in place of the missing transverse steel. A similar provision is not included in the 2015 IBC for prestressed piles (which actually displace soil), even though experimental research performed

Table 1. Minimum lateral tie and spiral spacing requirements for auger cast piles (SDC C).

The term Dp denotes pile diameter.

Pile Location

Minimum lateral tie spacing

From bottom of pile cap to 3Dp

Minimum of: 6 inch center to center 8 longitudinal bar diameters

Remainder of pile length where longitudinal steel is present

16 longitudinal bar diameters

Table 2. Minimum lateral tie and spiral spacing requirements for auger cast piles (SDCs D, E, and F).

Pile Location and Condition

Minimum lateral tie spacing

From bottom of pile cap to 3Dp in Site Class A through D

Must satisfy the following ACI 318 Sections: 21.6.4.2, 21.6.4.3, 21.6.4.4 -Note that 21.6.4.4(a) transverse spiral reinforcement ratio may be reduced 50% - Note that ACI 318 Equation 10-5 need not apply per IBC Section 1810.3.2.1.2

From bottom of pile cap to 7Dp in Site Class E through F (also, from 7Dp at interfaces of hard or stiff soil layers to soil layers that are liquefiable or are composed or soft-to-medium stiff clay)

Must satisfy the following ACI 318 Sections: 21.6.4.2, 21.6.4.3, 21.6.4.4 - Note that ACI 318 Equation 10-5 need not apply per IBC Section 1810.3.2.1.2

Remainder of pile length where longitudinal steel is present

Minimum of: 12 inch center to center 12 longitudinal bar diameters 0.5Dp

pile types. The code does, however, refer to “approved” methods. It should be noted that AASHTO and state specific bridge standards present different resistance factors for each pile type. Specifically, and consistent with an LRFD philosophy, these codes uniformly assign larger resistance factors for prestressed piles as compared to cast in place piles and thereby assigns greater design capacity to the prestressed piles of the same size. An example is provided below.

Example Resistance Factors (typical of bridge construction): Driven Pile (with PDA): 0.85 Driven Pile (with Wave Equation): 0.75 Shafts with Single Test (IGM): 0.55 Shafts (other): 0.45 The 2015 IBC recognizes through reference of “approved” methods that the geotechnical reliability of auger cast and prestressed piles are different as a result of both construction and testing considerations (i.e., prestressed piles are tested piles by nature of installation techniques).

However, it is the author’s opinion that the language is not strong enough. A survey of geotechnical engineers practicing on the east coast suggests that common practice is to consider the geotechnical capacity of both foundation types as identical for the same size pile when doing

building reports, yet to conclude the prestressed piles are substantially stronger when writing bridge reports. A simple additional phrase such as “approved methods including an appropriate consideration of geotechnical resistance factors” eliminates this problem.

3) The 2015 IBC does not mandate or recommend specific subsidence demand factors for different pile types. The code does, however, refer to “approved” methods. It is well established in the literature that as a result of higher soil-pile friction stresses, auger cast piles are subjected to significantly higher load effects from soil subsidence than are prestressed piles of the same size. Modification for this effect is not as easy to resolve as item 2 above, but it is the author’s opinion that geotechnical engineers should be made aware of this discrepancy.

4) Placement of longitudinal steel and rebar cages in semifluid grout is one of the biggest issues with auger cast piles. Experience has shown that there is no way to ensure that rebar cages of any reasonable length are set in the grout column in accordance with the construction documents. Durability and bending strength can be severely compromised when the cage is not set with the level of precision specified elsewhere in ACI 318-11 for other

STRUCTURE magazine

12

March 2016

types of reinforced concrete members. When the cracking moment is exceeded, a reinforced cage should always be used since the placement of one vertical bar is even more prone to misplacement and its impact on bending strength is negligible. Also, the designer should note that the 2015 IBC does not (but likely should) require that the longitudinal reinforcement that extends to the location of the cracking moment extend one development length beyond this location. This detailing methodology is inconsistent with traditional ACI 318 design approaches for moment resistance in reinforced concrete members.

The most critical discrepancy between auger cast and prestressed piles actually ties back to the initial discussion on seismic design and reinforcing requirements. Most notable is the required amount of spiral for both pile types considering Seismic Design Category (SDC) C and SDCs D through F. The amount of spiral required in the 2015 IBC for each pile type is summarized below.

Spiral Requirements for Auger Cast Piles – SDC C No specific volumetric ratio of spiral is required.

However, the minimum permitted spiral diameter is 3/8 inches and the minimum spacing of the transverse steel is specified in Table 1.

Spiral Requirements for Auger Cast Piles – SDC D through SDC F The volumetric spiral ratio ρs = volume of spiral/volume of core (measured out-to-out of spiral) must equal or exceed the following:

ρs,min = 0.12

fc ' fyh

where fyh is the yield stress of the spiral reinforcement. Minimum spacing of the transverse steel is specified in Table 2.

Spiral Requirements for Prestressed Piles – SDC C Sufficient transverse reinforcement must be provided in the upper 20 feet of the pile length such that the spiral reinforcement index exceeds the following:

ρs,min = 0.12

fc ' fyh

where fyh is the yield stress of the spiral reinforcement not to be taken as larger than 85,000 psi. For the remaining length of the pile, half the minimum transverse reinforcement specified above must be provided.

Spiral Requirements for Prestressed Piles – SDC D through SDC F In the pile’s ductile region, which includes up to the top 35 feet of the pile length, the spiral reinforcement index must equal or exceed the following:

ρs,min = 0.25 0.12

(

(

)(

)

fc ' Ag P –1 0.5+1.4 ≥ max fyh Ach fc 'Ag

fc ' P 0.5+1.4 fyh fc 'Ag

) and 0.12 ff ' c

yh

where Ach is the cross-sectional area of the confined core (measured out-to-out of spiral), Ag is the gross cross-sectional area of the pile, and P is the factored compressive load on the pile using either IBC Equation 16-5 or 16-7 as applicable. Note that the minimum spiral reinforcement index need not be taken as greater than 0.021 (i.e., ρs,min ≤ 0.021).

Comparison of Spiral Requirements A quick comparison of the spiral requirements presented above manifests that the quantity

of spiral required and the length of pile over which this spiral is required is much greater for prestressed piles as compared to auger cast piles. For example, why is the SDC D volumetric ratio required for auger cast spiral equal to that required for prestressed piles in SDC C? Also, why is the expected length of curvature ductility demand so much greater for prestressed piles when no soil structure interaction model comparing the two pile types would suggest this would be the case?

The spiral requirements for prestressed piles are based on extensive flexural ductility tests performed on actual piles and conclusions made in the Recommended Practice for Design, Manufacture and Installation of Prestressed Concrete Piling (1993). It is unclear to this author why auger cast piles have such relaxed prescriptive spiral requirements.

Th e Recommended Practice for Design, Manufacture and Installation of Prestressed Concrete Piling (1993) is undergoing major revisions at the time of this writing. However, it is important to note that new research performed for PCI (Fanous et al., 2010) is being used to justify that even more spiral should be used in SDCs C through F. PCI has established required curvature ductility demands of 12 and 18 in areas of moderate and high

seismicity, respectively, as target values for design, and will also encourage the designer to use performance based design to justify more accurate quantities when advanced soil structure interaction modeling is included as part of the design process.

It is the author’s recommendation that unless performance based design methodologies are used, both pile types have the same length of pile segments detailed for the same prescriptive ductility capacities so that both piles can be assumed to provide the same level of safety in response to the design earthquake.

Acknowledgements The author wishes to thank the Precast/ Prestressed Concrete Institute (PCI) for partially funding this study comparing driven prestressed concrete piles to augercast piling. During this last year of study, the questions and input from the committee members of PCI has not only enhanced the knowledge of those that participated but also allowed this author to coalesce a better understanding for how to approach the need for legacy methods and performance-based seismic provisions here in my home state of South Carolina.▪

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Codes and standards updates and discussions related to codes and standards

O

ver the past twenty years, the seismic design value maps referenced by the building codes have undergone revolutionary changes, affecting the information they portray, the way they are developed and the design procedures that reference them. Many structural engineers, noting these constant changes and the effects on their designs, question why this happens and if it is necessary.

This article presents a historic review of major developments in seismic design value mapping and a look forward to potential future changes.

Prior to 1993, U.S. building codes adopted seismic maps that portrayed design values in the form of seismic zones (Figure 1). The five seismic zones, each of which covered broad regions of the U.S., were based on the historic occurrence of damaging earthquakes, or the lack thereof; Zone 0 representing places where earthquakes had never been experienced, and were therefore deemed unlikely to occur; Zone 4 where major damaging earthquakes had historically occurred one or more times; and the intermediate zones covering places where ground shaking may occasionally have been felt, but where damage due to this shaking seemed less likely.

Originally, seismic zones and design forces were not quantitatively tied to anticipated ground accelerations. In the early 1900s, engineers in Italy, Japan and the U.S., without specific knowledge of the accelerations produced by earthquakes or the dynamic response of structures subjected to ground shaking, concluded that structures should be seismically designed to withstand 10% of their weight, applied as a lateral force should provide adequate protection. At that time, only buildings in places known to have severe earthquakes, including Los Angeles, Tokyo, and San Francisco, were designed for such forces. Later as building codes evolved, code-writers specified general use of the 10% criteria for structures in zones of severe seismic risk, and arbitrary fractions of this requirement for zones of lesser seismic risk. Still later, as engineers began to understand structural dynamics and spectral response, the 10% rule was adjusted to account for the reduced response of more flexible structures.

At the same time, engineers observed that some types of structures, for example, unreinforced masonry bearing wall buildings, performed more poorly in earthquakes than others, e.g., frame structures. In response, codes specified structural quality factors to adjust the required design forces based on the type of structural system and the perceived enhanced capability of some structural systems to provide better performance. Finally, in the 1970s, code writers

Seismic Design Value Maps Past, Present and Future

By Ronald O. Hamburger, S.E., SECB

Ronald O. Hamburger (rohamburger@sgh.com) is a senior principal at Simpson Gumpertz & Heger in San Francisco. He has been active in the development of seismic requirements of building codes and standards since the 1980s.

He presently chairs the ASCE 7 Committee and also the Project 17 Committee discussed in this article.

14 March 2016

concluded that some structures, by reason of their occupancy and use, were more important than others and should be designed stronger, resulting in introduction of an occupancy importance factor. Throughout the 1970s and 1980s, seismic design forces were determined by the formula: V = ZIKCSW where Z is a zone factor, having a value ranging from 0 (in Zone 0) to 1 (in Zone 4); I, the occupancy importance factor, still with us today; K, a structural quality factor varying from 0.67 for structures having complete vertical load carrying space frames to 1.33 for “box” type structures wherein the shear walls carried most of the structure’s weight W; and C, a period (or height) dependent force coefficient, specified to result in a design forces equal to 10% of the structure’s weight for rigid structures of ordinary occupancy and ordinary framing located in Zone 4; and S a site factor, which increased design forces on soft soil sites. A nearly identical version of this formula was in use, without the occupancy importance and site factors, since the 1940s.

Following the 1971 San Fernando earthquake and observation that some code conforming structures performed poorly, C was adjusted to provide an “ordinary” design force of 13% of the structure’s weight, and then following similar observations in the 1994 Northridge earthquake, 18% of the structure’s weight. Seismic design value maps continued to specify the Z coefficient, and geographic boundary of seismic zones based on the historic occurrence of earthquakes in broad regions. There was relatively little science underlying these requirements. Engineers determined design forces in an imprecise, but simple, way and the design forces in a given zone remained constant from code cycle to cycle, unless an extraordinary event, like a major damaging earthquake, suggested that force levels should be raised.

In the 1970s, the ATC 3-06 project initiated scientific quantification of mapped design values, abandoning seismic zone maps and adopting spectral response acceleration maps in their place. The ATC 3-06 report declared that design earthquake ground motions represented events with a 475-year mean return period and that the Zone 4 design motions had effective peak ground accelerations of 0.4g, a broad generalization given the huge variation in actual seismic risk across Zone 4. The ATC 3-06 report also recommended separate maps for short period and 1-second response accelerations, respectively denoted Aa and Av. Despite these conceptual differences, the ATC 3-06 maps resembled and retained much of the coarseness of the original seismic zones (Figure 2). Building codes adopted the ATC

2A

3

2B

1

0

2A

4 3

0 4 2B

1

2A

1

1

2A

4

1

2A 2B

2A 2A

1

Map Area Coeff Aa

2B 0

Figure 1. 1988 UBC Seismic Zone Map.

Figure 2. Aa map from ATC 3-06.

3-06 recommendations slowly, retaining seismic zone maps into the 1990s. The codes did, however, adopt the concept that design ground motion represented 475-year shaking, and that design ground motions in Zone 4 had effective peak ground acceleration of 0.4g.

Paralleling the approach recommended

by ATC 3-06, the 1988 Uniform Building Code (UBC) abandoned the base shear force formula in use since the 1940s in favor of a form similar to that found in present building codes. The Z and S parameters were dropped from the base shear equation but used to compute spectral response coefficients Ca and Cv, which replaced them. The structural quality factor, K, was abandoned in favor of the response modification coefficient Rw. Conceptually Rw represented the level of ductility, overstrength and damping inherent in different structural systems and the judgmentally determined ability of such structural systems to safely exhibit inelastic response. However, the Rw values were calibrated so that required design forces for different structural systems remained essentially unchanged from that required by earlier codes. Seismic zones were retained, with the Ca and Cv values derived over broad regions from the seismic zone factor and also the site characteristics. The base shear formula became more scientific, as did the commentary describing its basis, but design force levels changed little.

Starting in 1985, the Building Seismic Safety Council (BSSC) began to publish the NEHRP Recommended Provisions for Seismic Regulation of Buildings based on the ATC 3-06 report. This included a base shear formula like that contained in the 1988 UBC except that Rw was replaced with R to provide strength-level rather than allowable stress design level forces, and a larger number of structural systems, with

unique R values, were defined. Unlike the 1988 UBC, the NEHRP Recommended Provisions included separate maps for Aa and Av. In 1993 both the BOCA and Standard Building Codes, used throughout the eastern U.S., adopted seismic design criteria and maps based on the NEHRP Recommended Provisions while the Uniform Building Code (UBC) retained the earlier criteria developed and maintained by the western states structural engineering associations, most notably SEAOC and SEAW. However, by 1994 it was clear that the three model code development organizations should collaborate to publish a single nationally applicable code, and that seismic requirements would be based on the NEHRP Recommended Provisions. Preparing for this merger, the Federal Emergency Management Agency hosted a series of joint Building Seismic Safety Council (BSSC) / United States Geologic Survey (USGS) projects to develop updated seismic design

Figure 3. Ss map for the Western U.S. adopted in ASCE 7-05.

STRUCTURE magazine

15

March 2016

7 6 5 4 3 2 1

0.40 0.30 0.20 0.15 0.10 0.05 0.05

value maps for the new code. SEAOC and SEAW representatives also participated in these projects.

A key consideration in these joint projects was that 475-year return periods, while sufficient to capture most events likely to occur in the Western U.S. were not sufficient to capture credible ground shaking events in the eastern U.S. To avoid a major earthquake catastrophe in the eastern U.S., from a repeat of the 1811-1812 New Madrid earthquakes or the 1898 Charleston earthquake, it was felt that return periods on the order of a few thousand years were needed. A first attempt to develop nationally applicable seismic design maps and procedures, termed the Seismic Design Values Panel, failed to reach consensus.

Ultimately, the second joint USGS/BSSC project (Project 97) developed a compromise solution resorting to 2,475 year return periods throughout the U.S., but capping the mapped values, deemed excessively high in regions near major active faults, with deterministic estimates of maximum likely ground shaking. Project 97 also introduced the concept of separately mapping short period and 1-second period response accelerations, now termed Ss and S1, in the form of contours, and developed the design procedure keyed to these values that still underlies ASCE 7 and the IBC today. The resulting maps appeared in ASCE 7-98, ASCE 7-02 and were referenced by the 2000 and 2003 editions of the International Building Code (IBC), then with slight revisions, were updated, republished and referenced by ASCE 7-05, and the 2006 and 2009 IBC editions (Figure 3).

Significantly, the new maps showing contours for the Ss and S1 values were based on probabilistic seismic hazard analysis for every point on a 2 km by 2 km grid (approximately 1.2 miles by 1.2 miles) across the United continued on page 17

upon which the maps are based and dissatisfaction with the ever-changing design requirements for buildings. Further, as the definition of the maps has become more complex, designers have lost an understanding of the intent of the map, and what they represent. Importantly, the maps portray precision in the design values that is inappropriate, given the substantial uncertainty in the values portrayed. As a result of these concerns, the ASCE 7-16 committee was reluctant to adopt the latest edition maps.

Recognizing these concerns, FEMA has sponsored a new joint USGS/BSSC project, termed Project 17, to once again evaluate the basis for the seismic design value maps referenced by the building codes. In addition to issues associated with updating the scientific basis for the maps, Project 17 will also address the issues of mapped value stability and portrayal of inappropriate levels of precision. The resulting maps produced by this effort will be referenceable by ASCE 7-22 and the 2024 edition of the IBC. Project 17 initiated in 2015 with a planning eff ort that included educational webinars and outreach to the structural and geotechnical engineering professions.

The resulting work plan includes a twoyear effort that will focus on: 1) balancing the precision of mapped values against the uncertainty in their determination, with an intent to provide stability in future codeadopted maps, 2) the appropriate risk (or hazard level) to use as the map basis, 3) more accurate determination of spectral values for soft soil sites and long period structures, and 4) if still required, how to set deterministic caps on design values in regions close to major active faults. Other issues, including characterization of basin effects on spectra shape and values, consideration of strong shaking duration, and the use of simulated rather than recorded events as a basis for the maps, were considered; but they were deemed beyond the project’s scope given the available resources. If additional resources or data become available, the project may address these.

The intent is for Project 17 to provide direction to USGS on generation of the next issue of design value maps by late 2017. This will enable USGS to provide maps to BSSC for adoption into the NEHRP Recommended Provisions in time to support the development of ASCE 7-22 and for inclusion in the 2024 IBC. In the interim, ASCE 7-16 and the 2018 and 2021 editions of the IBC will incorporate seismic maps based on the USGS 2014 seismic hazard database.▪

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States. These hazard analyses used the best available scientific knowledge to assess the values of these accelerations for a 2,475-year return period. Contours varied by 0.05g, and in regions close to major active faults were closely spaced such that, over a distance of a few kilometers, specified design values varied significantly. Not stated, nor generally understood by structural engineers, was the fact that the mapped values were inherently uncertain. While the mapped contour values represented “best” or mean estimates of the hazard at the 2,475-year return period, the statistical calculations underlying their derivation produced coefficients of variation of 60% or more. This uncertainty was a result of several factors, including a lack of knowledge as to the true locations of faults, the magnitudes of potential earthquakes on those faults and their activity rates, as well as unpredictable randomness in the prediction of ground acceleration values as a function of magnitude and distance. Scientific opinion as to the true value of these factors changed frequently and as these opinions changed, so too did the mapped values of ground shaking adopted by the building codes, such that with each successive building code edition, mapped values have alternatively increased and decreased, often without any real improvement in the certainty with which they have been calculated. As the mapped values of ground shaking parameters have changed, so have the required design forces for buildings, sometimes significantly from one code edition to another.

In 2007, FEMA partnered with BSSC and USGS once again to conduct a review of the basis for the seismic hazard maps and determine if updates to this basis were appropriate. Project 07, as the effort was called, resulted in conversion of the maps from a uniform hazard (2,475-year return period) for ground shaking to a uniform risk (1% annual probability of collapse) basis.

Further, Project 07 decided that rather than using geomean motion as the basis for the maps, maximum direction motions should be used. The resulting revised maps appeared in ASCE 7-10, IBC-2012 and the pending IBC-2015. In 2014, USGS made revisions to the seismic hazard model underlying the maps, employing updated science, and published a new series of maps for reference

by ASCE 7-16 and IBC 2018. Again, the specified design values in some locations changed significantly.

Noting the fluctuation in specified design values that occurs from code edition to code edition, structural engineers have expressed disbelief in the validity of the science

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Structural Performance performance issues relative to extreme events

T

he next edition of ASCE 7 Minimum Design Loads for Buildings and Other Structures, ASCE 7-16 (ASCE, 2016), is expected to be published in September 2016, in time for adoption into the 2018 International Building Code (IBC) (ICC, 2018).

For that edition, ASCE 7-10 (ASCE, 2010) has been modified to include a new Section 12.10.3, Alternative Design Provisions for Diaphragms including Chords and Collectors, within Section 12.10, Diaphragm Chords and Collectors. The new section provides for an alternative determination of diaphragm design force level, which is mandatory for precast concrete diaphragms in buildings assigned to SDC C, D, E, or F. The alternative is permitted to be used for other precast concrete diaphragms, cast-in-place concrete diaphragms, and wood sheathed diaphragms on wood framing. Section 12.10.3 does not apply to steel deck diaphragms. ASCE 7-10 has also been modified to add a Section 14.2.4, containing detailed seismic design provisions for precast concrete diaphragms including a connector qualification protocol.

Chapter 14 of ASCE 7-10 is currently not adopted by the 2012 or 2015 IBC (ICC, 2012, 2015). Steps directed towards the inclusion of ASCE 7-16 Section 14.2.4 in the 2018 IBC are now being taken.

Both changes originated in the 2015 NEHRP Recommended Provisions for Buildings and Other Structures (FEMA, 2015). This article is devoted to a discussion of ASCE 7-16 Section 12.10.3, Alternative Design Provisions for Diaphragms including Chords and Collectors.

Alternative Diaphragm Seismic Design Force Level of ASCE 7-16

By S. K. Ghosh, Ph.D.

S. K. Ghosh is President at S. K. Ghosh Associates Inc., Palatine, IL and Aliso Viejo, CA. He chaired Issue Team 6 on Diaphragms of the Building Seismic Safety Council Provisions Update Committee for the 2015 NEHRP Provisions. He can be reached at skghoshinc@gmail.com.

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

Basis and Overview The seismic design of structures has long been based on an approximation of the inelastic response of the seismic force-resisting system.

The approximation reduces the results of an elastic analysis in consideration of the reserve strength, ductility, and energy dissipation inherent in the vertical elements of the seismic force-resisting system. In 1978, ATC-3 (ATC, 1978) provided design force reduction factors based on consideration of inelastic behavior of the vertical elements of the seismic force-resisting system and the performance of structures in past earthquakes. The primary assumption leading to these factors is that yielding in the vertical elements of the seismic force-resisting system is the primary mechanism for inelastic behavior and energy dissipation. Starting with the 1997 Uniform Building Code (UBC) (ICBO, 1997), the actual forces and displacements that might occur in the vertical elements in a design-level seismic event were

18 March 2016

recognized with the introduction of a seismic force amplification factors, Ω0, and deflection amplification factor, Cd, respectively. In contrast, the design requirements for the horizontal elements of the lateral force-resisting system (the diaphragms) have been established by empirical considerations related to anticipated behavior of the vertical elements, rather than explicitly considering behavior of the diaphragms. For established diaphragm construction types, this empirical approach has been generally satisfactory. Satisfactory system performance, however, requires that the diaphragms have sufficient strength and ductility to mobilize the inelastic behavior of the vertical elements.

In order to help achieve the intended seismic performance of structures, the designs of horizontal and vertical elements of the seismic force-resisting system need to be made more consistent. Analytical results, as well as experimental results from shaketable tests in Japan, Mexico, and the United States, have shown that diaphragm forces over much of the height of the structure actually experienced in the design-level earthquake may at times be significantly greater than code-level diaphragm design forces, particularly where diaphragm response is near-elastic. Overstrength and ductility of the diaphragm, however, may account for satisfactory diaphragm performance. ASCE 7-16 Section 12.10.3 ties the design of diaphragms to levels of force and deformation capacity that represent actual anticipated behavior.

ASCE 7-16 Section 12.10.3 presents a near-elastic diaphragm force as the statistical sum of first mode effect and higher mode effects (Rodriguez et al., 2002). The first mode effect is reduced by the R-factor of the seismic force-resisting system, but then amplified by the overstrength factor, Ω0, because vertical element overstrength will generate higher first mode forces in the diaphragm.

The effect caused by higher mode response is not reduced. In recognition of the deformation capacity and overstrength of the diaphragm, the elastic diaphragm force from the first and higher modes of response is then reduced by a diaphragm force reduction factor, Rs. With the modification by Rs, the proposed design force level may not be significantly different from the diaphragm design force level of ASCE 7-16 Sections 12.10.1 and 12.10.2 for many practical cases. For some types of diaphragms and for some locations within structures, the proposed diaphragm design forces will change significantly, resulting in noticeable changes to resulting construction. Based on data from testing and analysis, and on building performance observations, it is believed that these changes are warranted.

The alternative design force level of Section 12.10.3 is based on work by Rodriguez, Restrepo, and Carr (Rodriguez et al., 2002), verified by more recent work by Fleischman et al. (Pankow, 2014).

Table 1. Diaphragm Design Force Reduction factor, Rs

ShearControlleda

FlexureControlleda

-

1.5

2

EDO 1, b

0.7

0.7

BDO 2, b

1.0

1.0

RDO

1.4

1.4

3.0

NA

Diaphragm System Cast-in-place concrete designed in accordance with ASCE 7-16 Section 14.2 and ACI 318 Precast concrete designed in accordance with ASCE 7-16 Section 14.2.4 and ACI 318 Wood sheathed designed in accordance with ASCE 7-16 Section 14.5 and AF&PA (now AWC) Special Design Provisions for Wind and Seismic

3, b

-

EDO is precast concrete diaphragm Elastic Design Option.

BDO is precast concrete diaphragm Basic Design Option.

3 RDO is precast concrete diaphragm Reduced Design Option. a Flexure-controlled and Shear-controlled diaphragms are defined in ASCE 7-16 Section 11.2. b Elastic, basic, and reduced design options are defined in ASCE 7-16 Section 14.2.4.

1 2

Current Diaphragm Seismic Design Force Level

Diaphragm Design Force at Any Level

Current seismic design forces for diaphragms in ASCE 7-10, to be retained in ASCE 7-16, are a function of the design forces acting on the vertical elements and are, therefore, reduced by the R-factor that applies to the vertical elements. Upper- and Lower-bound limits on the forces are added, as shown in Equation 1.

The alternative diaphragm design force is given in Equations 2 and 3.

∑Fi 0.2SDSI ew px ≤ Fpx

i=x n

wpx ≤ 0.4SDSI ew px

Equation 1

∑wi i=x

Alternative Diaphragm Seismic Design Force Level The alternative diaphragm seismic design force level is for buildings in which response of the vertical elements of the seismic force-resisting system dominates the overall structure behavior. It is not meant for buildings in which the seismic response is dominated by the diaphragms (as can occur in big-box buildings with flexible diaphragms).

The latter buildings are treated in FEMA P-1026 (FEMA, 2015a).

STRUCTURE magazine

≥ 0.2SDSIewpx

Equation 3

where: Cpx is the diaphragm design acceleration coefficient at Level x.

Rs is the diaphragm design force reduction factor.

Cp0 is the diaphragm design acceleration coefficient at the structure base.

Cpn is the diaphragm design acceleration coefficient at the top of the structure.

Rs is given in Table 1.

The definitions of flexure-controlled and shear-controlled diaphragms in ASCE 7-16 Section 11.2 are as follows: Flexure-Controlled Diaphragm: Diaphragm with a flexural yielding mechanism, which limits the maximum forces that develop in the diaphragm, and having a design shear strength or factored nominal shear capacity greater than the shear corresponding to the nominal flexural strength.

Shear-Controlled Diaphragm: Diaphragm that does not meet the requirements of a flexure-controlled diaphragm.

The precast concrete diaphragm design options are defined in ASCE 7-16 Section 14.2.4 as follows: Basic Design Option (BDO): An option where elastic diaphragm response in the design earthquake is targeted. continued on next page NEW

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where Fpx is the diaphragm design force at Level x.

Fi is the portion of the seismic base shear, V, induced at Level i. wi is the weight tributary to Level i. wpx is the weight tributary to the diaphragm at Level x.

SDS is the design spectral response acceleration parameter at short periods.

Ie is the seismic importance factor.

This empirical seismic design force level has generally resulted in satisfactory performance of diaphragms in earthquakes, as evidenced by a lack of observed severe damage. Material-specific factors related to diaphragm over-strength and deformation capacity may account for the adequate performance.

Results using analysis tools not available when the empirical rules were first established, indicate that the level of force required for design of diaphragms in new code-compliant buildings may not ensure development of inelastic mechanisms in the vertical elements of the seismic force-resisting system. This is particularly true where diaphragms have limited ductility and displacement capacity, as dramatically illustrated by the response of several shear wall structures during the Northridge Earthquake.

Equation 2

engineering manual

n

Fpx = Cpx wpx Rs

Elastic Design Option (EDO): An option where elastic diaphragm response in the maximum considered earthquake is targeted.

Reduced Design Option (RDO): An option that permits limited diaphragm yielding in the design earthquake.

Distribution of Diaphragm Design Force along Height As per Equation 2, the distribution of diaphragm design forces over the height of the structure varies as a function of Cpx. For structures of three or more stories, Cpx varies linearly between Cp0 at the base and Cpi at 80% of structural height, hn, above the base.

It then varies linearly between Cpi at 0.8hn to Cpn at hn, as shown on the right hand portion of Figure 1. For one and two stories, Cpx varies between Cp0 and Cpn, as shown in the left had portion of Figure 1.

In order to determine Cpx, it is necessary to first determine Cp0, Cpi, and Cpn. Cp0 = 0.4SDSIewpx

Figure 1. Floor acceleration envelopes for calculating the design acceleration coefficient Cpx in buildings with n ≤ 2 and in buildings with n ≥ 3.

Equation 4

Cpi is the greater of values given by: Cpi = Cp0

Equation 5

Cpi = 0.9Γm1Ω0Cs

Equation 6

Table 2. Modal Contribution Coefficient Modifier, zs.

Description

where: Γm1 is first mode contribution factor Γm1 = 1 + 0.5zs (1– 1 N

)

Equation 7

zs = modal contribution coefficient modifier dependent on seismic force-resisting system, as given in Table 2.

Cpn = √(Γm1Ω0Cs)2 + (Γm2Cs2)2 ≥ Cpi Equation 8

zs–value

Buildings designed with Buckling Restrained Braced Frame systems defined in ASCE 7-16 Table 12.2-1

0.30

Buildings designed with Moment-Resisting Frame systems defined in ASCE 7-16 Table 12.2-1

0.70

Buildings designed with Dual Systems defined in ASCE 7-16 Table 12.2-1 with Special or Intermediate Moment Frames capable of resisting at least 25% of the prescribed seismic forces

0.85

Buildings designed with all other seismic force-resisting systems

1.00

where: Γm2 is higher mode contribution factor Γm2 = 0.9zs (1– 1 N

)

2

Equation 9

Cs2 is higher mode seismic response coefficient. Cs2 is the smallest of values given by Cs2 = (0.15N + 0.25)Ie SDS Cs2 = Ie SDS Cs2 =

Equation 11

Ie SDS 0.03(N – 1)

Cs2 = 0

Equation 10

For N ≥ 2 Equation 12 For N = 1 Equation 13

Validation To validate the alternative seismic design force level, coefficients Cpx were calculated for various buildings tested on a shake table. Figures 2 and 3 plot the floor acceleration envelopes

Figure 2. Comparison of measured floor accelerations and accelerations predicted by Equations 2 and 3 for a 7-story bearing wall building (Panagiotou et al., 2011).

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2 and 3, normalized to accelerations at the structure base. The proposed design equations predict the vertical distribution of accelerations in both the uppermost part of the building and in the lowest levels reasonably well.

Comparisons of Design Force Levels Comparisons of diaphragm seismic design force levels along the heights of a number of buildings of various materials and assigned to various SDCs have been made. A few representative cases are shown below. For more comparisons, see FEMA P-1051 (FEMA, 2016).

4-Story Perimeter Wall Precast Concrete Parking Structure in Knoxville, TN

Figure 3. Comparison of measured floor accelerations and accelerations predicted by Equations 2 and 3 for a 5-story special MRF building (Chen et al., 2013).

Figure 4. Comparison of measured floor accelerations with accelerations predicted by Equations 2 and 3 for steel BRBF and special MRF buildings (adapted from Choi et al., 2008).

and the floor accelerations predicted from Equations 2 and 3 with Rs = 1 for two buildings built at full-scale and tested on a shake table (Panagiotou et al., 2011; Chen et al., 2013). Cp0 is defined as the diaphragm design acceleration coefficient at the structure base, and Cpx is defined as the diaphragm design acceleration coefficient at Level x. The vertical distribution of measured floor accelerations is reasonably predicted by Equations 2 and 3.

Research work by Choi et al. (2008) concluded that buckling-restrained braced frames are very effective in limiting floor accelerations

in buildings arising from higher-mode effects.

This finding is reflected in the provisions of ASE 7-16 Section 12.10.3, where the mode shape factor zs has been made the smallest for buckling-restrained braced frame systems.

Figure 4 compares vertical distributions of average floor accelerations obtained from the nonlinear time-history analysis of four buildings (two steel buckling-restrained braced frame systems and two steel special moment frame systems) when subjected to an ensemble of spectrum-compatible earthquakes with floor accelerations computed from Equations

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The structure for this example is a 4-story perimeter shear wall precast concrete parking garage. The lateral force-resisting system (LFRS) in the transverse direction is composed of four perimeter precast walls, two at each end of the structure. The LFRS in the longitudinal direction consists of interior lite walls flanking the central ramp. The SDC is C. The comparison of diaphragm design force levels for the structure by ASCE 7-16 Sections 12.10.1 and 12.10.2 (marked ASCE 7), by ASCE 7-16 Section 12.10.3 (marked ASCE 7 Alt.), and by the 2015 NEHRP Provisions (labeled NEHRP), are illustrated in Figure 5 (page 22). EDO, BDO, and RDO in the figure stand for Elastic, Basic, and Reduced Design Options, respectively. A few changes were made to alternative design force level provisions of the 2015 NEHRP Provisions before they were adopted into ASCE 7-16.

Those differences are not important for the purposes of this brief article.

Steel-Framed Office Structure in Seattle, WA The structure for this example is a 12-story buckling-restrained braced frame office building in Seattle, WA. The SDC is D. The comparison of diaphragm design force levels for the structure by ASCE 7-16 Sections 12.10.1 and 12.10.2 (marked ASCE 7), by ASCE 7-16 Section 12.10.3 (marked ASCE 7 Alt., and by the 2015 NEHRP Provisions (labeled NEHRP), are illustrated in Figure 6 (page 22). All three sets of requirements produce the same diaphragm design forces through most of the height of the structure, because the minimum diaphragm design force controls, except that ASCE 7-16 Sections 12.10.1 and 12.10.2 produce slightly higher than minimum diaphragm design forces at and near the very top. continued on next page

4-Story Shear Wall Parking Garage Knoxville (SDC C)

Figure 5. Design force level comparisons for precast concrete parking structure. (All references to ASCE 7 and NEHRP are to ASCE 7-16 and the 2015 NEHRP Provisions, respectively) [Reproduced with permission from FEMA from the upcoming FEMA P-1051].

Cast-in-Place Concrete Shear Wall Residential Structure in Northern California The structure for this example is a 15-story reinforced concrete special shear wall residential structure in northern California. The SDC is D. The comparison of diaphragm design force levels for the structure by ASCE 7-16 Sections 12.10.1 and 12.10.2 (marked ASCE 7), by ASCE 7-16 Section 12.10.3 (marked ASCE 7 Alt., and by the 2015 NEHRP Provisions (labeled NEHRP), are illustrated in Figure 7. There is very little difference between the design force levels by ASCE 7-16 Section 12.10.3 and the Provisions. These force levels are higher than those given by ASCE 7-10 Sections 12.10.1 and 12.10.2 – throughout the building height for shear-controlled diaphragms and only near the top for flexurecontrolled diaphragms.

Conclusion

Figure 6. Design force level comparisons for 12-story steel-framed office structure (References to ASCE 7 and NEHRP are to ASCE 7-16 and the 2015 NEHRP Provisions, respectively) [Reproduced with permission from FEMA from the upcoming FEMA P-1051].

The alternative diaphragm seismic design force level of ASCE 7-16, mandated for precast concrete diaphragms in buildings assigned to SDC C and above and permitted for other precast concrete diaphragms, cast-in-place concrete diaphragms, and wood sheathed diaphragms on wood framing, departs from the current empirical approach and brings the design forces closer to reality as indicated by observations, test results, and analytical results. The reader is encouraged to consult the 2015 NEHRP Provisions (FEMA, 2015) and commentary for more detailed discussion of the alternative procedure and development of the diaphragm force reduction factor Rs.▪

Acknowledgments Kelly Cobeen, Jose Restrepo, and Mario Rodriguez contributed significantly to the research, understanding and codification of the material presented in this article.

Their efforts, along with those of the other members of Issue Team 6 of the Provisions Update Committee of the Building Seismic Safety Council, originators of the alternative diaphragm design force level, are gratefully acknowledged.

Figure 7. Design force level comparisons for 15-story concrete shear wall residential structure (References to ASCE 7 and NEHRP are to ASCE 7-16 and the 2015 NEHRP Provisions, respectively)[Reproduced with permission from FEMA from the upcoming FEMA P-1051].

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Code Updates code developments and announcements

T

he American Concrete Institute (ACI) Building Code Requirements for Structural Concrete (ACI 318) includes provisions for anchoring to concrete in Appendix D (ACI 318-02 through ACI 31811) and Chapter 17 (ACI 318-14). Anchorages designed to resist seismic load conditions require special consideration. This article discusses the changes between ACI 318-08 Appendix D seismic provisions and ACI 318-11 Appendix D/ ACI 318-14 Chapter 17 seismic provisions, and includes a brief discussion about the International Building Code (IBC) seismic anchoring provisions.

ACI 318 Seismic Provisions for Anchors ACI 318-08 seismic anchoring provisions are given in Part D.3.3, which is included in Part D.3 – General requirements. ACI 318-11 seismic anchoring provisions are given in Part D.3.3 – Seismic design requirements. ACI 318-14 seismic anchoring provisions are given in Section 17.2.3 – Seismic design. ACI 318-08 Part D.3.3 defines seismic anchor design as that which “includes earthquake forces for structures assigned to Seismic Design Category C, D, E, or F”. ACI 318-11 Part D.3.3.1 and ACI 318-14 Section 17.2.3.1 define seismic anchor design as that for “anchors in structures assigned to Seismic Design Category C, D, E, or F”.

The ACI 318 and IBC codes assume cracked concrete conditions for the design of cast-inplace and post-installed anchors because the existence of cracks in the anchor vicinity can result in a reduced ultimate load capacity and increased displacement at ultimate load. ACI 318 requires post-installed anchors to be qualified for seismic load conditions via testing in cracked concrete. Flexural crack widths corresponding to the onset of reinforcing yield under seismic loading are assumed to equal 0.02 inches. Post-installed anchor qualification standards are referenced in ACI 318-08 Part D.3.3.2, ACI 318-11 Part D.3.3.3 and ACI 318-14 Section 17.2.3.3.

ACI 318-08 seismic anchoring provisions must be satisfied for both tension and shear. In contrast to these provisions, ACI 318-11 and ACI 318-14 seismic anchoring provisions permit design for either tension, or shear, or both tension and shear.

Changes in the ACI 318 Anchoring to Concrete Seismic Provisions

By Richard T. Morgan, P.E.

Richard T. Morgan is the Manager for Software and Literature in the Technical Marketing Department of Hilti North America. He is responsible for PROFIS Anchor and PROFIS Rebar software. He can be reached at richard.morgan@hilti.com.

ACI 318-08 Seismic Provisions ACI 318-08 Appendix D seismic design consists of three options defined by the provisions given in Part D.3.3.4, Part D.3.3.5 and Part D.3.3.6. The

24 March 2016

provisions in the option selected must be satisfied for both tension and shear load conditions.

ACI 318-08 Appendix D seismic design criteria can be summarized as follows: • calculate nominal strengths corresponding to possible anchor failure modes per Part D.4.1.

• apply a strength reduction factor (φ-factor) to each nominal strength per Part D.4.1.2.

• apply a seismic reduction factor of 0.75 to non-steel design strengths per Part D.3.3.3.

The commentary RD.3.3.3 notes that the 0.75 factor is applied “to account for increased damage states in the concrete resulting from seismic actions.” When the design is controlled by nonductile anchor strengths, an additional reduction factor must be applied to the calculated anchor design strengths corresponding to brittle failure modes. This criterion will be covered when discussing Part D.3.3.6.

Part D.3.3.4 can be used if the anchorage design is governed by the steel strength of a ductile steel element. The design steel strength in tension, defined by the parameter φNsa, must be the controlling tension design strength compared to the non-steel tension design strengths defined by the parameter (0.75)(φNN). Likewise, the design steel strength in shear, defined by the parameter φVsa, must be the controlling shear design strength compared to the non-steel shear design strengths defined by the parameter (0.75)(φVN). Part D.1 – Definitions defines a ductile steel element as having a tensile test elongation of at least fourteen percent measured over a specified gauge length, and a reduction in cross-sectional area of at least thirty percent. Anchor elements that do not satisfy these criteria, or for which these criteria are not determined, are assumed to be brittle steel elements, which precludes them from design with the provisions of D.3.3.4.

Part D.3.3.5 can be used if the anchorage design is controlled by ductile yielding of the attachment.

The force calculated to yield the attachment must be less than or equal to the calculated anchor design strengths. Tension anchor design strengths are defined as φNsa for steel failure and (0.75) (φNN) for non-steel failure. Shear anchor design strengths are defined by φVsa for steel failure and (0.75)(φVN) for non-steel failure.

Part D.3.3.4 and Part D.3.3.5 are both predicated on a ductile failure mode controlling the anchorage design. The ACI 318 code recognizes, however, that an anchorage design controlled by a ductile failure mode may not be possible. For example, anchor spacing and edge distance, concrete member thickness, or base plate properties may preclude an anchorage design controlled by a ductile failure mode. Therefore, Part D.3.3.6 provides another option that waives any ductility requirement and permits the anchorage design to be controlled by a brittle failure mode. The provisions of Part D.3.3.6 include an additional

ACI 318-08 Appendix D Seismic Tension Provisions

ACI 318-08 Appendix D Seismic Shear Provisions

Design Strength in Tension (ΦNsa or ΦNn)

If anchor element is not ductile: фnonductile фsteel Nsa ≥ Nua

(0.75) фconcrete Ncbg ≥ Nua

(0.75) фnonductile фconcrete Ncbg ≥ Nua

(0.75) фconcrete Npn ≥ Nua

(0.75) фnonductile фconcrete Npn ≥ Nua

(0.75) фbond Nag ≥ Nua

(0.75) фnonductile фbond Nag ≥ Nua

(0.75) фconcrete Nsbg ≥ Nua

(0.75) фnonductile фconcrete Nsbg ≥ Nua

Concrete/Bond pry-out

Steel element

If anchor element is ductile: фsteel Nsa ≥ Nua

Concrete pry-out

D.3.3.6

Anchor element must be ductile for D.3.3.4 фsteel Nsa ≥ Nua

Concrete breakout

ACI 318 PROVISION

D.3.3.4 and D.3.3.5

Bond Strength

ACI 318 PROVISION

Nominal Strength in Shear (Vsa or Vn)

Side Face Blowout

RESISTANCE

D.3.3.6

Pull out/pull through

ACI 318 PROVISION

D.3.3.4 and D.3.3.5

Steel element

ACI 318 PROVISION

Concrete breakout

RESISTANCE Nominal Strength in Tension (Nsa or Nn)

Anchor element must be ductile for D.3.3.4 фsteel Nsa ≥ Nua

If anchor element is ductile: фsteel Vsa ≥ Vua If anchor element is not ductile: фnonductile фsteel Vsa ≥ Vua

(0.75) фconcrete Vcbg ≥ Vua

(0.75) фnonductile фconcrete Vcbg ≥ Vua

(0.75) фconcrete Vcpg ≥ Vua

(0.75) фnonductile фconcrete Vcpg ≥ Vua

Design Strength in Shear (ΦVsa or ΦVn)

Figure 1.

reduction factor which must be applied to anchor design strengths corresponding to brittle failure modes. For simplicity, this factor will be referred to in this article as φnonductile. φnonductile is applied to non-steel anchor design strengths (φnonductile 0.75 φNN and φnonductile 0.75 φVN) as well as to steel design strengths for anchor elements considered to be brittle (φnonductile φsteel Nsa and φnonductile φsteel Vsa). The default value for φnonductile is 0.4; however, it can vary depending on the design conditions being considered. Part D.3.3.6 notes that a φnonductile value of 0.5 can be used for “anchors of stud bearing walls” because this application

typically consists of multiple anchors capable of load redistribution. The 2009 IBC Section 1908.1.9 waives the use of φnonductile for anchorage of nonstructural components and anchors designed to resist wall out-of-plane forces.

Figure 1 summarizes ACI 318-08 Appendix D seismic calculations.

ACI 318-11 and ACI 318-14 Seismic Provisions ACI 318-11 anchor provisions are given in Appendix D – Anchoring to Concrete.

ACI 318-14 anchor provisions are given ADVERTISEMENT–For Advertiser Information, visit www.STRUCTUREmag.org

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in Chapter 17 – Anchoring to Concrete.

Other than the numbering system, there is no difference in content between ACI 318-11 Appendix D and ACI 318-14 Chapter 17 anchoring provisions. This section discusses ACI 318-11 Appendix D seismic anchor provisions and references the corresponding ACI 318-14 Chapter 17 section in parentheses.

ACI 318-11 and ACI 318-14 seismic anchor calculations do not have to be performed if the earthquake component of the factored load acting on the anchorage is less than or equal to twenty percent of the total factored load acting on the anchorage. Unlike ACI

318-08 Appendix D seismic provisions, ACI 318-11 and ACI 318-14 seismic anchor provisions permit design for either tension conditions, or shear conditions, or both tension and shear conditions.

ACI 318-11 and ACI 318-14 Seismic Tension Provisions ACI 318-11 Part D.3.3.4 – Requirements for tensile loading (ACI 318-14 Section 17.2.3.4) permits a tiered approach to seismic design. Part D.3.3.4.1 (Section 17.2.3.4.1) waives the requirement to design for seismic tension if the “tensile component of the strength-level earthquake force” is less than or equal to 20 percent of the total factored tension load. Equations consisting of factored load combinations are given in ACI 318-11 Section 9.2 – Required Strength (ACI 318-14 Section 5.3 – Load factors and combinations). The parameter E in these equations corresponds to the earthquake force component of the factored load.

When considering tension loads that include E, if the value calculated for E is less than or equal to twenty percent of the total factored load, no seismic calculations are required for tension. In this case, the tension design for the anchorage will be per Table D.4.1.1– Required Strength of Anchors (ACI 318-14 Table 17.3.1.1). If E is greater than twenty percent of the total factored tension load, Part D.3.3.4.2 (Section 17.2.3.4.2) requires the tension design for the anchorage to be performed using one of the options given in Part D.3.3.4.3 (Section 17.2.3.4.3).

ACI 318-11 Appendix D and ACI 318-14 Chapter 17 seismic tension provisions can be summarized as follows: • calculate design strengths corresponding to possible anchor failure modes per Table D.4.1.1 (Table 17.3.1.1).

• apply a seismic reduction factor of 0.75 to non-steel tension design strengths per Part D.3.3.4.4 (Section 17.2.3.4.4).

Seismic tension options include anchorage design controlled by the strength of the attachment (ductile or brittle failure), or anchorage design controlled by the anchor design strengths (ductile or brittle failure).

When the anchorage design is controlled by a brittle anchor failure mode, an overstrength factor (Ω0) must be applied to the earthquake component (E) of the factored load.

Part D.3.3.4.3(a) (ACI 318-14 Section 17.2.3.4.3(a)) provisions are only relevant to ductile anchor elements. A ductility check must first be performed. The purpose of this check is to provide a reasonable expectation,

based on nominal strengths, that the anchor element will have yielded when ultimate load is reached. The check requires the ratio (Nua,i /1.2Nsa), corresponding to steel failure, to exceed the ratio (Nua /NN), corresponding to non-steel failure. The check defines non-steel failure as “concrete-governed strengths”. Note that bond strength is considered a “concretegoverned” tension strength for purposes of this check. Nua,i corresponds to the highest-loaded anchor in tension, and the steel strength of the anchor is defined as 1.2 times the nominal steel strength (Nsa). Nua corresponds to the factored tension load, and NN corresponds to the nominal concrete breakout strength, nominal bond strength, nominal pullout strength or nominal side-face blowout strength for the anchorage. If (Nua,i /1.2Nsa) for steel failure is greater than (Nua /NN) for all possible non-steel failure modes, the ductility check is satisfied, and the calculated design strengths will be (φNsa) for steel failure and (0.75)(φNN) for non-steel failure modes.

ACI 318-11 Part D.3.3.4.3(b) (ACI 318-14 Section 17.2.3.4.3(b)) requires the anchorage design to be controlled by ductile yielding of the attachment. ACI 318-11 Part D.3.3.4.3(c) (ACI 318-14 Section 17.2.3.4.3(c)) permits the anchorage design to be controlled by the strength of a nonyielding attachment, e.g. crushing of a wood sill plate. The calculated design strengths for both options will be defined by the parameters (φNsa) for steel failure and (0.75)(φNN) for non-steel failure modes.

ACI 318-11 Part D.3.3.4.3(d) (ACI 318-14 Section 17.2.3.4.3(d)) requires anchor design strengths to be greater than or equal to the factored tension load inclusive of an Ω0 overstrength factor in the earthquake component (E) of the factored load.

ACI 318-11 and ACI 318-14 Seismic Shear Provisions ACI 318-11 Part D.3.3.5 – Requirements for shear loading (ACI 318-14 Section 17.2.3.5) also permits a tiered approach to seismic design. Part D.3.3.5.1 (Section 17.2.3.5.1) waives the requirement to design for seismic shear if the “shear component of the strength-level earthquake force” is less than or equal to 20 percent of the total factored shear load. When considering shear factored load equations that include E, if the value calculated for E is less than or equal to 20 percent of the total factored load, no seismic calculations are required for shear.

In this case, the anchorage design will be per Table D.4.1.1 (ACI 318-14 Table 17.3.1.1).

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If E is greater than 20 percent of the total factored shear load, Part D.3.3.5.2 (Section 17.2.3.5.2) requires the shear design for the anchorage to be performed using one of the options given in Part D.3.3.5.3 (Section 17.2.3.5.3).

Unlike ACI 318-08 Appendix D, ACI 318-11 Appendix D and ACI 318-14 Chapter 17 seismic shear provisions do not apply a reduction factor of 0.75 to non-steel design strengths. The calculated shear design strengths are defined by the parameters φVsa for steel failure, and φVN for non-steel failure modes. Another difference between ACI 318-11/ACI 318-14 seismic shear anchoring provisions and ACI 318-08 seismic shear anchoring provisions is that anchorage design based on yielding of a ductile anchor element is not offered as a seismic shear option in either ACI 318-11 or ACI 318-14.

ACI 318-11 Part D.3.3.5.3(a) (ACI 318-14 Section 17.2.3.5.3(a)) requires the anchorage design to be controlled by ductile yielding of the attachment. If this cannot be satisfied, ACI 318-11 Part D.3.3.5.3(b) (ACI 318-14 Section 17.2.3.5.3(b)) provides an option that permits the anchorage design to be controlled by the strength of a nonyielding attachment.

ACI 318-11 Part D.3.3.5.3(c) (ACI 318-14 Section 17.2.3.5.3(c)) requires the calculated anchor design strengths to be greater than or equal to the factored shear load inclusive of an Ω0 overstrength factor in the earthquake component (E) of the factored load. Figure 2 summarizes ACI 318-11 Appendix D and ACI 318-14 Chapter 17 seismic calculations.

IBC Seismic Provisions Versus ACI 318 Seismic Provisions The IBC references codes and standards that are considered part of the requirements of that particular IBC version. Chapter 35 – Referenced Standards in the 2012 IBC and 2015 IBC references ACI 318-11 and ACI 318-14 respectively. Chapter 19 – CONCRETE references ACI 318 anchoring provisions; however, the 2012 IBC was published prior to the publication of ACI 318-11, which resulted in the 2012 IBC seismic anchoring provisions given in Section 1905.1.9 referencing ACI 318-08 Appendix D seismic provisions instead of ACI 318-11 Appendix D seismic provisions.

This situation illustrates the importance of understanding the local codes, because jurisdictions may make amendments to the IBC model codes. For example, jurisdictions can amend the 2012 IBC Chapter 19 anchor provisions prior to adopting the 2012 IBC,

Figure 2.

or jurisdictions that have adopted the 2012 IBC without amendments can permit design per ACI 318-11 Appendix D via approval by the authority having jurisdiction. 2015 IBC Chapter 19 seismic anchoring provisions reference ACI 318-14 Chapter 17 seismic anchoring provisions.

Summary ACI 318 seismic anchoring provisions provide options for a design that is controlled

by the strength of the attachment or by the strength of the anchors. A design controlled by some form of ductile failure is preferable; however, a design controlled by a non-ductile failure mode is permissible if additional provisions relative to either the load or resistance calculations are satisfied.

Post-installed anchors must be qualified by testing for use with the seismic anchor provisions of ACI 318.

ACI 318-08 seismic anchoring provisions must be satisfied for both tension and shear

load conditions. ACI 318-11 and ACI 318-14 seismic anchoring provisions can be performed for tension only, shear only, or tension and shear.

The 2012 IBC seismic anchoring provisions reference ACI 318-08 Appendix D provisions instead of ACI 318-11 Appendix D provisions due to a difference in the publication date of each code. This discrepancy has been eliminated with harmonization of the 2015 IBC and ACI 318-14 Chapter 17 seismic anchoring provisions.▪

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Structural rehabilitation renovation and restoration of existing structures

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ging infrastructure and stricter seismic design guidelines cause many structures to need seismic upgrading. This is especially true in the Pacific Northwest, where the Puget Sound area is prone to large magnitude subduction zone earthquakes. Most recently, the 2001 Nisqually Earthquake, one of the largest earthquakes in Washington’s recent history, caused over $2 billion in damage and resulted in hundreds of injuries. The tremors could be felt as far away as Spokane in the eastern part of the state.

While seismic retrofits provide a significant reduction in life-safety and financial risks, most upgrades are only undertaken when they are mandated by jurisdictional or investor/lender requirements. Significant renovations or building improvements, building alterations, unsafe building ordinances, or requirements by financial institutions are the most common triggers.

Older buildings may go through a number of seismic upgrades during their lifespans. One recent project was the seismic retrofit of the Sunset House in downtown Seattle, the subject of this article, where Fiber Reinforced Polymers (FRP) were successfully employed.

Seismic Retrofit with Fiber Reinforced Polymers Using ASCE-41 to Retrofit a Multi-Story Concrete Building

By Sarah Witt and Greg Gilda, P.E., S.E., LEED AP

Sarah Witt is Director of Strengthening at Contech Services Inc. and is an active member of ACI Committee 440 (Fiber Reinforced Polymer Reinforcement).

Sarah can be reached at sarah@contechservices.com.

Greg Gilda has been active with the Structural Engineers Association of Washington and also volunteers and chairs the Construction Codes Advisory Board for the City of Seattle.

Greg can be reached at ggilda@dci-engineers.com.

Fiber Reinforced Polymers Fiber Reinforced Polymers are well recognized as an effective seismic retrofit material for existing concrete buildings. This segment of the strengthening industry is more than twenty years old and several successful projects have been installed. Some of these retrofitted buildings have experienced significant earthquakes and performed as designed, demonstrating the effectiveness of the technology. Extensive laboratory testing and actual earthquakes have led to the development of reliable design methodologies and guidelines for FRP to be used by the engineering community.

There are many methods available for the seismic retrofit of buildings including section enlargement with concrete and/or steel, adding more members such as new shear walls, beams or columns, base isolation or braced frames. While these methods are very effective at improving the building performance, they may result in lost space, long shutdown times during construction or cause a large impact on day-to-day operations of an occupied building. For more than twenty years in the United States, FRP has been used as an alternate method of structural reinforcing for existing buildings and has gained acceptance as a structural repair method throughout the world.

FRP materials consist of high strength fibers in a polymer matrix. The fibers provide the strength

28 March 2016

and stiffness, and the matrix provides load transfer and environmental protection to the fibers. The most commonly used fibers are glass and carbon.

FRP design is based on the composite properties of both of these materials when combined. Design professionals can ensure they specify well tested products by requiring the materials to have an ICC Listing. ICC AC 125, Acceptance Criteria for Concrete and Reinforced and Unreinforced Masonry Strengthening Using Externally Bonded Fiber-Reinforced Polymer (FRP) Composite Systems, has a comprehensive test program that includes environmental durability to ensure materials can be consistently designed and perform as expected.

The American Concrete Institute (ACI) has also published material specification for FRP, ACI 440.8R-13, Specification for Carbon and Glass Fiber-reinforced Polymer (FRP) Materials Made by Wet Layup for External Strengthening of Concrete and Masonry Structures. This is the first code document in the United States for externally bonded wet layup FRP materials.

FRP materials have high strength-to-weight ratios, which make them an ideal material for seismic retrofit. They do not add significant mass to a structure, while they can be designed to add ductility, confinement, moment or shear capacity to existing structural members. This allows for local strengthening without concerns of transferring the added weight to the foundation. FRP can also be designed to add strength without changing the stiffness, minimizing the extent of additional analysis of the structure after strengthening. FRP materials add minimal depth to the structure, with an average application less than ¼-inch thick.

FRP materials are designed as tension members that work in conjunction with the existing member. Currently, there are several design codes and recommendations for these materials throughout the world. In the United States, ACI 440.2R-08, Guide for the Design and Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures, provides a widely used design guideline for the materials. It was first published in 2002 and republished in 2008. However, this document currently does not cover seismic retrofit design.

Seismic Retrofit Design Standards At the time the Sunset House building project was undertaken, the International Existing Building Code (IEBC) had not been formally adopted as a standard in Seattle and the primary standards were ASCE 31-03 (evaluation) and ASCE 41-06 (retrofit). Seattle has now adopted the IEBC with local amendments and ASCE 41-13 now incorporates the previous two documents (ASCE 31-03 and ASCE 41-06) into one document.

The IEBC currently allows for two approaches to analyzing and strengthening existing buildings. The first is to review the structural system relative to the current International Building Code (IBC) with a reduction in force and detailing requirements in recognition of the strength of the original structures and the shorter expected remaining life-spans.

Alternatively, the second method permitted in the IEBC is ASCE 41-13, Seismic Rehabilitation of Existing Buildings.

In the past, engineers were confronted with a discrepancy between the force levels derived from the two approaches. The two approaches handle the material and structural behavior quite differently, making the IBC loads appear to be significantly less than the ASCE 41 loads. Generally, the structural solutions will be similar to the ASCE 41 approach, providing a more detailed approach for existing buildings. The ASCE 41 approach provides performance objectives based on the desired performance of the building. The Basic Performance Objective for Existing Buildings (BPOE) is intended to provide requirements for existing buildings that produce performance equivalent to the reduced IBC loads approach.

Currently, ASCE 41-13 does not address FRP as a material used as part of a seismic

Structural Evaluation of Sunset House using ASCE 41

Figure 1. Limited work area for installation of FRP for tension loads.

system, i.e. the deformation-controlled elements with m-factor force reductions, so some judgement and a solid understanding of the design approach used to produce the loads is required until FRP is included in future editions. FRP can currently be used as a forcecontrolled element based on the maximum load it will be subjected to.

The Sunset House building is a ten-story, 1970s era residential structure built with precast concrete plank floors and a combination of reinforced concrete and masonry shear walls (Figure 1). The seismic evaluation was required as part of a major rehabilitation of the building. This evaluation was performed using ASCE 41-06, Seismic Retrofit of Existing Buildings. A Tier 1 checklist, a Tier 2 evaluation and a Tier 3 linear dynamic analysis were used to identify and mitigate the seismic issues in the building.

ASCE 41 provides significant performancebased guidance to designers in evaluating and upgrading existing buildings, incorporating damage observations from past seismic events and laboratory testing. The seismic performance objectives can be customized depending on the occupancy, owner objectives, type of building, etc.

The retrofit of the Sunset House was performed to meet a Life-Safety Performance Objective with identified deficiencies in the tensile capacity of boundary details in some of the shear walls and in the shear capacity at some of the link beams in the corridor walls. While these deficiencies were fairly

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March 2016

isolated, potential impact to the performance of the structure would have likely been significant.

Design of FRP for Sunset House The location of the deficient structural elements relative to the corridors and living spaces significantly limited the size of the solutions used for the strengthening. After exploring a variety of potential strengthening methods, FRP was selected based on cost effectiveness, as well as the minimal architectural impact to the living spaces. FRP not only provided minimum impact to the space, but also provided the required strength without a significant increase in stiffness (an increase in stiffness on this project would have affected the distribution of lateral loads and would have likely required strengthening of additional elements). Other options, such as adding steel and concrete to the walls, were not feasible due to the space constraints.

The FRP was designed for the two major components of the retrofit design, adding shear and overturning capacity to walls and increasing the shear capacity of the coupling beams over several of the doorways. Increasing the overturning capacity of the shear walls required increasing the tensile capacity at the wall boundary connections. This was provided by adding bonded layers of carbon and glass fibers to connect the walls between the floor levels. Shear capacities at the walls were similarly strengthened with bonded glass fiber sheets applied to connect the top and bottoms of the walls to the floor slabs. Coupling beams’ capacities were increased by bonding FRP to the faces of the wall. The FRP was extended beyond the ends of the coupling beams to develop the strength of the fibers.

The seismic system m-factors used in ASCE 41 to represent the expected ductility are not specified for systems including FRP, such as shear walls. However, the design team felt that FRP was a good option. Design coordination among the project team allowed the FRP to be utilized and provided the best retrofit option for the building.

As ACI 440.2R does not have a seismic section, the design equations for wall shear from ICC AC 125 were the basis of FRP design for the shear walls. The shear wall boundary elements and coupling beams were design based on appropriate design strains developed from expected movement in the walls and the composite behavior with the FRP. Reduction factors were chosen by looking at the development of the design loads and commonly used FRP reduction factors.

Figure 2. Existing slab is cut back to allow the FRP to be installed continuously from floor to floor.

Installation The Sunset House was an occupied facility during the installation of the FRP strengthening. This created construction challenges for both scheduling and coordination among the occupants. Installation was phased with access to only certain floors and work areas at a given time. The corridors were narrow and access to each living unit needed to be maintained during construction (Figure 2).

Work began with protection of the work area and removal of the existing wall finishes.

One sided wall applications of FRP are bond critical in that all the force of the FRP is transferred through the bond into the substrate.

For bond critical applications, the surface preparation is the most important step in the installation process. It is necessary to control and capture the dust created by the mechanical abrasion of the concrete surface. The use of HEPA vacuums and negative air machines were used, enabling this process to proceed without the release of dust into the public areas, reducing the impact on the occupants.

Following surface preparation, the concrete surface is primed with a system compatible epoxy. As the facility was occupied, a primer epoxy was used that had no volatiles and no toxic odors. Additionally, negative air machines were used as ventilators to keep fresh air movement within the working areas.

Figure 3. Exterior view of Sunset House.

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Following the priming of the surface, the FRP fabrics are saturated with the epoxy matrix and placed onto the prepared wall surface.

The retrofit design included vertical fibers that needed to run continuously from upper floors down to the foundation to ensure load path continuity. While there are options to use steel plates or fiber anchors for transferring the forces through the slab, the design was detailed to cut back the slab and allow access for the fiber to be continuously installed on the wall and floor (Figure 3). While removing small sections of the slab, the existing rebar remained in place. Once the fiber was installed, the concrete was replaced and the connection between the wall and slab was not damaged. This avoided any steel plates in the corridor and other connection details that have a more significant profile.

The final installation of the FRP system was less than ½-inch thick and maintained the required width of the egress routes. The facility remained occupied and the work was completed on time. The FRP strengthening significantly increases the ductility of the critical connections, as the shear walls and coupling beams and resulting building performance is expected to be significantly better than it’s previous as-built condition.

Future Developments for FRP The Sunset House demonstrates how ASCE 41 and FRP can be effectively used in conjunction for seismic retrofit design. Moving ahead, the ACI 440 committee is working to incorporate a seismic design chapter into the ACI 440.2R document. This chapter will reference ASCE 41, providing more guidance to the design engineer on using FRP to retrofit structures. In addition, the industry needs to look at developing appropriate m-factors for these strengthened members.▪

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Structural PracticeS practical knowledge beyond the textbook

S

tructural engineers know the mechanics of the seismic provisions of the International Building Code (IBC) and ASCE 7-10. They know how to get Ss and S1 for a site and apply the equations to calculate a seismic response coefficient (Cs) that is used to calculate the seismic base shear, which is used to size the seismic resisting elements of the building. However, many do not understand the background behind the equations and the coefficients. The purpose of this article is to establish a foundation for a common understanding as an aid in discussing seismic concepts with owners, clients, and other engineers.

Earthquakes vs. Ground Motion Talking Point #1 Engineers design for a specific ground motion shaking intensity, not a specific earthquake.

Talking Point #2 The IBC mandates the Maximum Considered Earthquake (MCER) ground motion that must be considered in the design process.

Earthquakes cause the earth to shake. It is the ground shaking caused by the earthquake that causes building movement and damage. The code writers didn’t do us any favors by using the terms Maximum Considered Earthquake and Design Earthquake (sometimes referred to as the Design Basis Earthquake). Using these terms alone implies that we are designing for a specific earthquake. The MCER is a ground shaking intensity, given as a response acceleration, which is generally caused by a range of earthquakes of different magnitudes from several earthquake sources, or in some areas, a single earthquake from a dominant earthquake source. Engineers should be clear and use the phrase Maximum Considered Earthquake Ground Motion to emphasize that it is the ground motion shaking intensity we are designing for and not the earthquake.

Are You Communicating Seismic Concepts Correctly?

By Brent Maxfield, S.E.

Brent Maxfield is an active member of SEAU and is the Past President of the EERI Utah Chapter. He has an active interest in seismic topics. He can be reached at BrentMaxfieldPE@gmail.com.

This article is based on an article and is used with permission from the SEAU News, Seismic Communication: Let’s Get on the Same Page, May 2015.

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

Talking Point #3 Beginning in 2012, the IBC established a riskmodified MCER, which is based on a uniform risk of building collapse (1% in 50 years, or 1/5000 per year), and results in the same risk of a building collapse in New York, Atlanta, Seattle, and the rest of the country.

The Uniform Hazard MCE ground shaking intensity is no longer used in building design (see Talking Point #16). To get the MCER, the Uniform Hazard ground shaking intensity is increased or decreased until there is a 1% probability of building collapse in 50 years. In most areas of the country, the code shaking intensity is reduced from the Uniform Hazard, meaning that a shaking intensity matching the MCER is likely to occur more often than a shaking intensity matching the older MCE.

32 March 2016

Talking Point #4 The MCER is the ground shaking intensity that the IBC requires to be considered, but larger shaking intensities are possible. In reality, the MCER should be thought of as the minimum ground shaking intensity that must be considered.

Talking Point #5 A specific fault rupture or a specific magnitude earthquake may not cause an MCER level ground shaking intensity. A fault could rupture many times before it results in the MCER level ground shaking intensity being reached or exceeded. The maximum expected earthquake on a fault may not cause an MCER level ground shaking intensity, but conversely, it is also possible that it could cause the MCER level ground shaking intensity to be exceeded.

Talking Point #6 Even though the MCER is the considered ground shaking intensity, building design and performance is usually based on an intensity of 2/3 of the MCER (see Talking Point #8), which this article will call the design ground shaking intensity.

Code Performance Expectations It is important to understand the life safety performance expectations of the code and to be able to clearly communicate these to clients and building owners.

Talking Point #7 Although the term Life Safety has specific meaning to engineers, it can have other interpretations and be misconstrued by clients and owners.

Three key damage states not specifically mentioned but implied by the IBC are Immediate Occupancy, Life Safety, and Collapse Prevention.

A building owner who would be satisfied with a Collapse Prevention damage state may say that they want a Life Safety damage state because they misunderstand the terms. Here are simple definitions of performance expectations: • Immediate Occupancy: A building can be used after some cleanup occurs and can be occupied during the repairs to fix building damage.

• Life Safety: A building could have significant structural damage, but it has reserve structural capacity to resist aftershocks. The building may not be able to be occupied until after repairs are made.

• Collapse Prevention: A building has been pushed to the limits of its strength and stiffness and is on the verge of collapse. Aftershocks may cause the building to collapse.

Talking Point #8 For an accurate discussion of performance expectation, engineers must provide clients and building owners with a clear understanding of the expected building performance and the ground

shaking intensity at which that building performance is expected to occur.

The implied safety objective of the IBC is to achieve Life Safety if the building site experiences a ground shaking intensity equal to the design ground shaking intensity (2/3 MCER) and to achieve Collapse Prevention if the building site experiences a ground shaking intensity equal to the MCER. The IBC uses a Seismic Importance Factor of 1.5 for essential facilities to increase the strength of the building and reduce the ductility demand on the structure. The objective for an essential facility, such as a hospital, is to achieve Immediate Occupancy if the building site experiences a ground shaking intensity equal to the design ground shaking intensity (2/3 MCER) and to achieve Life Safety if the building site experiences a ground shaking intensity equal to the MCER. Note that a hospital may not be operational if it experiences the MCER ground shaking intensity.

components will be operational for the MCER ground shaking intensity. At MCER intensity, nonstructural elements may fall, causing localized deaths.

Code Ground Motions The following is a very brief description covering basic concepts about how the Ss and S1 values found on the USGS website are derived. These values could be based on either a deterministic or a probabilistic ground motion shaking intensity.

Deterministic Ground Motions Talking Point #12 A deterministic ground motion analysis for a specific fault will predict a range of ground shaking intensities from a specific magnitude earthquake.

Talking Point #13 It is only possible to predict a range of ground shaking intensities from a specific earthquake at a specific site. It is impossible to predict what the exact ground shaking intensity will be at a specific site. continued on next page

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Talking Point #9 The risk of collapse is reduced for buildings that are designed using Seismic Importance Factors of 1.25 or 1.5.

Seismic Importance Factors are intended to improve the building performance at the design ground shaking intensity (2/3 MCER) and at the MCER ground shaking intensity. This is accomplished by reducing the Response Modification Factor (R). Note that the Seismic Importance Factor is not applied to the demand coefficient (SDS or SD1).

Talking Point #10 The IBC allows for a very small risk of building collapse.

1) There is a 1% in 50 years probability (1/5000 per year) that a building will collapse due to a seismic event.

2) Up to 10% of buildings designed and constructed per the IBC could experience some collapse when subjected to the MCER ground shaking intensity.

Talking Point #11 Nonstructural components are designed for the design ground shaking intensity (2/3 MCER), and there are no performance goals for a MCER level ground shaking intensity.

At the design ground shaking intensity (2/3 MCER), components with an Ip = 1.0 can be expected to have major damage, but significant falling hazards are avoided.

Components with an Ip = 1.5 can have limited damage, but should remain functional.

There should be no expectation that essential

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This estimation of ground motion shaking intensity is called a deterministic approach. From a code standpoint, a deterministic ground motion looks at the 84th percentile response acceleration from all nearby active faults and selects the largest response acceleration (shaking intensity).

Probabilistic Ground Motions Talking Point #14 Probabilistic ground motions refer to the probability that a specific ground shaking intensity level will be exceeded. They do not refer to a specific magnitude earthquake (see Talking Point #1).

Deterministic ground motion predictions assume that a characteristic earthquake occurs, but they do not consider the likelihood of it occurring. Probabilistic ground motions add another dimension by considering the probability that a specific magnitude earthquake will actually occur. They are expressed as a probability that a specific level of ground shaking intensity will be exceeded in a specific period of time. For example, a 10% in 50 year ground shaking intensity means that there is a 10% probability that the shaking intensity will be exceeded in 50 years. It could also be stated that there is a 1/475 probability that in any one year the shaking intensity would be exceeded. A 2% in 50 year ground shaking intensity is larger and rarer. There is only a 1/2475 probability that the shaking intensity will be exceeded in any year.

Talking Point #15 The USGS acknowledges that a magnitude 7.5 earthquake could occur anywhere in the United States. If a probabilistic ground shaking intensity is low, it is not because there can

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When a specific magnitude earthquake happens, the ground shaking intensity around the region could vary greatly. Areas close by each other can have very different levels of ground shaking intensity. Recordings from previous same-magnitude earthquakes show that there is a wide range of possible shaking intensities.

Engineers and seismologists who have studied earthquake ground motions can only predict a range of shaking intensities from a specific earthquake based on fault type, distance, site conditions, and other factors because there is so much variability in these parameters. The ground motion prediction equations provide a median shaking intensity and a standard deviation, from which the range of shaking intensities can be calculated, and do not predict what the exact ground shaking intensity will be.

Predicting what the ground shaking intensity will be from a future earthquake is like predicting when a kernel of popcorn will pop. Some kernels pop early, some pop late. If you were given a kernel of popcorn, you could not predict when it will pop, but you could say that there is a 50% likelihood that it would pop before the median time. Or, you could say that there is an 84% likelihood that it would pop before the median + one standard deviation time. So the question is not, “When will it pop?” but “How likely is it to pop before a specific period of time?” Likewise, it is not appropriate to guess what the exact ground shaking intensity will be from a magnitude X earthquake. Instead, it is better to ask, “What is the ground shaking intensity level where there is an 84% likelihood that a magnitude X earthquake will cause a ground shaking intensity less than this level?”

This question could also be asked for 50%, or any other percentile.

never be a large magnitude earthquake; it is because the probability of a large magnitude earthquake is very low.

A probabilistic ground shaking intensity is based on the probability of various size earthquakes impacting a specific location. It considers both nearby and distant faults, and considers how many very small to large earthquakes have occurred in the area. It is a complicated process to calculate a probabilistic ground motion, and is beyond the scope of this article.

Deterministic Caps Talking Point #16 The older MCE (prior to the 2012 IBC) is a Uniform Hazard ground shaking intensity with a 2% in 50 year probability of being exceeded (1/2475 per year).

When a 2% in 50 year probabilistic ground shaking intensity is calculated for every location around the United States, areas with many active faults will have very high ground shaking intensities, and areas with no active faults will have much lower ground shaking intensities. Each location has the same probability that the calculated ground shaking intensity will be exceeded: 2% in 50 years, or 1/2475 per year. This creates what is referred to as a “Uniform Hazard.”

Because some areas of California have many active faults, it results in very high probabilistic ground shaking intensities. It was decided to cap the MCE based on the deterministic ground shaking intensity of the controlling nearby fault.

The MCER also uses a deterministic cap. The current cap is based on the 84th percentile deterministic ground shaking intensity.

Talking Point #17 When the MCER ground shaking intensity (see Talking Point #3) is based on the deterministic cap, then the shaking intensity will be lower than the risk adjusted Uniform Hazard ground shaking intensity. Because it is lower, the risk of building collapse is greater than the IBC objective of 1% in 50 years, and it is likely to occur more often than either the 2% in 50 year Uniform Hazard or the risk adjusted Uniform Hazard ground shaking intensities.

Concluding Thoughts This article only scratched the surface on many topics. Hopefully the information provided will whet our appetite to learn more and establish a common understanding for discussing seismic concepts.▪ USGS website: http://earthquake.usgs.gov/ designmaps/us/application.php.

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Structural teSting issues and advances related to structural testing

T

he 1933 Long Beach earthquake showed that unreinforced doublewythe masonry brick walls did not perform well. Consequently, California regulators imposed a requirement that double-wythe brick masonry be reinforced and grouted, and that the newly constructed masonry be destructively tested by drilling a core specimen horizontally through the wall to test the bond between the clay masonry unit and grout for shear capacity. The bond criteria for grout to masonry unit was arbitrarily set at 100 psi. In 1983, the bond criteria was changed to 2.5√f 'm psi, a value nearly equal to 100 psi.

Over the past 75 years, the requirement has morphed into application to single-wythe hollow unit masonry walls which was never the intent of the provision and ignores the benefit of webs and tapers in Concrete Masonry Units (CMU).

Additionally, there is discussion at the national level on whether or not destructively coring and testing the masonry cores is a worthwhile effort. The following analysis is based on current code provisions and puts the discussion into a rational perspective.

When a reinforced masonry wall is subjected to out-of-plane loads, the tension is carried by the reinforcement and the compression by the masonry. In this context, the masonry is a combination of masonry units, mortar, and grout. There are also shear stresses in the wall. The shear stresses are both perpendicular to the face of the wall, as

Coring of Concrete Masonry Walls: Is it Necessary?

By Richard M. Bennett, Ph.D., P.E.

Richard M. Bennett is a professor of Civil and Environmental Engineering and the Director of Engineering Fundamentals at the University of Tennessee, Knoxville. He served as Chairman of the Flexural and Axial Loads Subcommittee of the TMS 402/602 Code Committee from 2004 to 2010. From 2010 to 2013, he was the Vice-Chair of the Main committee and is currently Chair of the 2016 TMS 402/602 Committee. He can be reached at rmbennett@utk.edu.

Figure 2. Illustration of the destructive nature of coring. The first attempt hit reinforcement causing further damage.

36 March 2016

Figure 1. Stresses in a reinforced masonry wall.

well as parallel to the face of the wall. The shear stresses parallel to the face of the wall are similar to those that develop between the structural steel and the concrete in a composite steel/concrete slab beam. The stresses in the cross-section are shown in Figure 1.

The Masonry Society (TMS) 402 Code, Building Code Requirements for Masonry Structures, requires the wall to be designed to carry the shear forces perpendicular to the face of wall (2013 TMS 402 Section 8.3.5 for Allowable Stress Design, and Section 9.3.5.3 for Strength Design). There are no requirements in TMS 402 with regard to the shear stresses parallel to the face of the wall. However, the California Division of State Architect and the California Office of Statewide Health Planning and Development have requirements for core testing of masonry walls. The minimum average unit shear interface requirement between the grout and face shell has been arbitrarily set at 2.5√f m' psi. This requirement is presumably to verify that there is sufficient bond between the grout and the masonry unit to carry the shear stresses. The coring, shown in Figure 2, demonstrates the destructive nature of the testing. The question is whether this coring is necessary, and whether TMS 402 should even consider a similar requirement.

To answer the question on the necessity of coring, a variety of wall configurations were analyzed. All walls were considered to be fully grouted and simply supported. The analysis procedure was as follows: 1) Select a wall height, block size, reinforcement bar size, reinforcement bar spacing, axial load, and a specified compressive strength, f 'm. Type S Portland cement-lime mortar was assumed for all

walls. Wall weights were determined based on 125 pcf units, although this assumption has a negligible effect on the results. The axial load was assumed to act concentric with the wall. Any eccentricity to the axial load would reduce the out-of-plane load the wall could carry.

In some cases, loads were unrealistically high, being several hundred psf, but the load was still used.

Height t (ft) (inch)

Axial (k/ft)

wu (psf )

Bar Bar Spacing Size (#) (inch)

f'm (psi)

Shear (lb)

Shear stress (psi)

12

7.625

1

195.8

5

16

2000

1175

16.0

12

7.625

1

139.1

5

24

2000

835

10.7

12

7.625

1

110.3

5

32

2000

662

8.0

10

7.625

1

270.5

5

16

2000

1353

19.2

12

7.625

1

195.8

5

16

2000

1175

16.0

16

7.625

1

114.0

5

16

2000

912

12.0

20

7.625

1

72.5

5

16

2000

725

9.6

of close to 400 psf, an unrealistically high out-of-plane load.

To summarize, the analyses made several conservative assumptions, resulting in a very conservative analysis. To review, the conservative assumptions were: 1) The axial load is considered to act concentrically, resulting in the largest shear force for a given moment capacity.

2) The wall is loaded to the maximum out-of-plane load that it can carry. Typically, due to discrete reinforcement sizes and spacings, and prescriptive reinforcement requirements, walls are not loaded to the maximum out-of-plane capacity.

3) Any interlocking due to offset webs, block taper, etc. was neglected. The shear surface was considered to be planar.

Even with a very conservative analysis, the maximum shear stress was only 19.2 psi. The 19.2 psi was for a 10-foot high wall with unrealistically high out-of-plane loads. Under

typical load conditions, the shear stress was 16 psi or less. This shear stress is much less than the 100 psi that was the initial arbitrary California requirement, and also much less than 2.5√f 'm , which would be about 97 psi for f m' = 1500 psi and 112 psi for f m' = 2000 psi.

Based on the above results, two conclusions can be drawn.

1) No core testing is required. The shear stresses are very low. Additionally, the above analysis does not consider the benefit of the homogeneous concrete masonry unit which has a continuous connection between the cross web and face shell, taper of the CMU or interlock of overhanging mortar fins.

2) TMS 402 is justified in not requiring designers to check the shear stress at the grout/face shell interface. That will not control the design.

The complete report along with calculations and expanded tables can be viewed online at http://cmacn.org/PDF/Masonry_Chron_ Winter_2016.pdf.▪

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2) The wall was analyzed using the “slender wall procedure”, Section 9.3.5.4.2 of the 2013 TMS 402 Code, to determine the maximum out-ofplane load the wall could carry. In some cases, loads were unrealistically high, being several hundred psf, but the load was still used.

3) Based on the maximum out-of-plane load, the maximum shear force was calculated. From the maximum shear force, the shear stress at the interface between the grout and face shell was calculated. If the wall is treated as a traditional composite section, and the equivalent rectangular stress block is in the face shell, the shear force at the grout/face shell interface will be based on the yield force of the steel. If part of the equivalent rectangular stress block were in the grouted core, the shear stress at the interface would be reduced.

The shear stress can be obtained as the shear force divided by the shear area over half the wall height.

Typical results are shown in the Table. The first set of results is for the bar spacing being varied, increasing from 16 inches up to 32 inches. The second set of results is for different height walls. Many other conditions were examined, including varying the axial load, varying f 'm, varying the eccentricity of the axial load, and examining 12-inch walls with the bars offset from the center. Similar results were obtained in all cases.

A review of these results shows that the shear stress increases as wall height decreases. The highest shear stress is less than 20 psi for a 10-foot high wall, a typical story height. Most masonry walls are at least 10 feet high and the resultant shear stress was for an out-of-plane load

Typical results of analysis of a variety of wall configurations.

Structural analySiS discussing problems, solutions, idiosyncrasies, and applications of various analysis methods

I

t is no mystery that there are still uncertainties and lack of guidance when conducting nonlinear seismic analysis of structures. The challenge remains that the modeling, type of nonlinearities, and parameters required for analysis oftentimes vary from project-to-project and personto-person depending on the assumptions that are made in light of limited guidance. This article summarizes key points and relevant discourse from a panel session as a means of sharing information, advancing the practice, and closing the gap between research/development and practice associated with conducting nonlinear seismic analysis.

Motivation From the 2012 survey results compiled by the ASCE Subcommittee on Emerging Analysis Methods in Earthquake Engineering and published in Nonlinear Analysis in Modern Earthquake Engineering Practice (STRUCTURE, March 2014), four (4) major barriers for entry into nonlinear analysis were identified: 1) high complexity, 2) time consumption, 3) lack of clear guidance, and 4) communicating the benefit of advanced analyses to owners. The Subcommittee was led to charter more discussions and means to “close the gap” between research/development and practice about nonlinear seismic analysis. To further this effort, a panel discussion consisting of academics and practitioners was held at the 2015 Structures Congress in Portland, Oregon. Critical issues and challenges to date were discussed, and viewpoints were shared. This article summarizes some of the highlights from the session that may prove useful for structural engineers who are confronted with the challenge of conducting nonlinear seismic analysis, where several initial questions arise: • Does this design warrant advanced nonlinear analysis?

• Who will pay for it?

• Will the project finish on time due to the extended amount of time required to do such a complicated analysis?

• Is there confidence in the results produced based on assumptions made?

As structural engineers, we have a huge responsibility to society and the profession based on the designs we produce. As if that’s not a big enough charge, we know that these undertakings can be daunting at best.

Challenging Issues When Conducting Nonlinear Seismic Analysis

By Monique Head, Ph.D., Rakesh Pathak, Ph.D., P.E., Susendar Muthukumar, Ph.D., P.E. and Kevin R. Mackie, Ph.D., P.E.

Panel Discussion During the 2015 SEI/ASCE Structures Congress, seven (7) panelists from academia and industry gathered together to discuss some of the challenging issues facing the profession, such as

38 March 2016

modeling of nonlinear structural components, capturing geometric nonlinearities in response, pushover analysis, time history analysis, selection of ground motions, and how much modeling detail is enough to get reasonable results. The panelists, ranging in specific expertise centered on seismic analysis and design, were: • Ibrahim “Ibbi” Almufti, S.E., P.E., LEED AP, Associate at Arup • Finley Charney, Ph.D., P.E., Professor at Virginia Tech • Amir Gilani, Ph.D., S.E., Structural Specialist at Miyamoto International, Inc. • Walterio A. Lopez, S.E., Principal at Rutherford & Chekene • Weichang Pang, Ph.D., Associate Professor at Clemson University • Rafael Sabelli, S.E., Director of Seismic Design at Walter P Moore • James Daniel Dolan, Ph.D., Professor at Washington State University The general consensus for eight (8) major areas are summarized herein based on the following categories: 1) when to conduct nonlinear analysis, 2) the challenges, 3) justification to owners, 4) challenges for structural software industry, 5) nonlinear analysis validation, 6) need for more guidance, 7) pushover analysis and 8) “the future.”

When to Conduct Nonlinear Analysis All panelists unanimously pointed out that nonlinear analysis should be applied in situations where the building type is not regular or assumptions of code-based linear analysis are not valid anymore.

Everyone also agreed that analysis for retrofitting or presence of certain lateral-force resisting systems like viscous dampers, isolators, or any new type of lateralforce resisting system warrants nonlinear analysis.

The Challenges The primary point emphasized was the need to interpret results from advanced analysis, which requires experience and peer reviews. “The added cost and likely peer review time and expenses associated with nonlinear procedures may be a barrier of entry [Gilani].” This begged the next concern for determining when advanced analysis is even deemed necessary, especially given the processing time, time needed to interpret results and cost to do such analyses. The second challenge identified was the scaling of ground motions. All of these challenges were also identified from the 2012 survey, but rose as the top two challenges agreed among the panelists.

Justification to Owners Justification to owners is one of the starting points before an engineer can proceed to do nonlinear

seismic analysis in cases where it is not warranted. Each panelist had a slightly different perspective based on their experience and understanding. Almufti and Pang pointed out that simulated financial losses and downtime after a seismic event could be one of the motivation for the owners. Lopez indicated nonlinear analysis could provide potential savings in material/ schedule/etc. over the code prescribed linear procedures. Charney noted that justification may not be likely if the same building can be designed by satisfying all the code requirements. Sabelli and Gilani indicated that a lack of reliability of linear methods in certain situations could be the driving factor. Sabelli also indicated economy and design creativity for unconventional systems as justification for performing nonlinear analysis. Overall, everybody agreed that nonlinear procedures are time consuming and computationally demanding, and also require an added cost of peer review.

important to recognize, as it goes hand-in-hand with the need to interpret results. Moreover, further guidance was needed for validating models, assuming that the modeling results are reliable, while supporting documents from FEMA P-695 and ASCE 7/41 help with selection and scaling of ground motions [Pang]. Similarly, when following ASCE 7 Chapter 16 “where the system is to be modeled in 3D, subjected to 11 pairs of ground motions, and in cases where accidental torsion must be analyzed, this can increase to 44 pairs of ground motions with the use of scenario spectra as the target for ground motion

scaling, which can increase the required analysis by another factor of 2, 3 or 4…the time required to perform this analysis can be measured in hours or days for a single analysis for complex systems [Charney].” As such, future additions and enhancements to ASCE 7-16 and ASCE41 may (or may not!) be welcomed continuous improvement to aid advanced analyses.

Pushover Analysis “To do pushover analysis or not, that is the question!” The general impression from the

Challenges for Structural Software Industry

Nonlinear Analysis Validation This is the confidence building step which every engineer “must” go through during the nonlinear analysis. Unfortunately, there are no standard guidelines which one can follow, but only a set of “rules of thumb” derived from experience and fundamentals of structural analysis. One can start from viewing results like mode shapes, vibration periods [Pang, Charney], running sensitivity studies by varying time steps, changes in phenomenological definitions, hinge properties etc. [Gilani], validation of component behavior by comparing analytical and experimental results, matching initial conditions like dead and live load in columns [Lopez]. Gilani pointed out that it is due to the difficulty of such validation that the code requires peer review of design based on nonlinear analysis.

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More Guidance One new topic that surfaced from the discussion was the need for data management given the large amount of data produced when conducting advanced analyses. This is STRUCTURE magazine

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The general consensus to this question was that software must provide tools for efficient data management, post-processing and reduction in run time for nonlinear time history analysis.

panelists was that a pushover analysis should not be used as the sole measure and not be needed if one is performing a nonlinear time history analysis. It was noted that pushover analysis is less useful for drift or ductility demand, but rather to help proportion the structure to activate any intended ductile mechanism [Sabelli]. Furthermore, “pushover analysis is probably not appropriate for multi-mode buildings [Almufti].” Pushover analysis was also noted as “not being useful… do a response history, as collapse mechanisms are frequently misidentified, even for short buildings [Lopez].”

acceptance criteria for structural elements in new construction, since ASCE-41 is not intended for new construction [Lopez].”

While a fruitful discussion and exchange of information from experienced advanced analyses users took place, there is still more work to do to streamline this process and educate future engineers on advanced nonlinear analysis procedures. So the question still remains, do we “pay now or pay later?” For a complete listing of the panelists’ responses, please visit www.cece.ucf.edu/people/kmackie/ SEI-SEC/Panelist.html.

Acknowledgment

The Future Given recent discussions and even votes, the need for more education and training on advanced topics like nonlinear analysis cannot be overstated. Education, training, workshops, continuing education units, and other types of professional development are paramount. Training in school and other “proper training of engineers’, as noted by Graham Powell’s two articles in the November and December 2008 issues of STRUCTURE magazine, were reiterated by Lopez. Coupled with more knowledge would be the need for “better post-processing tools as well as

The findings and opinions presented herein are those of the authors and are not necessarily those of the American Society of Civil Engineers (ASCE). The support provided by the panelists, ASCE Subcommittee on Emerging Analysis Methods in Earthquake Engineering and Seismic Effects Committee is greatly appreciated. The ASCE Subcommittee on Emerging Analysis Methods in Earthquake Engineering and Seismic Effects Committee cordially invite all those interested in learning more about our activity and future pursuits to attend our meeting during the 2017 ASCE Structures Congress in Denver, CO.▪ ADVERTISEMENT–For Advertiser Information, visit www.STRUCTUREmag.org

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Monique Head, Ph.D., is an Associate Professor in the Department of Civil Engineering at Morgan State University in Baltimore, MD. She is the vice chair for the ASCE SEI Subcommittee on Emerging Analysis Methods in Earthquake Engineering. She can be reached at monique.head@morgan.edu.

Rakesh Pathak, Ph.D., P.E., is a Senior Research Engineer and program developer for bridge and building analysis products at Bentley Systems Inc. He is the chair for ASCE SEI Subcommittee on Emerging Analysis Methods in Earthquake Engineering. He can be reached at rakesh.pathak@bentley.com.

Susendar Muthukumar, Ph.D., P.E., is a Senior Engineer and Senior Associate with the Research & Development Group of Walter P Moore & Associates based in Austin, TX. He can be reached at smuthukumar@walterpmoore.com.

Kevin R. Mackie, Ph.D., P.E., is an Associate Professor and Associate Chair at the University of Central Florida, Orlando, FL. He is the chair of the ASCE/SEI Seismic Effects technical committee. He can be reached at kevin.mackie@ucf.edu.

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LINCOLN SQUARE

EXPANSION STEEL FIBER REINFORCED CONCRETE SOLVES SEISMIC DESIGN CHALLENGES

By Cary Kopczynski, P.E., S.E., FACI and Mark Whiteley, P.E., S.E.

B

ellevue, Washington continues to blossom into a vibrant, world class city. The Lincoln Square Expansion (LSE) broke ground in downtown Bellevue in June 2014 and, when complete in 2017, will add two 450-foot towers, a four level retail podium, and six levels of subterranean parking to Bellevue’s urban core. LSE is an excellent example of how innovative structural design can respond to high seismic requirements and still meet demanding architectural programs.

The 41-story hotel/residential tower will feature a W Hotel and upscale apartments. The 31-story office tower will feature Class-A office space with unparalleled views. Both towers integrate with a podium structure with retail shops and restaurants. The subterranean parking will include 2,200 new parking spaces and connect to adjacent existing underground parking via tunnels.

The hotel/residential tower is cast-in-place concrete with a mix of one-way and two-way post-tensioned slabs, with the office tower and retail podium framed in structural steel. Concrete shear walls resist seismic loads for both towers and the retail podium. The subterranean parking structure utilizes one-way post-tensioned slabs with wideshallow beams to create large open space and user-friendly parking.

Performance-Based Design Performance-Based Design (PBD) was key to the structural design of LSE. Essentially, PBD is a methodology for creating acceptable alternates to prescriptive building code requirements, contingent upon demonstrating that the proposed design meets code required seismic performance levels. This is generally accomplished through rigorous non-linear analysis and by checking the stiffness, ductility, and strength of critical elements.

Steel fiber reinforced concrete shear wall coupling beam under construction.

STRUCTURE magazine

Lincoln Square Expansion (LSE) in Bellevue, WA.

Since LSE consists of two towers and a common podium, careful attention was given to the seismic interaction between these structures. A combined nonlinear model consisting of both towers, the retail podium, and below grade parking was created. The stiffness assumptions for backstay effects between slabs and basement walls were important considerations, since they determined the effective fixity at the base of the towers. Two different sets of assumptions were made for these elements in the nonlinear analysis in order to effectively evaluate an upper and lower bound solution.

Steel Fiber Reinforced Concrete Reinforcing congestion has long been the bane of concrete construction in high seismic regions. Some of the most difficult and congested reinforcing is often in shear wall coupling beams. Traditionally, diagonal bars are used to reinforce these coupling beams, combined with tightly spaced stirrups and ties. This creates significant constructability problems since the diagonal bars clash with adjacent shear wall reinforcing.

For the LSE seismic system, PBD provided a means to implement steel fiber reinforced concrete (SFRC) in 341 of 392 total coupling beams. This is significant, since the use of SFRC for seismic design has heretofore been limited. The only prior use of SFRC in seismic coupling beams was in a 24-story tower in Seattle, for which CKC was also the structural engineer.

Modeling of key elements is critical to reliably predicting seismic behavior. The SFRC coupling beams in LSE were of particular importance. Both the initial stiffness and cyclic degradation used in the model were carefully calibrated against the hysteresis loops derived from dynamic lab testing. This hysteretic behavior was then used in the non-linear analytical models.

Dramix© steel fibers manufactured by Bekaert, with a fiber dosage of 200 lb/yd3 (120 kg/m3) of concrete, were used in LSE. The fibers are 0.015-inch (0.38mm) diameter by 1.18-inch (30mm) cold-drawn steel wire with hooked ends for anchorage. Fibers were delivered to the producer in subsets of thirty. The subsets were bonded with watersoluble glue that dissolved when mixed into the concrete, allowing the fibers to separate and disperse throughout the mix. After workability was confirmed at the site, a bucket was used to place coupling beam

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Diagonally reinforced concrete coupling beam (top) and SFRC coupling beam (bottom).

Analytical model calibration with physical testing.

concrete. A piece of “Stayform” at each end of the coupling beam restrained the SFRC from flowing into the adjacent core walls, which were pumped with high-strength concrete.

The study of SFRC started at the University of Michigan, with financial support from the National Science Foundation. Further research was funded by the National Science Foundation Network for Earthquake Engineering Simulation, and Bekaert, a Belgium based global supplier of steel fibers. The studies concluded that SFRC coupling beams without diagonal bars achieve similar or better performance, compared to prescriptively designed beams. With SFRC, the stiffness, strength and ductility of coupling beams is maintained, even though a significant quantity of reinforcing steel – including the diagonal bars – is eliminated. Use of SFRC in coupling beams can result in up to a 40 percent reduction in reinforcing and 30 percent net cost savings, compared to traditional coupling beam construction.

SFRC provides the structural engineering profession with a valuable tool for improving the constructability of reinforced concrete buildings in high seismic regions. The use of SFRC in LSE resulted in a coupling beam design that eased reinforcing congestion, facilitated faster construction, and reduced rebar tonnage.

Additional SFRC research is currently underway at the University of Wisconsin, funded by the Charles Pankow Foundation. The results of this testing are expected to broaden the range of fiber types, dosage rates, and aspect ratios available for use by designers of concrete buildings in high seismic regions.

Office Tower In the office tower, high-strength ASTM A913 Grade 65 structural steel was specified for the columns. This eliminated the need for custom fabricated shapes at the lowest levels where load demands exceeded the capacity of standard Grade 50 column shapes.

Mechanical, electrical, and plumbing (MEP) in the office tower was integrated with the structural steel framing. Penetrations in the steel framing allowed the mechanical systems to feed through the floor framing, which increased the floor-to-ceiling height and allowed for an additional leasable floor in the tower.

The office core was constructed of high-strength concrete, utilizing a high-production climbing form and scaffold system. Prior to starting structural steel erection, the core construction was advanced well ahead to allow simultaneous topping out of both the core and structural steel frame.

Hotel/Residential Tower

Foundations Other aspects of LSE are also noteworthy. For example, the hotel/ residential tower’s mat foundation, which was completed in February 2015 following the office tower’s mat pour a few weeks earlier, included 13,690 cubic yards of concrete. Both foundation mats utilized 80 ksi ASTM A615 reinforcing. A self-consolidating concrete mix was specified for the bottom 18 inches of the mats to ensure proper consolidation around heavy concentrations of rebar. The remainder was poured with a mix that was designed to minimize heat of hydration and temperature differential from the center to surface of the concrete, which was held to approximately 40 degrees F.

Subterranean Parking A system of one-way concrete slabs and wide-shallow beams was used in the subterranean parking structure to modulate with tower columns STRUCTURE magazine

above, creating column-free 50-foot spans throughout. The use of PT in both the beams and slabs lowered the floor-to-floor heights, resulting in reduced excavation and shoring depth. The garage layout partially carries through to the structures above, creating column-free space of 40 by 50 feet in the office tower. The hotel/residential tower was framed with a two-way flat plate system with spans of 25 to 34 feet. Parking columns and columns in the lower levels of the hotel/residential Tower utilized 14,000 psi concrete, specified with a test age of 90 days.

The combination of two-way PT slabs along with HSS columns, introduced to control deflections, allowed 14-foot cantilevers on both the north and south ends of the hotel with 8-inch thick PT slabs. The slabs were designed to allow stripping of formwork prior to installation of the HSS columns. For the residential levels above, which are set back approximately 60 feet, the slabs were designed with similar long spans and cantilevers.

Three foot cantilevered outrigger beams were used to lengthen PT slab spans and preclude the need for additional columns in the hotel/ residential levels. Sloped columns were used to eliminate transfer beams as the building transitions from the garage to the hotel levels.

Together, these elements eliminated all primary column transfers.

Additionally, a transition of large shear wall openings in garage levels to smaller openings in the hotel/residential corridors was implemented to maximize coupling of the tower shear walls while still accommodating wide drive aisle openings in the garage. continued on next page

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Structural/mechanical integration to increase floor-to-ceiling height.

To eliminate the need for concrete puddling at high strength columns, column concrete was poured continuous through the slabs. “Stayform” was first wrapped around the column reinforcing cage for the depth of the slab to restrain slab concrete from flowing into the column during the slab pour. Column concrete was then placed through the core during the column pour above, providing the required through-slab continuity. Added rebar through the slab/column joints supplements the slab load transfer through shear-friction.

Conclusion The Lincoln Square Expansion in Bellevue, Washington is as an excellent example of how innovative structural design can add value to a multi-use project. LSE both enhances the Bellevue urban landscape and advances the structural engineering profession. Through careful coordination among all parties, CKC’s unique solutions to the design challenges resulted in an efficient structural system, and use of SFRC created a new approach to high-rise building design and construction in high seismic regions.▪

Project Team Owner/Developer: Kemper Development Company Structural Engineer: Cary Kopczynski & Company Architects: HKS Architects / Sclater Architects Contractor: GLY Construction Geotechnical Engineer: Hart Crowser Cary Kopczynski, P.E., S.E., FACI, is Senior Principal and CEO of Cary Kopczynski & Company (CKC). Mr. Kopczynski serves on the Board of Directors of both the American Concrete Institute (ACI) and PostTensioning Institute (PTI), Chaired the PTI’s Technical Advisory Board for six years, and served on ACI Committee 318 for several building code cycles. Mr. Kopczynski can be reached at caryk@ckcps.com.

Mark Whiteley, P.E., S.E., is a Principal at Cary Kopczynski & Company (CKC), located in Bellevue, Washington, and the Senior Project Manager for Lincoln Square Expansion. Mr. Whiteley can be reached at markw@ckcps.com.

LSE under construction: steel office tower (left) and CIP concrete hotel/residential tower (right). Both towers will be 450 feet high.

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THE MODERN TEMPLE

MAKING A HISTORIC MONUMENT SHINE

By Brendan Ramos, S.E. and David W. Cocke, S.E., F.SEI, F.ASCE Figure 2. February 2015, Architect Carlos Posada from Gensler visits the Masonic Temple.

T

he renovation of a building offers an opportunity to rejuvenate a space and ensure the preservation and conservation of its history. The Masonic Temple in Glendale, California, once stood as the tallest building on Brand Boulevard and has housed six different Masonic organizations since its opening in 1928 (Figure 1). However, the building has been in decline since its founding Masons defaulted on their mortgage in 1934 when they lost control of the building through foreclosure. In recent times, the building was used for minor theatrical productions and theater storage, but it has not been utilized to its full original potential. Today, this historic landmark still towers over surrounding buildings, which include The Americana at Brand, a large outdoor shopping community across the street. Across town, on the west side of Los Angeles, recent construction in the Silicon Beach area (LA’s Silicon Valley) has provided a home to numerous innovative media and technology companies leading the modern creative office space trend. So, many have wondered, is it possible for an iconic building like The Masonic Temple to regain its glory and be relevant today?

Consisting of four full floors and a basement that extends beyond the building footprint, The Masonic Temple is approximately 64 feet wide, 125 feet long, and 105 feet tall. The floor area, including the basement, is approximately 52,500 square feet. The floors are two stories tall and within each floor is a mezzanine. The mezzanines are concentrated on the west side of the building.

STRUCTURE magazine

Figure 1. Circa 1927-1930, photo documenting 228-240 S Brand Blvd. in Glendale CA. Source: California State Library Picture Catalog.

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Figure 4. The Masonic Temple ETABS MT model.

Figure 3. October 1928, exposed steel frame before completion. Source: Masonry in Glendale: A Brief History.

Caruso Affiliated, one of the largest privately held developers in the country and based in Los Angeles, holds a passion for building commercial community spaces, such as The Americana at Brand and The Grove, as well as preserving historic structures. They partnered with the future tenant, CB Richard Ellis (CBRE); architect, Gensler; contractor, W.E. O’Neil; and structural engineer, Structural Focus, to bring a revolutionary state-of-the-art office concept to life. Transforming this historic building into a creative office space began with a creative tenant and a visionary architect (Figure 2), who, in turn, asked the structural engineer: “How do we get this done?”

At first glance, The Masonic Temple appears to be a purely concrete building. Its exterior concrete walls mask the true skeleton of the structure: a steel-framed building, as seen in Figure 3. The main girders, beams, and columns of the building are made of structural steel and typically encased in concrete fireproofing. The building is essentially a “concrete-dipped” steel structure. As with many other existing buildings, full documentation of the original construction was limited. Structural properties and reinforcing were determined or verified through testing, scanning and, in some cases, exploratory demolition.

The voluntary seismic retrofit of The Masonic Temple began with a structural assessment, identifying all the building deficiencies, utilizing the Seismic Evaluation and Retrofit of Existing Buildings standard sponsored and published by the Structural Engineering Institute (SEI) of the American Society of Civil Engineers (ASCE), also known as ASCE 41-13. This document is a tool that outlines a systematic procedure to evaluate building deficiencies and provides retrofit guidelines for the designer to follow for the building to meet the desired building seismic performance objective. For this building, the desired building seismic performance objective was Collapse Prevention. A structural model was built in ETABS to aid in understanding the behavior of the building (Figure 4). The major deficiencies identified included STRUCTURE magazine

vertical discontinuities (perimeter lateral force resisting elements did not continue through the basement), torsional irregularity (the west elevation’s lateral force resisting element was considerably “softer” than that of the east elevation), and unbraced mezzanines (several interior mezzanines were not adequately braced for lateral resistance).

After identifying the specific building deficiencies, the design team stepped back to look at the bigger picture. An unlimited number of retrofit schemes could have resulted in the destruction of the building’s historic nature, but the design team steered away from the more severe structural retrofits that would compromise the aesthetics and significant historic characteristics. The key to preserving the historical features of the building, while maintaining its structural integrity, was to implement the provisions of the 2013 California Historical Building Code (CHBC).

The Masonic Temple is listed in the Glendale Register of Historic Resources and, therefore, is subject to the governing code, 2013 CHBC. California is one of the few states with a code that provides specific regulations that govern the restoration, renovation, and conservation of qualified historical buildings. As stated in the Code, “The intent of the CHBC is to save California’s architectural heritage by recognizing the unique construction problems inherent in a historical building and providing a code to deal with these problems.” The target seismic performance of the CHBC is “collapse

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Figure 5. Rendering of interior office space. Courtesy of Gensler and Caruso affiliated.

March 2016

Figure 6. Construction photograph showing the modifications to the steel trusses needed to hang the rods.

prevention.” Under the umbrella of the CHBC, utilizing ASCE 41-13’s BSE-1E seismic hazard level (20% in 50 years) for collapse prevention as an aid to pinpoint the major deficiencies, a scheme was formulated to voluntarily increase the performance of the building.

A strategic retrofit scheme was implemented by understanding the building’s behavior and the major deficiencies that had been targeted. New shear walls were introduced in the basement directly below the existing east and west shear walls. A combination of new east elevation openings and strengthening of the west elevation piers shifted the center of rigidity of the building to reduce its irregularity. Moreover, a handful of interior mezzanines were demolished as they were no longer being used. A new interior shear wall was introduced to laterally support the remaining mezzanines. Overall, the CHBC’s load path requirement was met.

However, the structural retrofits focused only on the skeletal issues of the building. The structural scope that required the transformation of the building into a fully functional masterpiece was only half of the project.

The next step was to create an open office space that would meet the needs of the tenant. CBRE’s concept for a creative office space is unlike any other. The company is creating a new way of working with inspiring environments that exist beyond workstations (Figure 5, page 47 ).

“Floating” mezzanines with thin profiles and conference rooms on the uppermost floor were key to implementing the open office, which translated to a desire for no interior columns. Rather, large hanger rods supported by steel roof trusses above achieved the column-free space.

To accomplish this, the roof trusses above were analyzed using a twodimensional RAM Elements model for the additional loads imposed by the hanging mezzanine floor below. Supplementing existing angle truss members with new angles to create box sections strengthened compression elements. New steel plates increased the area of steel, increasing the tension capacity of tensile elements. Transforming double angles into “I” sections using steel plates increased the flexural capacity of truss members. Existing riveted connections were supplemented with welds to increase the connection capacity. As a result, the strengthened roof trusses, supporting exposed hanger rods and fireproofed with intumescent paint, provide the open concept aesthetic that columns could not offer (Figure 6).

Also, a two-story helical stair connecting three floors added to the structural complication for this tenant improvement. The nearly 15-foot-diameter stair was constructed of a 20-inch deep outer HSS stringer; it was first cut vertically in half, cambered to shape, then reattached to create the outer support. The tighter wound inner stringer consisted of a thick plate cambered to shape. A practical building element became a work of art uncommon in a typical workspace.

STRUCTURE magazine

Figure 7. Rendering of the Masonic Temple renovated. Courtesy of Gensler and Caruso affiliated.

Project Team Owner: Caruso Affiliated Structural Engineer: Structural Focus Tenant: C.B. Richard Ellis Architect: Gensler General Contractor: W.E. O’Neil The effective communication between the design team was critical for the timely delivery of a successful project. The Masonic Temple’s historic renovation was dictated by a fast-paced schedule that posed a challenge on a complex structural project. As in any historic project, surprises arose, and each unforeseen problem required a quick turnaround. While unforeseen conditions may be frustrating, the hidden elements aren’t always an unfortunate problem. For instance, while demolishing the existing balcony seating, a steel truss encased in the concrete balcony was uncovered. Guided by the owner’s desire to preserve the history of the building, the truss became part of the new design. The contractor’s thorough documentation of unforeseen conditions and the architect’s flexibility in design, coupled with the structural engineer’s out-of-thebox solutions are credited for the success of the project.

A design team with a common vision to preserve and rejuvenate The Masonic Temple delivered a now-relevant historic structure with the amenities of a modern creative office space (Figure 7). Long-neglected, The Masonic Temple has reclaimed its place in the heart of the city of Glendale.▪

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Brendan Ramos, S.E., is a Project Engineer at Structural Focus. He can be reached at bramos@structuralfocus.com.

David W. Cocke is the President at Structural Focus. He is an alternate member of the California Historical Building Safety Board and sits on the LA Earthquake Technical Task Force, as well as the Board of Directors of the Earthquake Engineering Research Institute and the Board of Governors at the Structural Engineering Institute of ASCE. He can be reached at dcocke@structuralfocus.com.

March 2016

INNOVATION I THE MEANS TO THE END

Courtesy of Brett Drury.

By Jay Love, S.E.

n 2007, Sutter Health brought together its Integrated Project Delivery (IPD) team to produce its new San Francisco flagship hospital, California Pacific Medical Center (CPMC). At the first team meeting, Sutter challenged the IPD team, consisting of SmithGroupJJR, the design sub-consultants, the general contractor, HerreroBoldt and its major subcontractors, to bring innovation to the project to provide the best hospital possible within the Sutter-established budget. Specifically, Sutter challenged Degenkolb Engineers as the Structural Engineer of Record to examine the various seismic force resisting systems available to select the highest performing system that provided the greatest value to Sutter. Working as a team, Degenkolb defined value as providing improved seismic performance at a lower cost on the same schedule. For this project, improved seismic performance was defined as: 1) Reduced floor inertial accelerations at the same interstory displacement.

2) Decreased inelastic demand in the primary structural columns and girders.

The new CPMC hospital at the Van Ness and Geary Street campus is currently under construction. The hospital, when finished in Q1 of 2019, will consolidate the acute care services from two older CPMC campuses whose older buildings must be replaced in accordance with the California Senate Bill 1953 Regulations that followed the 1994 Northridge Earthquake. The new hospital will provide Women & Infants and Adult Care services in eleven stories of programmed space.

Under the hospital, there will be two floors of parking. The central utilities plant is located on the top story, with HVAC equipment and emergency generators on the roof.

STRUCTURE magazine

The primary lateral force resisting system above grade is a welded steel moment resisting frame with a supplemental damping system.

Below grade, the perimeter concrete basement walls provide lateral resistance to the foundation. A total of 119 viscous wall dampers, or VWDs, provide the supplemental damping to the moment resisting system. Each floor above grade has a minimum of two dampers on two grid lines in each direction of the building. Additional dampers are installed on floors where the seismic response is greater, particularly at the mid-height stories of the building.

This system was jointly chosen by the IPD team after Degenkolb presented comparison designs for a conventional welded steel moment resisting system, a base-isolated system with a steel braced frame superstructure, and a damped steel moment resisting frame system. At this point in the process, the damped moment frame provided significant savings in steel material over both the conventional moment resisting frame and the based isolated system solution. In addition, the simplicity of the system was greatly preferred over the complexities of the moat system that would be necessary with the base-isolated solution. A third, but not inconsequential, consideration was that the viscous wall dampers could be strategically Figure 1. Schematic viscous located between the windows wall damper.

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on the exterior façade, providing unobstructed access to exterior light in the patient rooms.

Supplemental Viscous Damping Structural engineers practicing in earthquake engineering are accustomed to providing strength and stiffness to structures to resist dynamic loading and limit displacement. The common assumption is that damping is a constant, typically 5% of critical damping (although in some structures, the damping may be on the order of 2% or less.)

The standard equations of motion that govern the elastic analyses deal primarily with the interrelationship of first and third terms, mass and stiffness, to define the expected response of the structure, essentially ignoring the second, or damping, term. ma(t) + cv(t) + kx(t) = – mag Supplemental damping can increase the damping by a factor of four, and thereby decrease the required stiffness in order to achieve similar levels of interstory drift. This is especially significant when considering that the seismic design of steel moment resisting frames is frequently controlled by strict code-required interstory drift limitations, especially for Occupancy Category IV buildings such as hospitals. When designing a conventional moment frame structure with a 1% interstory drift limitation (ASCE 7-05, Minimum Design Loads for Buildings and Other Structures), the steel columns and girders are often much larger than required for strength in order to control the interstory drift. Supplemental damping can reduce the oversizing of the steel system yet still maintain strict drift limitations.

In the United States, supplemental damping has been used for seismic design, most commonly in conjunction with base isolation systems.

The supplemental damping in such systems provides some control of the lateral displacements at the isolation plane. These damping systems are commonly cylindrical fluid dampers that look much like car shock absorbers.

Viscous Wall Damper Development

Figure 2. Viscous wall damper in a moment resisting frame.

In 1992, viscous wall dampers were installed in the Sato Building in Tokyo. Today, the wall damper system has been used on more than 100 projects.

Dynamic Isolation Systems, Inc., well-known for its development of lead-rubber base isolators in the United States, had teamed with Aseismic Device Corporation, Ltd., a subsidiary of Sumitomo, to bring this technology to the United States. ADC, Ltd. was formed in 1996 to bring the various seismic force reduction technologies to market.

A typical viscous wall damper is comprised of a simple steel tank, or wall section, connected to the floor girder below, with a vertical steel plate or vane(s) that is inserted into the steel box and connected to the floor girder above. The vane is free to translate horizontally (Figure 2) through a polymer viscous fluid in the tank.

The viscous fluid, polyisobutylene, a synthetic elastomer, provides the velocity-related damping when the vane pushes its way through the fluid as the floors displace horizontally from one another. The elastomer fluid is non-toxic, odorless, non-flammable material with a viscosity of about 95,000 poise at room temperature. The damper output force, Fd, is proportional to the damping coefficient, Cw, and the velocity, v(t) raised to an exponent, a. The damping coefficient and velocity exponent are both experimentally determined. The wall damper force depends on interstory velocity, displacement and, to a lesser degree, temperature. For buildings such as hospitals where the internal temperature is maintained by sophisticated HVAC systems with backup power, the temperature dependence is relatively small.

Viscous wall dampers, shown schematically in Figure 1, were developed in Japan in the late 1980s by engineers at Sumitomo Construction Company, Ltd. (Arima, 1988). As part of the development, a fourstory full scale prototype test frame was built on a shake table in the Building Research Institute in Tsukuba City in order to compare the viscous wall damper system with conventional braced frames Analysis and Design structures, steel moment frame structures, and a lead-rubber bearing base-isolated system. The shake table tests showed a 50% reduction The analysis and design of the CPMC hospital were based on the in floor accelerations at the roof when compared to the conventional 2007 Edition of the California Building Code (CBC). The CBC refersystems. The tests also showed a 66% reduction ences ASCE 7-05 requirements for minimum in relative displacement compared to the condesign loads including gravity, seismic, and ventional moment frame. (The displacements wind forces. In addition to the conventional were similar between the viscous wall damped requirements found in Chapter 12 – Seismic system and the braced frame system, as would Design Requirements for Building Structures, be expected for stiff braced frames.) the design had to also comply with Chapter In addition to shake table testing with scaled 16 – Seismic Response History Procedures, earthquake accelerations, the viscous wall Chapter 18 – Seismic Design Requirements damped prototype building experienced four for Structures with Damping Systems, and real earthquakes between late 1987 and early Chapter 21 – Site-Specific Ground Motions 1988, with magnitudes varying from 4.5 to for Seismic Design.

6.7 at distances of 44 to 56 miles (70 to 90 Early analytic studies were based on perforkm). The engineers used these real-world earthmance data provided by Japanese engineers quakes to validate the results obtained from from ADC, Ltd. It was quickly apparent the shake table tests. that additional full-scale test data would be Figure 3. Production test rig in the DIS shop.

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Langan-Treadwell & Rollo (Langan) were used. Langan provided Maximum Considered Events (MCE) and Design Earthquake (DE) target spectra for both probabilistic and deterministic earthquakes.

Given the proximity to the San Andreas Fault, 6.8 miles (11 km) away, the deterministic event controlled the seismic design for this site. Using these spectra, Langan provided ten ground motion records that were scaled to closely match the MCE and DE target spectra over the range of periods for the building. The Peer Review Panel, as well as the California Geologic Survey (CGS), reviewed and approve the target spectra, record selection, and scaling factors for Degenkolb’s use.

Full Scale Prototype Testing Figure 4. Shipping dampers from the DIS plant in Nevada.

required to extend the Japanese data to larger system velocities and displacements.

Early in the design, the team met with the Office of Statewide Health & Planning Development (OSHPD), the Authority Having Jurisdiction (AHJ) over the design of acute care hospitals in California, to present the basic design concepts for a damped moment resisting frame system. As this system had never been used on a hospital in California before, nor for that matter in the United States, OSHPD raised various concerns. However, by the end of the meeting, OSHPD made it clear that the team would have to thoroughly demonstrate through testing and analysis that the new hospital would perform as well or better than a conventional California hospital.

An independent panel of three design professionals, required by ASCE 7-05, Chapter 18, was convened to peer review the site-specific seismic criteria, the preliminary design of the damping system, and the final design of the entire lateral force resisting system. As part of the preliminary design process, a project-specific Structural Design Methodology and Criteria document was developed to more fully describe how the design would meet the requirements of ASCE 7-05.

The Peer Review Panel reviewed and approved the Design Criteria development, providing valuable input during the process. In addition, the Peer Review Panel also reviewed test results from full-scale prototype testing programs described below.

In order to design the damped moment frame system, the team elected to use the nonlinear response history procedures, outlined in Chapter 18 of ASCE 7 and more fully developed in the Design Criteria document. The analysis was done with two 3-D nonlinear models using the PERFORM software package from Computers and Structure, Inc. The model included nonlinear elements to represent the Viscous Wall Dampers, girder connections and columns of the primary lateral system. Two models were used to bound the results of the analyses based on the expected properties of the wall dampers. The first model represented the upper bound damper properties, while the second model represented the lower bound properties of the dampers. The upper bound model generated more force in the dampers with less deformation in the girders and columns. The upper bound model results were used to check force-controlled elements such as collectors and columns subjected to high overturning forces. The lower bound model generated less force in the dampers, thus making the moment resisting frame more prominent in the force resisting system. The lower bound model was used to check the deformationcontrolled girder elements.

For both upper and lower bound models, a suite of ten ground motion records selected by the owner’s geotechnical consultant, STRUCTURE magazine

The first full scale testing of a 6-foot x 11-foot tall ‘pre-prototype’ damper took place at the UC San Diego Caltrans Seismic Response Modification Device (SRMD) Test Facility in May of 2008. This damper was put through a series of 26 tests, including in-plane sinusoidal tests and bi-directional earthquake response history tests.

Twenty tests were in-plane sinusoidal tests of generally three to five cycles at displacements ranging from 0.5 inches to 3.4 inches, the estimated MCE interstory displacement. By varying the input frequency, velocities from 0.7 inches per second to 15.3 inches per second were achieved. The six bi-directional earthquake response history tests were based on both Design Earthquake and MCE level response history results taken from Degenkolb’s nonlinear response history analyses for two different ground motions. The intent of these tests were to compare actual test results with the input motions developed from analysis models.

Using the results of the pre-prototype test program, the analytic damper model was evaluated and calibrated to continue with full analysis of the building. Given the varying floor-to-floor heights, three basic viscous damper sizes were selected, 7x9 feet, 7x10 feet, and 7x12 feet, to standardize the analysis as well as the fabrication processes.

(The 7-foot width dimension was not an arbitrary selection; because many of the wall dampers would ultimately be located on the building exterior, the engineers worked with the architect to determine the available space between bedroom windows, approximately 8 feet.

In doing so, the team successfully hid the dampers behind the solid portions of the exterior façade.)

Two additional sets of prototypes were tested at UC San Diego, including three 7- x 12-foot dampers and two 7- x 9-foot dampers.

Testing protocols similar to the pre-prototype test program were followed with addition tests at 1 inch per second to provide calibration data for future production testing at the DIS fabrication plant. These prototype type tests established the Target Force, F0 at 0 displacement and the Target Energy Dissipated per Cycle, or EDC. Both the Target Force and EDC values were calculated from the average of the last three cycles of a five cycle test sequence.

Viscous Wall Damper Production DIS manufactured and production-tested the viscous wall dampers at its Nevada facility. DIS designed and constructed a test rig (Figure 3, page 51) to expedite the final production testing approval process. To facilitate the cutting and welding, DIS constructed fabrication jigs in its shop that assure each damper met the precise tolerances necessary to obtain consistent performance.

The Production Testing Program was defined in the Design Criteria document to test a proportion of the total number of fabricated dampers. For each size damper, the first five production devices were tested at 1 inch per second to a displacement of 2 inches for five cycles to

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Courtesy of Brett Drury.

compare against the Target Values established in the Prototype Test Program. The allowable tolerance for production testing was +/-10% on the entire group of dampers and +/-15% on any individual damper.

If all five dampers met the target values within +/-15%, then only 50% of the next ten devices would be tested. If those devices met the target values, then only 40% of the next ten devices were tested and so on. Overall, with all tested production dampers meeting the Target Force and Target EDC values, 49 of the 119 (approximately 40%) production dampers were tested. On average, the production

dampers were less than 5% below the Target Force, F0, and were less than 5% above the Target EDC as established in the Prototype testing. The production testing was witnessed by an independent testing laboratory in the DIS shop.

After successful production testing, DIS shipped the wall dampers to the project steel fabricator/erector, The Herrick Corporation, in Stockton, California. While the fluid is very viscous, it is necessary to ship the dampers in the upright position (Figure 4), a practical shipping limitation to the maximum vertical height that can be fabricated and

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delivered on a project. Herrick took on responsibility for scheduling and shipping the dampers to the site in close coordination with the steel girders that connect to the dampers.

Site Erection Due to tight site conditions (See STRUCTURE magazine, December 2015 issue – Working at a Congested Urban Site.), Herrick shipped the dampers to the site “just in time” to coordinate with the primary frame erection. The tower crane lifted each damper, approximately 10,000 pounds apiece, to its final location on the frame, setting and connecting the damper to the bottom girder with high-strength bolts.

Conclusions Sutter originally challenged the team to bring value to its new hospital by finding innovative solutions. Given that the viscous wall damper system was developed over 25 years ago and used extensively in Japan, one might argue the team was not that innovative. However, the team needed a great deal of perseverance, technical excellence, attention to detail, and ultimately, support from the client to bring this system to the United States for the first time in a California hospital.

Was the team successful on this project? In a word, yes. Viscous wall dampers substantially decreased the floor inertial accelerations, especially the upper floors of the structure where seismic accelerations are typically greatest in conventional buildings. The viscous wall dampers saved a substantial amount of steel framing by controlling interstory drift. Based on the nonlinear analyses, the viscous wall dampers are expected to absorb nearly 90% of the earthquake energy

at the Design Earthquake level. Without viscous wall dampers, a steel moment resisting frame would have required 50% to 60% more steel in terms of tonnage, and more moment frames on more column lines in the building. Factoring in the cost of the viscous wall damper with the structural steel, the owner saved 25% of the cost of the structural steel system. And lastly, including the damper testing program, the review and approval process, and damper fabrication, the construction schedule was not compromised by the use of the dampers.

Is there a viable future for viscous wall damper systems in the United States? In several words, a qualified yes. Throughout this project, many people asked, “Why has it taken so long to bring this technology to the United States?” We can look at the slow implementation of base isolation in the United States for some similar impediments to implementation (Arendt, 2010). In order for this technology to gain greater acceptance in the U.S.: 1) Building owners, the decision makers, must place greater value on the seismic performance of their structures, 2) Engineers must understand the technology and its benefits to effectively demonstrate that value to the owner, 3) Relevant codes must be reviewed for improvements that would ease the use without reducing safeguards, and 4) Professional associations must promote dissemination of information for higher performance levels.

Accomplishing a first in the U.S. was not an easy path, but the results have made the journey worthwhile.▪ Jay Love, S.E., is the Structural Engineer of Record at Degenkolb Engineers. He can be reached at rjlove@degenkolb.com.

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www.buysuperstud.com/DeflectionClips or call 800.477.7883 US Patent 6213679; other patents pending

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STEEL

2016

Looking Good for Steel-Related Companies

By Larry Kahaner

W

ith a robust 2015 behind them, companies involved in steel construction look forward to an even stronger 2016.

“Business is strong, and 2015 was another recordbreaking year for SidePlate, both in number of new projects and overall contract value,” says Jason Hoover, Eastern Regional Business Manager, Industry Outreach Executive for SidePlate Systems, Inc (www.sideplate.com). “To keep up, we’re currently hiring engineers for our main office in California and a few other positions across the country.”

Terry Kubat, Partner/Software Engineer at IES, Inc. (www.iesweb.com) agrees. “We were blown away by our growth over the past two years. I think that a lot of structural engineers are starting to see IES as a new source of tools. Even though we’ve been in business since 1994, our software-quality has improved tremendously since about 2011 due to modern software development practices and the hiring of Adam Stordahl, P.E. Recently we’ve added another structural engineer, Garrett Drake, P.E. as well. Having smart, young professionals with more recent field experience is really helping IES better meet customer needs. We regularly see new customers adding IES tools alongside (or instead of ) some older or more historically popular products.”

Hoover says that his company’s latest innovation is the SidePlate Bolted Special Moment Frame, which is a slight variation on their field-bolted connection, which has been very popular over the past few years. “The biggest advantage of the SidePlate Bolted SMF is speed. There is no field-welding involved, so there are no weld inspections, and the beams literally drop into place. It could hardly be easier. We recently completed five full-scale lab tests that validated the seismic performance of the connection, and we’re starting to roll it out on high-seismic, R=8 projects.”

STRUCTURE magazine

He adds: “Many of our projects are now coming down to speed versus a few years ago where there was more of a focus on simply reducing construction costs. The SidePlate Bolted connection has really opened a lot of eyes and proven that we can do both: significant time and hard-cost savings. And for structural engineers, they’re continually being asked to do more for less. At no charge, SidePlate’s engineers offer a second set of eyes and can offer design options that SEs would otherwise be unable to offer their clients.” Says Hoover: “We continue to fight misconceptions that SidePlate is simply another 1:1 connection option, or that we manufacture something. The reality is that our designs improve the entire lateral system, and the connections themselves are built by any steel fabricator. We partner with the entire project team, from the structural engineer to fabricators and erectors, to ensure everything goes smoothly.”

IES’s Kubat notes that his company has a number of product upgrades in the works for 2016. “We are completely overhauling our flagship product VisualAnalysis to better take advantage of multi-processor architectures, parallel processing and background threading. The bottom line for customers will be a much improved interface for creating structural models, and running analysis and design checks. In addition, we are improving other products such as our free VARevitLink for two-way BIM communications with Autodesk’s popular Revit product. We are expanding our VAConnect tool for steel connection design that works stand-alone or integrates with VisualAnalysis.”

He would like SEs to know that IES has always tried to lead the way with user-interface improvements that greatly enhance an engineer’s learning curve. “It is extremely difficult for engineers to change software tools for a variety of reasons, such as time to continued on page 59

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March 2016

“ St��ct�ral desig� is what we do. IES tools help our engineers do it well. ” Structural Software for Professional Engineers IES VisualAnalysis

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STEEL evaluate and cost. We work very hard at IES to overcome these problems on all fronts: small-sized downloads, completely self-service free trials, and free technical support during your trial-period.

A typical engineer can get up and running with VisualAnalysis in less than 30 minutes, and be working productively on a real-world project with the trial product.”

Kubat concludes: “Sometimes I am amazed at the questions engineers will ask us – ‘Does VisualAnalysis do both lateral and gravity in the same model?’ Our response: Of course it does, and has for 22 years. Why would you want something different?”

At New Millennium Building Systems (www.newmill.com), they see a clear trend on learning to think like a building owner.

“Traditionally, this was the primary role of the architect, but increasingly it will be the role of every design-impacting participant on the project,” a spokesperson says. “This will be especially true on larger projects and multi-building campuses requiring blends of steel building systems.”

Why take the owner’s point of view? “Thinking like the building owner means thinking deeper about how to achieve the design intent while taking into account the cost of the total project, short-term and long-term,” the spokesperson says.

New Millennium now offers an expanded range of steel building systems. This includes cladding systems that bring a highly aesthetic range of architectural options, and exposed architectural deck ceiling systems that bring high esthetics, acoustic control, long-spans, and MPE integration. “We also now design, engineer and manufacture

“Thinking like the building owner means thinking deeper about how to achieve the design intent while taking into account the cost of the total project, short-term and long-term.” long-span composite floor systems. Having all these system options brings new meaning to value-add,” says the spokesperson. “We can offer the right system or systems based on lowest costs over the life of a project. When we are involved early on a project, we can show the pros and cons of the currently contemplated building system versus other system options, concrete or steel. If a project will benefit from a long span approach, we offer different steel building system options for this. We can help you look holistically at a range of ifthen cost scenarios. We can remove the risk of a more near-sighted value-added recommendation that can actually value-subtract from an owner’s total-project perspective.”

Marinos Stylianou, CEO at S-FRAME Software Inc. (www.s-frame.com), says that their Release 11.2 of the S-FRAME suite of products is being very well received by the structural engineering community.

“It delivered significant updates and new functionality to our analysis and design suite of products: S-FRAME, S-STEEL, S-PAD, S-CONCRETE, S-CALC, S-VIEW and S-FOUNDATION.” Adds Stylianou, “R11.2 includes important new functionality and many

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14-NMBS-18_struc-horiz.indd 1

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1/5/16 1:56 PM

STEEL enhancements designed to improve our client’s user experience in four key areas: First, connectivity to industry standards REVIT, AutoCAD, TEKLA, MS EXCEL, MS ACCESS. Second, the ability to automate repetitive tasks through a modern approach to automation using macros and the Python scripting language from within our new products. Three, tighter integration of analysis & design to further improve productivity gains by staying within one GUI for analysis, steel and concrete design; and fourth, additional advanced material models and analysis capabilities to handle the most demanding modeling requirements including partial releases and material failure which are important in performance-based design studies.”

He says: “We’ve also partnered with ADAPT Soft and provided our clients an integrated solution between S-FRAME’s S-CONCRETE and ADAPT’s Builder, and the results from 2015 are very encouraging. The oil and gas industry has been negatively impacted by lower oil prices, but our clients are well diversified and global. We’ve seen strong growth in the U.S. and Asia, especially from companies that chose to use our advanced structural analysis and design platform in particular for newer trend-setting tall buildings.” (See ad on page 4.)

“Business has never been better,” says Stuart Broome, Engineering Business Manager (USA) for Tekla (www.tekla.com), a Trimble Company. “There seems to be plenty of work out there and a shortage of engineers. Firms are more open than ever to look for ways to improve productivity and at Trimble we feel we are very well placed to provide a full solution, from concept through design to construction and operation.”

Broome says that Tekla Structures v21, was released in 2015, brings much more drawing and information management capabilities to their BIM solution. “Tekla Structures is well known around the world as being the most widely used and complete solution for steel and concrete detailing, but is less well known as a structural engineer’s tool for producing construction documents and general arrangement drawings. We’d like to change that.”

Version 21 included many new features specifically for engineers to make drawing production more efficient than ever before, says Broome. “Because all of the detail is contained in the actual model, there is no need for additional 2D line work – even dimensions and labels are automatically produced on the drawings. Improvements have been made to the Organizer tool to allow Engineers to create quick, custom reports automatically after creating the model.

Creating material take-offs and checking model accuracy has never been easier. Lastly, a change management feature has been added to enable engineers to keep track of what’s changing with each iteration of their own models or their client’s models, making tracking those changes a much less laborious task.”

The company last year also released their latest versions of TEDDS and Tekla Structural Designer (TSD). (See ad on page 3.)

“Although Lindapter (www.lindapter.com) has been pioneering steelwork connections for over 80 years, we are still coming up with innovative solutions for connecting steel, such as our Hollo-Bolt that allows a faster technique for connecting Hollow Structural Sections (HSS),” says Marketing Manager Wayne Golden. “We are delighted to announce that the Hollo-Bolt is the first and only expansion bolt for structural steel to receive full seismic approval from ICC-ES. The significance of this is that Structural Engineers STRUCTURE magazine

now have an alternative to drilling, bolting and welded HSS connections that is independently approved for use in all seismic design categories A through F.” Golden adds: “Hollo-Bolt is quickly and conveniently installed by simply inserting the fastener into a pre-drilled hole and tightening with a torque wrench, which ultimately saves the contractor time and money. This unique connection also provides the highest resistance to tensile loading in accordance with AC437, while ensuring compliance with the 2012 international building code.

The Hollo-Bolt has been used as a structural connection on projects all over the world with some impressive applications, such as securing the HSS framed roof at the Kimmel Center in Philadelphia.

However, since getting the full seismic approval, more and more structural engineers appreciate that the Hollo-Bolt is a viable connection solution for large-scale projects on the West Coast. We are now seeing the Hollo-Bolt used on major projects including the Wilshire Grand Center, Los Angeles and the ARTIC transit center in Southern California. There’s been a huge step-up in the use of Lindapter’s products, and it’s rewarding to see the fruits of our labor, especially when we put so much R&D into these products.”

At The Steel Network (www.steelnetwork.com), Nabil Rahman, Director of Product Development, wants SEs to know about StiffWall, a pre-engineered system intended to simplify and optimize the design and installation of strap bracing shear walls to resist wind or seismic forces and provide required lateral stiffness.

“The system eliminates the need for plywood, OSB, or steel sheet sheathed shear panels, all of which require excessive and complex fastener schedules. The system also eliminates the need for corner gusset plates traditionally used in strap bracing shear walls. In the StiffWall, the load path for shear forces through floor slabs is simplified by using corner boot connections and through bolts.

StiffWall has been effectively used in residential and commercial low and mid-rise cold-formed steel applications. The product is designed and manufactured to meet the performance requirements of each project.”

Says Rahman: “The StiffWall system is composed of panels where each panel connects two floors vertically. For a multi-story building, the number of panels for a single StiffWall system equals the number of stories. The panel consists of several structural components, which are the vertical end columns (vertical chords), the diagonal strap

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STEEL bracing, the corner boot connections, and the floor-to-floor through bolts. StiffWall Column/Boot Assemblies come preassembled for ease of installation.” He adds: “SteelSmartSystem software provides an intuitive, complete design and detailing solution for TSN’s StiffWall, and all your light steel framing design needs.” (See ad on page 83.)

At RISA Technologies (www.risa.com), Vice President, Operations, Amber Freund says they are continuing to hear from engineers that design projects and construction are growing. “We are hoping to see this growth continue in 2016,” she says.

“New for RISA-3D v14.0 is time history analysis capabilities.

This allows the user to enter their loading as a function of time and then analyze the structure over a time period to see how it reacts, including momentum and accelerations,” says Freund. “This feature is ideal for vibrating equipment on a structure. For years, RISA-3D has been used for industrial design, and engineers wanted to be able to look at the response of their structure to equipment vibration. The introduction of time history gives RISA-3D users the design tools they need to complete these industrial designs.”

As for trends, Freund notes: “We are still seeing BIM as a major topic with owners, architects and engineers. The implementation of BIM varies within the different construction industries and individual companies, but it is definitely being used more in the design process.” (See ad on page 84.)

Thomas Van Lann, CEO of CloudCalc, Inc. (www.clouldcalc.com) says that, because CloudCalc is a new concept, it will require

education before the engineering community fully understands the benefits it offers. “CloudCalc provides a cloud-based structural engineering software application. Like a lot of other engineering software, it can analyze steel structures for static and dynamic loads and check the results against the AISC ASD and LRFD codes.

But what differentiates us from the others is that CloudCalc is not tied to an individual PC. Because it runs in a browser (such as Chrome or FireFox), on any computer or Android tablet, an engineer can access, modify, and rerun his or her analysis from anywhere – a client’s office, a job site, or a hotel’s business center.

We expect to soon have a phone-based version which will let engineers check stress levels and make minor modifications whenever or wherever they wish.”

The concept for CloudCalc came from understanding customer needs. Says Van Laan, “Earlier in my career, I spent nearly 15 years as CEO of COADE, Inc., a very successful engineering software developer. During that time, our customers were very vocal in regards to their wish list: software that enabled collaboration among their distributed project team; device independent software that supported mobile devices, in order to permit more informed decisions in meetings or on site; flexibility to balance software licenses against the needs of cyclic project workloads; reduced dependence on IT support; shorter update cycles; and software that would appeal more to the app-driven lifestyles of the younger generation. With the expected generational shift-change expected in this industry,

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STEEL the Facebook-ers, tweeters, Instagram-ers of today will be leading the engineering projects of tomorrow.”

Tim Ellis, Simpson Strong-Tie (www.strongtie.com) Product Manager, says his company is offering their Strong-Rod Systems for seismic and wind design solutions for light-frame, mid-rise wood construction.

“Simpson Strong-Tie Strong-Rod continuous rod tiedown systems feature code-listed components and optimized rod-run assemblies, giving designers cost-effective and code-tested options for lightframe, mid-rise wood construction,” he says.

“The Strong-Rod Anchor Tiedown System for shearwall overturning restraint (ATS) and Strong-Rod Uplift Restraint System for roofs (URS) address many of the design challenges specifically associated with mid-rise buildings that must withstand seismic activity or wind events. These systems are designed to restrain both lateral and uplift loads, while maintaining reasonable costs on material and labor,” Ellis says.

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Historic structures significant structures of the past

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his is the first in a series on early suspension bridges. It starts with James Finley (STRUCTURE, November 2008), who designed and built the first iron chain suspension bridge (1801-1802) with a horizontal deck across Jacob’s Creek just south of Mount Pleasant, Pennsylvania on Old Route 119. Chain bridges with a deck resting directly on the chain had been built for years, but were only for pedestrian traffic. Finley’s bridge was designed for carriages and wagons, as well as pedestrians, cattle, etc. It was a small bridge with a central span of only 70 feet, a deck 12 feet 6 inches wide and towers 14 feet high. His chains consisted of hand forged wrought iron loops much like paper clips, except the bars were square rather than round in section. His towers were built of wooden timbers and continuous stringers, and a heavy wooden rail stiffened the wooden deck. Its suspenders were also wrought iron loops of varying lengths. The bridge, first of its kind anywhere in the world, was built for $600 with the cost shared by the two abutting counties, Fayette and Westmoreland. He warranted the bridge to last for fifty years. A local newspaper, dated May 22, 1802, wrote: “The Bridge, which Judge Finley (near this place) had undertaken to erect across Jacob’s creek, at the expense of Fayette and Westmoreland counties, near Judge Mason’s on the great road leading from Uniontown to Greensburg, is now completed. Its construction is on principles entirely new, and is perhaps the only one of the kind in the world. It is soley supported by two iron chains, extended over four piers, 14 feet higher than the bridge, fastened in the ground at the ends, describing a curve line, touching the level of the bridge in the center…

The projector has made many experiments to

Schuylkill Falls Chain Suspension Bridge (1809)

By Frank Griggs, Jr., Dist. M.ASCE, D.Eng., P.E., P.L.S.

Dr. Griggs specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He was formerly Director of Historic Bridge Programs for Clough, Harbour & Associates LLP in Albany, NY, and is now an independent Consulting Engineer.

Dr. Griggs can be reached at fgriggsjr@verizon.net.

ascertain the real strength of iron, and asserts that an inch square bar of tolerable iron in this position will bear between 30 and 40 tons; and, of course, less than one-eighth part of the iron employed in this bridge would be sufficient to bear the net weight thereof, being about 12 or 13 tons.”

Another bridge like it was not built for six years, when his plan was used for a replacement bridge in 1807 for Timothy Palmer’s bridge at the Little Falls on the Potomac River near Georgetown.

He was awarded a patent on the bridge on June 17, 1808 as patent no. X883. It was the seventh patent issued for a bridge following patents by Palmer, Burr and Pope. All records of the patent were lost in a fire in 1836, but a lengthy description of how he designed and built his bridges, published in The Port Folio in June 1810, was probably similar to his patent application. This article, in the writer’s opinion, was the finest piece of engineering literature of its time. He wrote of his bridge: “The bridge is solely supported by two iron chains, one on each side, the ends being well secured in the ground, and the chains raised over piers of a sufficient height erected on the abutments at each side, extended so slack as to describe a curve, so that the two middle joists of the lower tier [cross beams] may rest on the chains. The other joists of the same tier are attached to the chains by iron pendants of different lengths so as to form a level of the whole.

In order that the chain may support as much weight as it could bear, when hung with the weight attached to the end of it, the piers must be so high as to give the chain a sinking or curve of the one full seventh of the span. The ends of the chains must descend from the tops of the piers with the same inclination that they take inwards, until each end reaches the bottom of a digging, large enough to contain stones and other materials sufficient to counterbalance the

Image from The Port Folio for a 200-foot span that was part of a multiple span bridge like the Schuylkill Falls Bridge. Lower right and left indicate how Finley would connect his suspenders to the chain. The lower middle detail is at the top of the wooden tower showing the links connected to a saddle with pins.

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had built major bridges across the Hudson, Mohawk and Delaware Rivers. Finley was critical of Palmer’s Permanent Bridge, writing, “An estimate on these principles for a bridge of 500 feet between the abutments, with only one pier, will not amount to seven thousand dollars, exclusive of abutments and pier. Compare this with the Philadelphia Schuylkill bridge of the same extent, which cost sixty-five thousand dollars after the abutments and the two piers were completed; total expense, three hundred thousand dollars.” He also compared his bridge to two iron bridges in England, the

Coalbrookdale and Sunderland Bridges, and indicated how little iron his bridge, with iron in tension, used compared to those with iron in compression. As to stone, he wrote, “May I venture to glance at the grand, the majestic arch of solid stone, with any idea of contrast between it and our simple contrivance? Happy for me, utility, economy and despatch, are the ruling passions of the day, and will always take preference of expense, idle elegance and show, until the minds of men become contaminated with vanity or some worse passion.” He capped off his argument with, “It is remarkable that in

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weight of the bridge and what may chance to be thereon. The chains, if only one to a side, must be made with four branches at each end, to be let down through as many stones, and to be bolted below. These stones are laid flat on the bottom of the digging; other flat stones may be placed thereon, to bind and connect the whole, that they may have the same effect as a platform of one piece; four or more joists will be necessary for the upper tier [stringers] – to extend from end to end of the bridge – each will consist of more than one piece; the pieces had best pass each other side by side, so that the ends may rest on different joists on the lower tier. The splice will then extend from one joist to another of the lower tier, and must be bolted together by one bolt at each end of the splice…”

In his article, he indicated that his plan was not only used on the Schuylkill Falls Bridge but ones over the Potomac River at Cumberland and Georgetown, the Brandywine, two near Brownsville and one over the Neshaminy Creek north of Philadelphia.

A bridge across the rapids or Schuylkill Falls was authorized by the Pennsylvania Legislature on February 22, 1808 with Robert Kennedy and Conrad Carpenter as the proprietors. The only bridge across the Schuylkill River near Philadelphia was Timothy Palmer’s Permanent Bridge (STRUCTURE, October 2013) at Market Street, built in 1804. The falls were about five miles upstream from the Permanent Bridge. John Templeman, then working with James Finley as his exclusive agent, built the bridge across the river in 1809. In the Port Folio article Finley wrote, “In March 1808 I entered into an agreement with Mr. John Templeman of Georgetown, Maryland, by which he was to receive one half of all the monies arising from what permits or patent rights he could dispose of for and during the term of five years. All contracts to be in my name, and the money payable only to my agent in the city of Washington, who should pay one moiety over to Mr. Templeman.” His normal patent fee was $1 per foot of bridge. (Note: In the early 1800s, the term moiety was commonly used to denote a half.)

Finley must have had a hard sell to convince Kennedy and Carpenter to utilize a relatively untried design, when they could have contacted Timothy Palmer who had built major wooden bridges across the Merrimack, Schuylkill, Potomac and Delaware Rivers, or Theodore Burr who

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a science that has been maturing for thousands of years, and in which nothing is undertaken but by those who have been regularly brought up to the business, we should hear of so many misfortunes, and so much want of skill! Upon the whole, will it not be allowed that the best material has been chosen, (iron) the strongest and cheapest metal in the world – and applied in that way in which it possesses an hundred fold more power than it does in other positions?” Evidently his argument was successful, as he was given the contract to build the superstructure.

The actual span lengths are not clear. Finley wrote the bridge was 306 feet long with one pier. Templeman wrote, “The bridge…at the falls of Schuylkill has three spans, two of which are 153 feet long, and could have been extended much farther.” Svinin’s image seems to back up Templeman’s description.

Some sources indicate the spans were 200 and 100 feet, which makes some sense as Svinin’s drawing indicates the two suspension spans were of different lengths. Since the short span was of beam construction, maybe Finley did not consider it part of his chain bridge. Why they would build the extra pier so close to the abutment and why they made the spans of different lengths, if they did, is not clear from the record. The most likely dimensions are the ones given by Templeman.

The chains were made of 1½-inch-square iron bars wrought into links of between 8 (Finley’s panels were 8 feet long and the center panel was horizontal and 8 feet long) and 12 feet in length. The vertical suspenders were also loops made of 1½-inch square bars. The suspenders were attached to 5-by 10-inch wooden cross beams spaced 8 feet apart. The stringers probably were about 3 by 12 inches. The wooden deck was 2½ inches thick. Templeman further wrote, “the chain bridge at the falls of Schuylkill cost only 20,000 dollars including two abutments and two piers of hewn stone” and claimed “the chains and uprights of which will last a couple of hundred years, – the first being painted, and the latter covered with a roof – There will be of course only the flooring and railing of the Bridge, subject to decay – the renewal of which will amount to mere nothing.” In 1811, Finley also wrote a follow up 13 page pamphlet entitled, “A description of the chain bridge: invented by Judge Finley of Fayette County Pennsylvania with data and remarks illustrative of the power, cost, durability, and comparative superiority of this mode of bridging.”

In September, just after the bridge opened, some suspenders broke and a portion of the floor fell into the river. This was quickly repaired. The bridge then collapsed in 1810

Svinin – Schuylkill Falls Bridge Finley Design, built by Templeman, from Picturesque United States of America, 1811, 1812, 1913, Being a Memoir of Paul Svinin.

after only being up for about one year. The United States Gazette wrote “as a drove of cattle were passing across the bridge at the falls of Schuylkill, the works suddenly gave way and part of the superstructure fell into the river.” Finley, after inspecting the failure, and in order to maintain public confidence in his bridges, wrote a letter published in the January 17, 1811 United States Gazette in which he noted that the contractor had built the bridge contrary to his requirements.

The Finley statement is in part as follows: “Having recently been informed of some disaster befalling the chain bridges it becomes necessary for me to make a few cursory remarks, as to the cause of such failure.

The breach of the Schuylkill Bridge, by a drove of cattle, is an occurrence that deserves some attention. In giving a short explanation I invite the strictest scrutiny, and pledge my veracity for the correctness of the following statement.

And, the first thing to be observed is, that it was not a link nor any part of the chain properly so called, that broke; but an ill judged clip or coupling piece, with which two parts of the chain were joined together. Now it is indispensably necessary that these open hooks be removed, as well as some other improper substitutes, and let the connecting parts be as strong as any part of the chain itself; and two hundred tons burthen will make no considerable impression upon it. Great allowance must be made for the undertaker in this case; the principles being but little understood at that time; the workmen had never seen any thing of the kind, and had scarcely the shadow of information.”

Finley also followed up on Templeman’s March 14, 1809 Circular in a letter to the United States Gazette, in which he wrote: “But the chain bridge is a doubtful thing.

What then is the wooden frame bridge, consisting of two or three hundred tons of timber? View the upper and under framing

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– its hundreds of ties and bracings in every direction; what is the task of each and how much more can they bear? What burden can the bridge support independently of its own weight – or is not the huge mass of which it is composed a sufficient burthen almost for it; – In short the whole structure is so complex that nothing but loose conjecture can say any thing about it. Not so with the chain bridge, there are sufficient data for the strength and burden of its parts…let me further remark that, while the wooden bridge is the more in jeopardy the farther it is extended; the chain bridge on the contrary, becomes the more strong and secure the greater its extent.

(US Gazette, January 17, 1811)” The bridge failed again under a load of ice and snow on January 16, 1816. The United States Gazette again reported on the failure: “The Chain Bridge at the Falls of Schuylkill fell down about five o’clock on Wednesday morning. This unfortunate occurrence is said to have been occasioned by the great weight of snow which remained on it, and a decayed piece of timber. There was no person on the bridge when it fell.”

Some of Finley’s bridges lasted much longer.

Perhaps the best known was the bridge across the Merrimack River outside of Newburyport, Massachusetts. Timothy Palmer had built a wooden bridge across the river in 1792 (STRUCTURE, June 2013). By 1812 some decay had set in, and the long span of 160 feet on the southerly side of the river was replaced with a 244-foot bridge built by John Templeman. It had a partial collapse in 1827, but was rebuilt and used until 1907 when a lookalike bridge was built. Thomas Pope described this bridge in his 1811 book Treatise on Bridge Architecture, and brought it to the attention of the world.

The original Jacob’s Creek Bridge was replaced in 1833 after a life of 31 years.

While Finley’s bridges were unprecedented in terms of their design, they were a little ahead of their time. They were built by some men who did not understand Finley’s design, resulting in many failures. His designs, however, were picked up in England and France and modified. In England, Samuel Brown and Thomas Telford designed iron chain suspension bridges with horizontal decks in the 1820s. Their chains, however, were not built up of loops but were links connected by pins. Joseph L. E. Cordier, in his 1820 book Historie de la Navigation Intérieure, described the Finley Bridge. The Sequin Brothers in France picked up on the idea but used wire cables instead of iron chains. For his work, Finley is frequently called the Father of the Modern Suspension Bridge.▪

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CASE BuSinESS PrACtiCES

business issues

Advancing Technology How CASE Can Help Resolve the Friend or Foe Dichotomy

By Pedro Sifre, S.E.

T

hroughout history groups of people have created organizations to improve the lot of the members of the group and of society at large.

Factors that triggered the formation of such organizations vary. The medieval guilds were a response to the need to maintain a competitive edge and quality standards in the face of competition facilitated by increasing mobility within Europe. Ethnic-centric charities in the United States were providing mutual aid to recently arriving immigrant groups. Some of these still exist and spread their good works well beyond ethnic boundaries. Engineering societies have provided a way of establishing and maintaining standards and for disseminating knowledge among their members. The Council of American Structural Engineers (CASE) is one example of professionals joining together to address challenges posed by exercising our profession in an environment made more complex by, among other things, increasing litigiousness, galloping technology, and generational changes.

The rise of NCSEA, CASE, SEI and other organizations driven by the hard work of individuals is the best way to amplify voices and channel our energies to help manage the evolution of the profession and educate our young people with resources and knowledge sharing. CASE Tools and other CASE publications provide valuable resources that help polish our business practices and navigate ever increasing risks in this complex world. This article provides a few examples of stimulusand-response cycles that continue to motivate the work of CASE, and of how CASE can help us negotiate the evolution of our business practices.

Recently, the author was in a conference call concerning existing conditions at an old building where the concrete was deemed to be too weak for a new fall arrest anchor installation. On the line for the contracting side were the construction manager, the roofing contractor, the fall arrest subcontractor, and the subcontractor’s supplier and engineer. On the Owner’s side were the Owner, the Owner’s project manager (OPM), the facilities manager, and the safety manager.

On the design side, in addition to the author’s firm, were the architect, envelope

consultant, code consultant, and an OSHA rules consultant. From these conversations came word-by-word minutes to be edited by all parties and reviewed by who knows how many legal departments.

This is but one example of the complexity of relationships and communications that we navigate on a daily basis; a complexity that has been facilitated by the very technology that was supposed to make things easier. The phone call reference is a composite of real cases. But this type of conference call often has ten to twenty people on the line with more than 10 lines; something that would have been impossible 25 years ago. The callers saved the time and cost of travelling to a meeting, but somehow there was not a feeling of technology’s blessing that day. And the same was probably true of the people within earshot of countless speaker phones. Blaring speaker phones in cube farms are yet another unintended consequence of virtual meeting technology: a shortage of meeting rooms.

BIM platforms have created the benefits of three-dimensional modeling and the ability to visualize and coordinate multiple disciplines.

For structural engineers, the problem is that the modelling time is out of proportion with what is required for merely conveying design intent. With all their limitations, 2-D CAD drawings were such an efficient shorthand for conveying design information that most engineers still use BIM to create drawings that are facsimiles of CAD drawings. The problem comes with the increasing “clock speed” of revisions that architects can make. For structural engineers, we not only have to track the changes, but also redesign, alter section cuts and the calculations behind them, and then rearrange the annotations that accompany plans and sections.

As a benefit, BIM forces us into integrated thinking of the structure. For all but larger projects, it is now more difficult to establish assembly line work processes where one person is designing footings, another one is designing columns and developing the column schedule, etc. This is, of course, facilitated by design software. The down side is that a lot of entry level engineers used to gain valuable experience and insight from working on these assembly lines. And, the

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author argues, some of that feel is being lost by working with more automated platforms.

Couple that loss of insight with the greater “clock speed” of changes, and consider how they strain our management of design, analysis, and drawing production software Consider also challenges in training young engineers that have to enter a world in which the developments in the profession leave more work crammed into the same amount of time and the unrealistic expectations that technology creates.

Some of the unrealistic expectations are driven by technology. But another element is at play as well. The author calls it the increasing layering of project teams. Where there used to be a contractor and a few subs, now there is a construction manager and three times as many subs. Where there used to be an owner, there often is an Owner’s Project Manager (with the OPM having increasingly more consultants, especially on public projects). On design teams, we now see countless consultants. The layers add inefficiencies to communication which are complicated by more minutes and more email correspondence. The multi-layered project network is more fragile too. Consider how many times we are in a position that we have to answer a question instantly because not doing so is going to put somebody at a dead stop, which can set off a chain reaction of disproportionate delays. Immediate effects include 24-7 connectivity and endless RFIs, which have evolved from their original purpose of clarifying contract documents into numbered correspondence. How many people remember that the “I” used to stand for interpretation?

Another factor adding complexity to our work is the proliferation of uninsurable

Pedro Sifre (pjsifre@sgh.com) is a Senior Principal at Simpson Gumpertz & Heger, Boston, MA. He serves as a member of the CASE Toolkit Committee.

A listing and description of all CASE publications can be found on the CASE website, www.acec.org/case. All tools are free of charge for CASE member firms.

Tools are available to non-member firms for nominal fees. If you are interested in joining CASE refer to the website or contact Heather Talbert, htalbert@acec.org.

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contract clauses requiring us to accept defense obligations not tied to negligence, performance beyond the standard of care, and warranty language. It seems that the more seminars we attend alerting us to these clauses, the more we see them. I can imagine Owners attending similar seminars telling them the flip side of the argument. It is as if the “seminarization” of the profession has fueled a contract language arms race between owners and designers.

An element of technology – Powerpoint – has facilitated the “seminarization” trend.

Twenty years ago, developing a presentation with real slides was a monumental task limited to a few experts with access to production resources and time. The rest of us had to manage with overhead transparencies or “flimsies”. This and less expensive media have benefitted us by facilitating communications among structural engineers.

These woes of technology did not exist twenty-five years ago. Neither did the benefits, including this medium and this forum. For dealing with issues discussed in this article, there are multiple tools provided by CASE. For example, Tool 1-1 that helps manage risks and prevents claims; Tool 1-2 assists in developing a culture of quality; and Tool 3-4 for project work plans. The Tool 4-series provides communications tools that are useful for communicating within more complex teams. Tools 5-2 and 5-3 help manage the complexities of tasks and computation in the design office. The contract document tools in the Tool 9- series help ensure contract document quality and completeness. The CASE National Practice Guidelines offer a wealth of information that allows us to tackle assignments within the ever increasing complexity we have to navigate on a daily basis. CASE contracts and guidelines assist the engineer in developing scopes of work and agreements that protect the engineer and minimize risk.▪

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Nothing to install on your machine. Register for a free evaluation today.

Concrete Masonry Association of California and Nevada (CMACN) Phone: 916-722-1700 Email: info@cmacn.org Web: www.cmacn.org Product: CMD12 Design Tool for Masonry Description: Structural design of reinforced concrete and clay hollow unit masonry elements in accordance with provisions of Ch. 21-1997 UBC, 2001 through 2013 CBC or 2003 through 2012 IBC and 1999 through 2011 Bldg. Code Requirements for Masonry Structures (TMS 402/ACI 530/ASCE 5).

Decon USA Inc. Phone: 800-975-6990 Email: sales@deconusa.com Web: www.deconusa.com Product: Studrails Description: Our highly acclaimed Studrail design software has been updated and improved to incorporate the latest national and international design codes. An easy to understand user interface and clear graphics combined with a powerful calculation tool to produce a versatile platform for efficient and economical design. Free download at the website.

Product: Jordahl Anchor Channels Description: Allows a user friendly and safe calculation for anchoring in concrete with JTA anchor channels. The software features a technical and economical optimization of the design for each individual connection. 3D graphics are easy to use and allow a fast and clear input of all data. continued on next page

SOFTWARE UPDATES

news and information from software vendors

Design Data

Integrity Software, Inc.

Powers Fasteners

Phone: 800-443-0782 Email: sales@sds2.com Web: www.sds2.com Product: SDS/2 Software Description: Provides automatic detailing, connection design, and other data for the steel industry’s fabrication, detailing and engineering sectors. SDS/2 ‘s data sharing between all project partners reduces the time required to design, detail, fabricate and erect steel.

Phone: 512-372-8991 Email: sales@softwaremetering.com Web: www.softwaremetering.com Product: SofTrack Description: Reduce or eliminate your Bentley® Quarterly Overage Invoices. Control Bentley usage, receive alerts of idle usage including optional termination (to free licenses). Block unlicensed Bentley products from being used.

Phone: 845-230-7533 Email: Mark.Ziegler@sbdinc.com Web: www.powers.com Product: Powers Design AssistPowers Design Assist (PDA) Description: PDA now includes ACI 31811 and CSA A23.3 design provisions for mechanical, adhesive and cast-in place anchors.

Download or update to version 2.3 for free today at website to take advantage of the most current code standards.

Dlubal Software, Inc.

Phone: 323-592-3299 Email: EdLosch@gmail.com Web: www.LoschSoft.com Product: LECWall – Precast Concrete Column and Sandwich Wall Panel Design and Analysis Description: Prestressed and/or mild reinforcing. Flat, hollow-core or double tee configurations. Column design, handling analysis, multi-story capability, zero to 100 percent composite.

Phone: 267-702-2815 Email: info-us@dlubal.com Web: www.dlubal.com Product: RFEM Description: Dlubal Software is a worldwide leader in structural analysis and design software. Our highly sophisticated yet user-friendly programs with additional design modules will cater to each engineer’s individual project requirements. Seamless workflow is offered with BIM integration, accurate non-linear finite element analysis, and precise module design capabilities.

Hilti, Inc. Phone: 800-879-8000 Email: us-sales@hilti.com Web: www.us.hilti.com Product: Hilti PROFIS Anchor and PROFIS Rebar Description: PROFIS Anchor performs design calculations for Hilti post-installed anchor systems and cast-in-place anchors using the anchoring to concrete provisions of ACI 318.

PROFIS Rebar performs calculations for Hilti adhesive anchor systems and post-installed reinforcing bars using the development and splice provisions of ACI 318.

IES, Inc. Phone: 800-707-0816 Email: info@iesweb.com Web: www.iesweb.com Product: IES Building Suite Description: What if structural software installed in five minutes was simple, versatile and productive? What if it stayed out of your way during analysis and design? After 22 years, the IES Building Suite does it all: great value, simple licensing and 2-hour email support!

Solve your next problem in minutes.

Losch Software Ltd.

Pile Dynamics, Inc. Phone: 216-831-6131 Email: gbeim@pile.com Web: www.pile.com/pdi Product: GRLWEAP, PDI-TOMO, SPT Analysis Software Description: The revamped SPT Software, standard on SPT Analyzers and a Pile Driving Analyzer® option, quickly summarizes SPT hammer calibration results for reporting.

PDI-TOMO is new software used for crosshole sonic logging data 3D visualization and interpretation. GRLWEAP, the classic wave equation analysis of pile driving program, underwent a significant update.

POSTEN Engineering Systems Phone: 510-275-4750 Email: sales@postensoft.com Web: www.postensoft.com Product: POSTEN X Description: The most efficient & comprehensive post-tensioned concrete software in the world that, unlike other software, not only automatically designs the tendons, drapes, as well as columns, but also produces highly efficient, cost saving, sustainable designs with automatic documentation of material savings for LEED. The others simply Analyze – POSTEN DESIGNS.

Product: Powers Submittal Generator (PSG) Description: PSG is a submittal and substitution online tool that helps contractors create submittal packages in just a few steps and allows them to include all applicable code reports and technical details with a few clicks.

Contact us for a free demonstration!

RISA Technologies Phone: 800-332-RISA Email: info@risa.com Web: www.risa.com Product: RISA-3D Description: New for RISA-3D v14.0 is a Time History analysis capability. This allows the user to enter their loading as a function of time, and then analyze the structure over a time period to see how it reacts, including momentum and accelerations. This feature is ideal for vibrating equipment on a structure.

SCIA, Inc., a Nemetschek Company Phone: 410-207-5501 Email: info@scia.net Web: www.scia.net Product: SCIA Design Forms Description: Integrate custom checks into your FEA workflow. Script custom calculations that can run as standalone checks or link to SCIA Engineer’s FEA workflow.

Having the ability to write your own checks inside your FEA software is a real game changer. Try it for free!

Product: SCIA Engineer Description: Looking to migrate to or improve your 3D design workflows? SCIA Engineer offers an easy way to plug structural analysis and design into today’s BIM workflows. Tackle larger projects with advanced non-linear and dynamic analysis. Plug into BIM with IFC, and bi-directional links to Revit, Tekla, and others. Free demo!

All Resource Guide forms for the 2016 Editorial Calendar are now available on the website, www.STRUCTUREmag.org. Listings are provided as a courtesy. STRUCTURE® magazine is not responsible for errors.

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news and information from software vendors

Software UpdateS

Simpson Strong-Tie®

Structural Engineers, Inc.

Trimble

Phone: 800-925-5099 Email: web@strongtie.com Web: www.strongtie.com Product: Simpson Strong-Tie CFS Designer™ Software Description: Software for cold-formed steel designers automates product selection and helps navigate the complicated design provisions of AISI, while offering more robust design tools. The program has an upgraded user interface that makes input faster and more intuitive. CFS Designer is the new version of LGBEAMER software.

Phone: 540-731-3330 Email: tmmurray@floorvibe.com Web: www.FloorVibe.com Product: FloorVibe v2.20 Description: Performs calculations for all procedures in AISC DG11 Floor Vibrations due to Human Activity. Analyze floors for walking and rhythmic activities and for supporting sensitive equipment. Beam and steel joist data bases are included along with expert advice in real time. A demo version is at website.

Phone: 770-426-5105 Email: kristine.plemmons@Trimble.com Web: www.tekla.com Product: Tedds Description: Perform 2D frame analysis, access a large range of automated structural and civil calculations to U.S. codes and speed up your daily structural calculations.

Product: Simpson Strong-Tie Steel Deck Diaphragm Web App Description: Enables engineers to quickly and efficiently identify the best design and fastener solutions for steel decks given shear and uplift loads. The app, which is accessible from any web browser, provides diaphragm shear strengths of a steel deck when using Simpson Strong-Tie screws.

Phone: 610-280-9840 Email: sales@strumis.com Web: www.strumis.com Product: StruMIS Steel Fabrication Software Description: The complete management information and production system for every steel fabrication company; minimize overheads and costs, maximize productivity and profitability; in every step of the steel fabrication process.

StruMIS LLC

All Resource Guide forms for the 2016 Editorial Calendar are now available on the website, www.STRUCTUREmag.org. Listings are provided as a courtesy. STRUCTURE® magazine is not responsible for errors.

Product: Tekla Structural Designer Description: Software that helps engineering businesses win more work and maximize profits.

From the quick comparison of alternative design schemes to cost-effective change management and seamless BIM collaboration, Tekla Structural Designer can transform your business.

WoodWorks® Software Phone: 800-844-1275 Email: sales@woodworks-software.com Web: www.woodworks-software.com Product: WoodWorks® Design Office Suite Description: Designs perforated and segmented shearwalls; generates loads; rigid and flexible diaphragm distribution methods. SIZER: designs beams, columns, studs, joists up to 6 stories; automatic load patterning. New version coming this spring!

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InSIghtS

new trends, new techniques and current industry issues

Fastener Corrosion

By Mersedeh Akhoondan, Ph.D. and Graham E.C. Bell, Ph.D., P.E., FNACE

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asteners are typically low cost items; however, their failures may result in catastrophic and costly consequences. A recent case of bolt failure that has received enormous media attention was associated with the San FranciscoOakland Bay Bridge, the second busiest bridge in the nation. While investigations continue on the primary causes of bolt failures, corrosion was recognized as the foremost contributor. Corrosion is a naturally occurring phenomenon, the direct cost for which is a staggering $558 billion in the U.S. – approximately 3.1% of the U.S. GDP (2015).

Corrosion can be defined as the deterioration of materials (usually metals) by electrochemical interaction with their environment. Causes of corrosion initiation, rates of propagation, mechanisms of reaction and forms of corrosion products vary based on the type of materials, applications, and environmental exposures. Therefore, the risk of corrosion and preventive measures should be studied on a case by case basis. This article provides preliminary information on four common corrosion issues associated with ferrous mechanical fasteners (e.g., steel bolts and studs) and effective mitigation measures.

Additional information and specifications can be sought through NACE International (The Worldwide Corrosion Authority) and ASM International (Materials Information Society).

Top common forms of corrosion in threaded ferrous (steel based) fasteners are stress related corrosion (in the presence of cyclic or constant loading), hydrogen embrittlement, galvanic corrosion, and crevice corrosion.

Corrosion fatigue, in the presence of cyclic loading, and stress corrosion cracking (SCS), in the presence of constant stress, are known as the most common modes of failure for carbon and low alloy steel, as well as stainless steel alloys. Under tensile-stresses and exposure to corrosive environments (the presence of moisture and aggressive ions such as chloride), pits and cracks are formed within the microstructure of alloys reducing the expected yield strength. The higher the heat-treating temperature of the material (and the lower the ductility), the more susceptible it is to stress related corrosion. The first line of defense in controlling stress corrosion issues is to be aware of the issue in the design and construction stages. Controlling the environment,

which may involve removal of aggressive ions, the use of inhibitors, or protective coating (e.g. polymeric coating), are possible methods of alleviating such corrosion problems.

Hydrogen embrittlement (HE) occurs due to the ingress of free hydrogen into a metal’s microstructure, reducing ductility and the load-bearing capacity of the metal. The sources of hydrogen are encountered in the manufacturing processes of steel (e.g., pickling), fabrication (e.g., coating and plating), as a by-product of other corrosion reactions, and improper use of cathodic protection for corrosion protection. The failure, typically, initiates at the points of greatest stress concentration (e.g., for bolts and screws at the fillet radius under the head). It has been shown that the parts with hardness of Rockwell C36 or below are less likely to fail from embrittlement. HE is recognized as a contributing factor to the SCS. The HE of fasteners has received significant attention in recent years, following high profile failure cases, while studies continue on identifying causes of failures. The best practice is quality control of fasteners prior to use in critical structures. Care should be exercised in specifying high-strength martensitic low alloy bolts in hydrogen rich environments.

Galvanic corrosion, a.k.a “bimetallic corrosion”, occurs when two dissimilar metals come in contact in the presence of an electrolyte (moisture containing salt). A galvanic cell is created and the most active (anode) of the two metals corrodes to protect the least active (cathode) metal. The best practice is to use similar fasteners as the joining metal.

If not feasible, the surface area of the anode (least noble metal) should be greater than the cathode. It is a poor design practice to use carbon steel fasteners (least noble) in a stainless steel or copper assembly. Stainless steel fasteners (more noble metals) typically can be used in carbon steel assemblies, since the carbon steel will act as an anode. Application of non-conductive coatings on both metal surfaces has proven to reduce the severity of galvanic corrosion. Galvanic series tables are available as useful guides for compatibility of metals to be joined.

Crevice corrosion is a form of localized attack that most often occurs in narrow fissures where oxygen access is poor and a stagnant electrolyte solution is present.

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Galvanic corrosion.

Crevice corrosion.

Crevice corrosion is insidious and can rapidly progress, resulting in the attack of metal within the crevice areas (e.g., beneath bolts and cracks already initiated due to stress related corrosion). Possible mitigation methods are: avoiding formation of a crevice, sealing crevices with caulking compounds or elastic joint seals and painting surrounding cathodic surfaces, inspecting regularly for signs of crevices, and applying cathodic protection.

While corrosion resistance of fastener materials is important, the entire assembled joint should be considered to prevent corrosion issues. A careful evaluation of structural design, working stresses, expected service life, type of exposure, metallurgical characteristics of fasteners, and joint assembly is required to prevent future corrosion issues.▪ Dr. Akhoondan is a Corrosion EIT with HDR Engineering. She specializes in environmental compatibility of concrete reinforcing alloys and durability forecasting of infrastructures. She has designed and managed corrosion testing programs and performed failure analysis and metallographic evaluations of metal components.

Dr. Bell is Sr. Vice President and National Director for HDR’s Condition Assessment and Rehabilitation Business Class. He is certified by NACE International as both a Corrosion and Cathodic Protection Specialist, and was named a NACE International Fellow in 2015.

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

award winners and outstanding projects

Spotlight

Design on Stage: Goldsmith Theater

By Samuel Mengelkoch, S.E. and David W. Cocke, S.E., F.SEI, F.ASCE Structural Focus was an Award Winner for the Wallis Annenberg Center for the Performing Arts Goldsmith Theater project in the 2015 NCSEA Annual Excellence in Structural Engineering awards program (Category – New Buildings $10M to $30M).

C

ompleted in October 2013, the Wallis Annenberg Center for the Performing Arts project includes the restoration and seismic upgrade of the historic Beverly Hills Post Office building and the construction of the new 500-seat state-of-the-art Goldsmith Theater. The structural engineering effort was comprehensive and meticulous, requiring complex foundation work, careful restoration of the historic structure, ground-up design of the new theater, and precise coordination with the architect, acousticians, and mechanical engineers to ensure the highest acoustic performance.

In 2004, the Annenberg Foundation and other donors funded the preservation of the historic Beverly Hills Post Office building, and the creation of the Wallis Annenberg Center for the Performing Arts. Initial design concepts sought to place the main theater inside the historic structure. It soon became clear however, that in order to create a theater of sufficient size, along with the complex programming required to support it, the post office building simply was not large enough.

Esteemed Los Angeles architect Zoltan Pali and his firm, Studio Pali Fekete architects [SPF:a], began developing the concept for a new structure on the site to complement the historic building and expand the capability of the performing arts complex. SPF:a determined the best way to honor the historic post office building was not through mimicry or direct reference, but through a contemporary design inspired by the historic function of the site and the beauty of its traditional neighbor.

The Goldsmith Theater is sunken into the earth, making possible a full size, 60-foot fly tower without visually overbearing the adjacent two-story post office building or the Beverly Hills City Hall across the street. The entrance to the theater is at the same level as the basement of the post office building, and a grand stair leads guests from the main level of the post office down to the Goldsmith Lobby. In the Goldsmith Theater, the trap level underneath the stage is approximately 30 feet below street level. The foundation

design carefully considered the impact to and from a new subterranean parking structure built immediately to the east of the site, extending 40 feet below Crescent Drive.

Rising some 50 feet above street level, the new Goldsmith Theater fly tower was constructed of a hybrid steel and concrete system. Requiring very tall unbraced concrete walls, the engineering team at Structural Focus determined the most efficient design was achieved by allowing the walls to span between composite columns and beams. Wide flange members encased in concrete provided the composite behavior needed to withstand the large out-of-plane seismic forces generated within the walls. In addition, the internal steel ‘skeleton’ could be erected first, establishing a convenient frame to which formwork could be attached to receive the shotcrete walls.

Once the walls were sufficiently cured, shotcrete was applied on the reverse side to fully encase the beams and columns.

An essential design criterion for the project was the acoustic isolation of the theater stage and house, particularly from the massive mechanical equipment needed to condition the space. The team very much wanted to keep the HVAC units off the roof. Site limitations and complex programming restricted the team’s options, and the architect made the bold move of cantilevering multi-level mechanical platforms off the north and south ends of the fly tower structure, some 25 feet above grade. The engineers devised a simple steel frame structure with two-story diagonal tension struts to carry the weight of the equipment and platform back to the concrete fly tower walls. The performance space inside benefits from this dramatic gesture, and the cantilevers give the fly tower its signature form.

The creation of the Goldsmith as a separate building freed the post office building for versatile and unique programming. The extra space allows the Center to host educational events for young and aspiring performers alongside the world-class performances in the Goldsmith Theater. To connect the Center

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to the surrounding community and encourage casual visitors, a pedestrian pathway weaves through the site, between the new and historic structures, from Canon Drive east toward the stunning Beverly Hills City Hall just across Crescent Drive. An exterior sunken sculpture garden welcomes guests into the new Goldsmith lobby, and allows the Center to feature the artworks among vibrant landscaping and dazzling water features.

In a nod to the original function of the post office, the fly tower is wrapped in a skin textured to draw inspiration from thousands of hand-sorted letters, irreplaceable hand-written notes that brought people together years ago. The copper-colored façade complements the warmth of the post office, and invites and excites theatergoers at sundown.

The launch of the Wallis has electrified the community, and establishes a new benchmark venue for theatergoers in Southern California.

The pairing of the historic Post Office building with the strikingly modern design of the Goldsmith Theater is one of the most successful projects of its kind in Southern California.

Creative engineering solutions inspired by truly world-class architecture have built a landmark that Beverly Hills and all of Southern California will treasure for years to come.▪ Samuel Mengelkoch is an Associate with Structural Focus. He chairs the Image & Public Relations Committee, Structural Engineer’s Association of Southern California (SEAOSC).

David W. Cocke is the President at Structural Focus. He is an alternate member of the California Historical Building Safety Board and sits on the LA Earthquake Technical Task Force, as well as the Board of Directors of the Earthquake Engineering Research Institute and the Board of Governors at the Structural Engineering Institute of ASCE.

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News form the National Council of Structural Engineers Associations

NATIONAL

2016 NCSEA Membership Partnering Organizations CASE

SEI

Associate Members AISC

Insurance Institute for Business & Home Safety

American Wood Council

International Code Council

Bentley Systems, Inc.

Metal Building Manufacturers Assn.

Fabreeka International

Simpson Strong-Tie

Five Star Products

Steel Tube Institute

Affiliate Members AZZ Galvanizing

Independence Tube Corporation

Red Seat Software

Bekaert

ITW Commercial Construction North America

RISA Technologies

Lindapter USA

SidePlate Systems, Inc.

Blind Bolt Cast Connex Corporation Cold-Formed Steel Engineers Institute Construction Tie Products, Inc. DECON USA

Microsol Mitek Builder Products Nemetschek Scia New Millenium Building Systems

Design Data Headed Reinforcement Corp. (HRC) Hilti, Inc.

Pieresearch

SE Solutions, LLC Steel Deck Institute Steel Joist Institute Strand7 Tekla Vulcan Materials Company

Powers Fasteners

Sustaining Members Dunbar, Milby, Williams, Pittman & Vaughan

ARW Engineers Ballinger Barker Drottar Associates LLC Barter & Associates

Engineering Solutions, LLC Gerald E. Kinyon Gilsanz Murray Steficek

Bennett & Pless, Inc.

NCSEA News

NCSEA recognizes and thanks its Partnering Organizations and the following companies, organizations, and structural engineering firms for their Associate, Affiliate and Sustaining memberships in 20152016. For information on becoming an Associate, Affiliate or Sustaining member, contact Susan Cross at 312-649-4600, ext. 204, or email scross@ncsea.com. A listing of all of these members, including contact information, can be found at www.ncsea.com/members/more.

Blackwell Structural Engineers Burns & McDonnell

The Harman Group, Inc. The Haskell Company Holmes Culley

Cartwright Engineers

Mercer Engineering PC Morabito Consultants, Inc. NSA Construction Group, Inc. O’Donnell & Naccarato, Inc. Omega Structural Engineers, PLLC Ruby & Associates, Inc. Simpson Gumpertz & Heger Inc. Sound Structures, Inc.

Cowen Assoc. Consulting Structural Engineers

Joe DeReuil Associates

Criser Troutman Tanner Consulting Engineers

KBR

Structural Design Professionals, PLLC

Krech Ojard & Associates

Structural Engineers Group, Inc.

L.A. Fuess Partners

STV, Inc.

Land & Structure

TGRWA, LLC

LBYD, Inc.

Thornton Tomasetti

LHB Inc.

Wallace Engineering Structural Consultants

Jon Brody Structural Engineers

CTL Group DCI Engineers Degenkolb Engineers DiBlasi Associates, P.C. Dominick R. Pilla Associates DrJ Engineering

Mainland Engineering Consultants Martin/Martin, Inc.

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Stability Engineering

Wheaton & Sprague Engineering, Inc.

2016 Structural Engineering Summit

Disney’s Contemporary Resort · Orlando, Florida · September 14th-17th STAY CONNECTED BY USING THE HASHTAG #NCSEASummit

April 28, 2016 Fire Damage and Post-Fire Assessment of Structural Wood Members

Brian Kukay, Ph.D., P.E., Associate Professor, Montana Tech ST RU

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Diamond Reviewed

More detailed information on the webinars and a registration link can be found at www.ncsea.com. Subscriptions are available! 1.5 hours of continuing education. Approved

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50 states through the NCSEA Diamond Review Program. www.ncsea.com.

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NCSEA April 5, 2016 Seismic Design of Special Reinforced Masonry Shear Walls Gregory R. Kingsley, Ph.D., P.E., PEng, President & CEO, for CE credit in all KL&A Inc., Structural Engineers and Builders

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Chris Kimball, P.E., S.E., Kimball Engineering

The NCSEA Lateral SE Exam Review Course assists engineers in preparing for the SE Exam and includes instruction with an emphasis on building design: • Including sessions on exam strategy and bridge design • Key topics of structural code • Efficient analytical methods • New material in the 16-hour Structural exam • Typical exam questions • Problem-solving techniques • Exam day skills • 24/7 playback within a 6 month period – study anytime There are significant discounts available for groups taking the course. Log on to www.ncsea.com under Education, for more information and to register for the course. Online access to the vertical course held in February is also available.

STRUCTU

March 22, 2016 IBC Chapter 17: Special Inspections and Testing

NCSEA Lateral SE Exam Review Course set for March 19-20

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NCSEA Webinars

Kent has worked on a total of eight theme parks, from the beginning of design through to construction, in addition to many individual attractions. He is a licensed structural and civil engineer in California and a licensed professional engineer in Florida. Kent also has a Ph.D. from the University of Southern California and has authored several technical papers.

The NCSEA Structural Engineering Summit draws together the best in the structural engineering field and looks forward to welcoming Kent Estes as its keynote speaker. In addition, the Summit will feature technical educational sessions specific to structural engineering, nontechnical business sessions, social and networking events, the NCSEA Excellence in Structural Engineering Awards, and the trade show.

News from the National Council of Structural Engineers Associations

NCSEA is pleased to announce that Kent Estes, Ph.D., S.E., Walt Disney Imagineering, will be the keynote speaker for the 2016 Structural Engineering Summit, September 14-17, at Disney’s Contemporary Resort in Orlando. His presentation, Structural Engineering for Walt Disney Theme Parks, will provide a glimpse into the process of designing Disney theme parks, with a specific focus on structural engineering. Since 1952, Walt Disney Imagineering has been responsible for the design and construction of all Disney theme parks, starting in California, followed by Florida, and now featured in France, Japan, Hong Kong, and its newest location, Shanghai, China.

Estes recently completed nearly four years in Shanghai, where he collaborated closely with the Chinese design institutes and construction companies. His presentation at the Summit will discuss Disney’s unique construction practices, local building infrastructure, and building codes, as well as the varying soil conditions and mitigation measures. In each locale, the constants of safety, Disney quality, and guest experience must be assured at the highest level.

After the introductory discussion, the presentation will highlight and discuss many attractions and facilities, from the 1950s for the original Disneyland in Anaheim, California to the present construction just completed in Shanghai, China. Theme parks incorporate a variety of facilities, including retail, show elements, restaurants, theaters and rides; and there is a definite overlap between static buildings and moving, vibrating ride systems that must be considered. Perhaps the most unique aspects of the structural engineering are the vibration and fatigue considerations for the ride facilities. At the same time, there are many unique and iconic structures that serve the purpose of transporting guests to other times and places.

Kent Estes has been in the practice of structural engineering for more than 40 years. For over 30 of those years, he worked on Disney theme parks on three continents in the architectural and engineering arm of The Walt Disney Company, Walt Disney Imagineering. As mentioned above, he has just moved back from Shanghai, China. His role there was as the Design Manager for the Main Entry area where he managed a team of designers, architects and engineers. In addition, he was the Lead Structural Engineer for the park, hiring and overseeing a team of local Chinese structural engineers.

NCSEA News

Disney engineer to present keynote at 2016 Summit

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CALL FOR PROPOSALS NOW OPEN

Structural Columns

The Newsletter of the Structural Engineering Institute of ASCE

2017 Structures Congress April 6 – 8, 2017, Denver, Colorado Now accepting individual paper and complete session proposals for consideration.

Structures Congress 2017 is a forum to advance the art, science, and practice of structural engineering. The SEI National Technical Program Committee (NTPC) is seeking proposals for complete sessions and abstracts for individual papers to be presented at Structures Congress 2017.

Criteria used by the NTPC to review and evaluate the proposals include – but are not limited to – the following: • Technical Content • Presentation Quality • Applicability to the audience, i.e., what the audience will take away from the presentation To submit a proposal for consideration at Structures Congress 2017, visit www.structurescongress.org. The conference website will have detailed information and step by step power points to assist you.

Abstracts and Session Proposals should focus on topics and subtopics consistent with the list below. A complete list of subtopics is available at www.structurescongress.org:

Major Topics • Blast and Impact Loading and Response of Structures • Bridges and Transportation Structures • Buildings • Business and Professional Practice • Education • Forensic • Natural Disasters • Nonbuilding and Special Structures • Nonstructural Systems and Components • Research • Sharing Claim Experiences SEI encourages submissions from practitioners, educators, researchers, structural engineers, bridge and building designers, firm owners, codes and standards developers, and others.

All approved presenters are required to register and attend the conference. Please make sure this is possible before you make a submission.

Final papers are optional but strongly encouraged. Final papers will be evaluated by NTPC. Final papers that meet all the necessary format, copyright, permissions and content requirement will be published in the conference proceedings.

The due date for abstract and session proposals is June 2, 2016. Visit the congress website, www.structurescongress.org, to submit your proposals.

Questions? Contact Debbie Smith at dsmith@asce.org or 703-295-6095

Local Activities National Capital Section Structural Committee

Maryland Chapter

The ASCE National Capital Section is currently looking for individuals interested in the newly developed Structural Committee. The Structural Committee is intended to offer those with an interest in structural engineering the opportunity to network with other professionals, to sharpen their technical and professional skills and to grow professionally.

Those interested in becoming a participant can add their name and contact information to a list being developed www.asce-ncs.org/index.php/committees/structural. For any additional questions or inquiries contact Paul Parfitt at pparfitt@wje.com.

Errata SEI posts up-to-date errata information for our publications at www.asce.org/SEI. Click on “Publications” on our menu, and select “Errata.” If you have any errata that you would like to submit, please email it to Jon Esslinger at jesslinger@asce.org.

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The SEI Maryland Chapter recently hosted two speakers who addressed timely topics in structural engineering at the chapter’s general meetings. In December, Chiara Rosignoli of ParsonsBrinkerhoff presented a talk about the Dynamic Response of Simple Span Bridges. January’s speaker, Troy Brown of the Michigan State University College of Law, spoke about Comparative Ethics in Engineering. For more details see the full story on the SEI News page.

Get Involved in SEI Local Activities Join your local SEI Chapter, Graduate Student Chapter, or Structural Technical Groups (STG) to connect with colleagues, take advantage of local opportunities for lifelong learning, and advance structural engineering in your area. If there is not an SEI Chapter or STG in your area, talk with your ASCE Section/ Branch leaders about the simple steps to form an SEI Chapter.

Visit the SEI website at www.asce.org/SEI and look for Local Activities Division (LAD) Committees.

March 2016

Wood Education Committee Activities The Wood Education Committee has organized a ten session mini-symposium for the 2016 World Conference on Timber Engineering, August 22 – 25, 2016, in Vienna, Austria. The symposium focuses on initiatives and higher education activities related to wood design. Committee members Peggi Clouston, Michelle Kim-Baron, and Mikhail Gershfeld are the organizers of this mini-symposium. To learn more, visit the conference website at: http://wcte2016.conf.tuwien.ac.at.

In addition, the committee has proposed several ASCE Continuing Education webinars over the past year. On March 7, 2016, the first webinar, addressing the special considerations in design of large wood diaphragms, will be presented by John Lawson, C.E., S.E., M.ASCE. John has been researching this subject for many years and has proposed significant changes to the design philosophy and procedures. Learn more and register for this webinar at: www.asce.org/continuing_education.

Fire Protection Committee Call For New Members

SEI Futures Fund Call For art science.

Proposals – Due June 1

MAKE YOUR MARK GO THE DISTANCE

The job of a structural engineer and is both an

Structural engineers design the buildings where we live, work, go to school, and play, and the bridges we cross everyday.

As buildings reach greater heights and bridges span further distances, structural engineers must design these structures with materials such as steel, concrete, masonry, and timber to resist all forces. These forces include gravity, earthquakes, hurricanes, explosions and much more. All of this is considered to create the architectFutures and Fund client’s vision while creating a safe place for the public.

Investing in the Future of Our Profession

The SEI Futures Fund (SEIFF) invites proposals for new initiatives in line with SEIFF strategic areas that benefit the structural engineering profession and/or SEI as a whole, Makebe your mark by and would not otherwise funded outvisiting of SEI Division or www.ncsea.com and www.asce.org/SEI operating funds.

Strategic areas: • Promote student interest in structural engineering • Support younger member involvement in SEI activities • Enhance opportunities for professional development • Invest in the Future of the Profession Guidelines: • Proposed activities must benefit the structural engineering profession and/or SEI as a whole, and would not otherwise be funded out of SEI division or operating funds.

• Proposed activities must be completed in the year funded, and may not include standing committee operating expenses.

• It is recommended that proposals be vetted through the Committee or Chapter’s Executive Committee as appropriate.

For more information and a detailed proposal format, please contact Suzanne Fisher at sfisher@asce.org. Proposals are due June 1, 2016. Visit www.asce.org/structural-engineering/ sei-futures-fund to donate.

STRUCTURE magazine

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Make Your Mark Poster Now Available Celebrate structural engineering and inspire pre-college students to pursue structural engineering as a career. Include the Make your Mark poster in your pre-college outreach. The poster is produced by the National Council of Structural Engineers Associations (NCSEA) and SEI. The latest poster features a photo of a bridge designed by Santiago Calatrava in Reggio Emilia, Italy. Complimentary posters are available upon request to Suzanne Fisher at sfisher@asce.org. Be sure to include the number of posters you are requesting and where they should be sent. For more resources and ideas for outreach with young students, visit the ASCE Pre-College Outreach webpage at www.asce.org/pre-college_outreach.

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March 2016

The Newsletter of the Structural Engineering Institute of ASCE

The SEI Fire Protection Committee is the leading U.S. committee on structural fire engineering. This committee has developed and successfully balloted a new section for the 2016 edition of the ASCE/SEI 7 standard entitled Appendix E: PerformanceBased Design Procedures for Fire Effects on Structures. Appendix E provides requirements for structural fire engineering analyses, and specifies mandatory (occupant life safety) and discretionary (e.g., limitation of damage) performance objectives. Appendix E is currently open for public comment. As a companion design guideline for Appendix E, the Committee is currently developing ASCE/SEI Guideline: Structural Fire Engineering. This guideline will provide recommendations for structural fire engineering analysis techniques and input parameters, as well as design example cases. The committee is currently seeking enthusiastic new members to contribute to the guideline. If you are interested, please contact the committee chair, Kevin J. LaMalva, at kjlamalva@sgh.com, or apply at www.asce.org/ structural-engineering/sei-tad-committee-application.

Structural Columns

Committee Activities

CASE in Point

The Newsletter of the Council of American Structural Engineers

JUST RELEASED: Updated Special Inspections Guideline CASE has updated and released CASE 962-C: Guidelines for International Building Code – Mandated Special Inspections.

The updates to this guideline include: • Bringing it current with the requirements of the 2012 International Building Code (IBC) • Describing the roles, responsibilities and qualifications of the parties involved in the special inspection and testing process • How to prepare and conduct a special inspection and testing program • Who pays for the special inspections and test

CASE Tools are developed and released with the sole purpose of ensuring CASE members manage risk and safety when engaging in structural engineering projects. We encourage you to download them and incorporate them into your business.

You can purchase this and the other Risk Management Tools at www.acec.org/bookstore.

Follow ACEC Coalitions on Twitter – @ACECCoalitions.

WANTED Engineers to Lead, Direct, and Get Involved with CASE Committees!

If you’re looking for ways to expand and strengthen your business skillset, look no further than serving on one (or more!)

CASE Committees. Join us to sharpen your leadership skills – promote your talent and expertise – to help guide CASE programs, services, and publications.

We have a committee ready for your service: • Risk Management Toolkit Committee: Develops and maintains documents such as business practices manuals and policies for engineers under CASE’s Ten Foundations for Risk Management.

Please submit the following information to htalbert@acec.org • Letter of interest • Brief bio (no more than 2 paragraphs)

Expectations and Requirements To apply, you should • be a current member of the Council of American Structural Engineers (CASE) • be able to attend the groups’ two face-to-face meetings per year: August, February (hotel, travel reimbursable) • be available to engage with the working group via email and conference call • have some specific experience and/or expertise to contribute to the group Thank you for your interest in contributing to your professional association!

CASE Risk Management Tools Available Foundation 3 Planning – Plan to be Claims Free Tool 3-1 A Risk Management Program Planning Structure This tool is designed to help a Firm Principal design a Risk Management Program for his or her firm. The tool consists of a grid template that will help focus one’s thoughts on where risk may arise in various aspects of their engineering practice and how to mitigate those risks. Once the risk factor is identified, then a policy and procedure for how to respond to that risk is developed. This tool contains 10 sample risk factors with accompanying policies and procedures to illustrate how one might get started. The tool is designed to insert custom risks and policies to tailor it to individual firms.

Tool 3-2 Staffing and Revenue Projection Firms are provided a simple to use and easy to manipulate spreadsheet-based tool for predicting the staff that will be necessary to complete both “booked” and “potential” projects. The spreadsheet can be further utilized to track historical staffing demand to assist with future staffing and revenue projections.

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Tool 3-3 Website Resource Tool This tool lists website links that contain information that could be useful for a Structural Engineer. A brief description of the website is also included. For example, there is information about doing business across state lines, information regarding the responsibility of the Engineer of Record for each state, links to each State’s Licensing Board, etc.

Tool 3-4 Project Work Plan Templates Preparing and maintaining a proper Project Work Plan is a fundamental responsibility of a project manager. Work Plans document project delivery strategies and communicate them to the team members. Project Managers will use this template to create a project Work Plan that will be stored with the project documents.

You can purchase these and the other Risk Management Tools at www.acec.org/bookstore.

March 2016

Alternative Delivery Systems (ADS) Optimizing Your Firm to Win More Work with Less Risk March 24, 2016; Washington, D.C.

ACEC Annual Convention April 17-20, 2016 Washington, DC Top 10 Reasons to Attend the 2016 Annual Convention and Legislative Summit 1) Former White House Press Secretary and Co-Host for Fox News’ daily hit show “The Five” – Dana Perino on Political Outlook & the 2016 Elections 2) New Panel on State Infrastructure Funding – Insights from Former DOT Secretaries with Barry Schoch, Former PennDOT Secretary, McCormick Taylor; Paula Hammond, Former Washington State DOT Secretary, WSP|Parsons Brinckerhoff; and Ananth Prasad, Former Florida DOT Secretary, HNTB 3) Leadership and Influence with Daniel Pink, Worldrenowned business thought leader and author of three New York Times bestsellers 4) Congressional Issues Briefing and Panel on Prospects for Bipartisanship – moderated by Major Garrett, CBS News Chief White House Correspondent 5) NEW Panel on How Big Data Will Transform Your Company – Moderated by John Doering, President, J. Doering & Co., LLC; Michael Corkery, President/CEO, Deltek; Gary Hall, Chief Technology Officer, Cisco; Capt.

Jay Bitting, U.S. Naval Academy, School of Engineering 6) Incredible sleeping room rates at the headquarters hotel, Washington Marriott Wardman Park, will sell out fast!

STRUCTURE magazine

Applying Expertise as an Engineering Expert Witness – SAVE the DATE! May 19-20, 2016; Chicago, IL Engineers are often asked to serve as expert witnesses in legal proceedings – but only the prepared and prudent engineer should take on these potentially lucrative assignments. If asked, would you be ready to say yes? Developed exclusively for engineers, architects, and surveyors, this unique course will show you how to prepare for and successfully provide expert testimony for discovery, depositions, the witness stand, and related legal proceedings.

Applying Expertise as an Engineering Expert Witness is a focused and engaging 1½ day course that will run you through each step of the qualifications, ramifications, and expectations of serving as an expert witness.

For more information about the course and/or to register, www.acec.org/calendar/calendar-seminar/applyingexpertise-as-an-engineering-expert-witness-chicago-il.

Donate to the CASE Scholarship Fund!

The CASE scholarship, administered by the ACEC College of Fellows, is awarded every year to a deserving student seeking a Bachelor’s degree, at minimum, in an ABET-accredited engineering program. Since 2009, the CASE Scholarship program has given over $17,000 to engineering students to help pave their way to a bright future in structural engineering.

Your contribution today will help CASE and ACEC increase scholarship funds to promising students who need them most.

This is an exceptional opportunity to encourage growth in the structural engineering profession and ensure that the highest caliber of students become the future of our industry.

Your monetary support is vital in helping CASE and ACEC increase scholarships to those students who are the future of our industry. All donations toward the program may be eligible for tax deduction and you don’t have to be an ACEC member to donate! Contact Heather Talbert at htalbert@acec.org to donate.

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CASE is a part of the American Council of Engineering Companies

As capital project delivery increasingly move from traditional Design-Bid-Build to Alternative Delivery System (ADS), it’s critical that engineering firms clarify their roles, responsibilities, and overall project influence when there is no longer a direct communication or contractual relationship with project owners.

Otherwise, firms will quickly find themselves out of the decision making loop and at heightened risk for liability issues.

This full-day workshop will show you how to strengthen your firm’s standing in the ADS project process so that you’re better able to: • Protect your hard-earned reputation • Be compensated fairly • Minimize liability and risk • Provide innovative and effective solutions to the project owner • Educate owners, contractors, and even the public on the contribution engineers make to the success of any ADS project If your firm is considering ADS work, this is a must-attend presentation to help you win more profitable, high-quality projects while protecting your firm from unacceptable financial, liability, performance, or reputational risks.

For more information and/or to register, www.acec.org/ calendar/calendar-seminar/alternative-delivery-systems-adsoptimizing-your-firm-to-win-more-work-with-less-ris.

7) Registering early will keep an extra $100 in your bank account! Early-Bird registration fees end on Monday, March 21st!

8) 2016 Engineering Excellence Awards Gala – honoring the Nation’s Top Engineering Achievements!

9) Celebrate the music of Billy Joel with Mike DelGuidice and BIG Shot – featuring current members of Billy Joel’s band – at the Welcome Dinner.

10) The Annual Convention and Legislative Summit is the A/E Industry’s Foremost Policy and Political Lobbying Event, where over 1,000 of the industry’s leaders will meet this Spring!

For more information and/or to register, www.acec.org/ conferences/annual-convention-2016/registration.

CASE in Point

ACEC Business Insights

Structural Forum

opinions on topics of current importance to structural engineers

The Engineering Way of Thinking: Adaptation

By William M. Bulleit, Ph.D., P.E.

I

n this, the fourth and final column of a series (“The Idea,” December 2015; “The Future,” January 2016; “An Analysis,” February 2016), I ask you to consider the engineering way of thinking (EWT) as a relatively formal way of adapting to a constantly changing environment (in the broad sense) by enabling variation and selection as safely as possible under sometimes significant uncertainty. I will emphasize two sources: Engineers and Ivory Towers, by Hardy Cross (1952); and Adapt: Why Success Always Starts with Failure, by Tim Harford (2011). Cross is a well-known engineer (think moment distribution) from the mid-20th century, and Harford is an economist today.

Cross understood the EWT even in 1952: “They [engineers] use any fact or theory of science, whatever and however developed, that contributes to their art.” He also understood that engineering goes beyond science: “Engineers are not, however, primarily scientists. If they must be classified, they may be considered more humanists than scientists. Those who devote their life to engineering are likely to find themselves in contact with almost every phase of human activity.” 21st century engineers need to think more like Cross, recognizing that no matter how specialized our day-today engineering becomes, we use heuristics that – when generalized – can be useful in a wide range of endeavors.

The EWT is broad enough to allow engineers to design prototypical systems with relatively low uncertainty, such as engines; non-prototypical systems with high uncertainty, such as buildings subjected to seismic effects; and even vast systems with extreme uncertainty, such as the economic system of the United States. The way we go about implementing the EWT is to enable variation and selection – e.g., developing new designs, and then choosing the best based on failures, which can range from simply not meeting a particular criterion to complete system collapse.

Harford has something to say about this.

He describes three principles of adapting, which sound like techniques that engineers have used for decades: first, “try new things, expecting that some will fail”; second, “make failures survivable: create safe spaces for failure or move forward in small steps”; and third, “make sure you know when you’ve failed, or you will never learn.”

From a bird’s eye view, these two authors, separated by about 60 years, have both given a fair description of the EWT. Be interested in and learn as much as you can about anything that might improve your day-to-day engineering, but have wide horizons about what you learn, because you never know what you might need. Your practice will present you with problems that require you to enter areas where you have never designed before, and possibly areas where no one has ever designed. The results will be a form of variation.

When structural engineers engage in this kind of variation, they usually become more conservative and try to vary what they have done in the past as little as possible; i.e., “move forward in small steps.” We generally have little trouble knowing when our designs have failed due to the nature of our systems – if it deflects too much, it failed; if it collapses, it failed – however, recognizing failure is not so easy for all systems, particularly social systems. Is the “War on Drugs” a failure? Is the Affordable Care Act (ACA) failing? If we pick on the ACA, we can see that it attempts something new, yet does not expect any failures. There is no safe space for failures, and it certainly is not a small step.

How can we even tell if it fails? Those who developed it will never admit to any failure.

The ACA does not follow any of Harford’s three principles of adapting.

It would have been (at least arguably) more consistent with those principles to allow the states to develop healthcare plans of their own, as was already happening. Then there would have been 50 experiments, leading to a range of variation. The steps of change

would have been smaller, failures would have been smaller, and comparisons among the states would have better shown which approaches failed, thus allowing selection.

Certainly this is not the only alternative, but it would have better followed the EWT for large-scale social systems. Of course, if we were really to follow the EWT, we would use models to choose the paths of variation and other heuristics to help with those decisions. The full EWT has not yet been used for these types of scenarios.

My intention with these four columns is to get more engineers to think more broadly about how their knowledge can and should be used to enhance not only the technological aspects of our world, but also the natural and social aspects. The techniques that engineers use every day can be generalized. Admittedly, some will not work for social systems, but then we will just need to develop more techniques. Who in 1950 would have visualized nonlinear finite element analyses that can be used to examine the behavior of a steel building subjected to a suite of ground motions scaled to whatever magnitude the designer needed?

Where can the EWT go if we all choose to put our minds to it? Furthermore, what could we accomplish if we were to start training all individuals to some extent in engineering, much like we already do in areas like English, mathematics, history, and science? In our technological society, the EWT may actually be the most important way of thinking that there is. How else will we properly adapt to a rapidly changing natural and social environment?▪ William M. Bulleit (wmbullei@mtu.edu) is a professor in the Department of Civil and Environmental Engineering at Michigan Tech in Houghton, Michigan, and the vice chair of the SEI Engineering Philosophy Committee.

Structural Forum is intended to stimulate thoughtful dialogue and debate among structural engineers and other participants in the design and construction process. Any opinions expressed in Structural Forum are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, C 3 Ink, or the STRUCTURE® magazine Editorial Board.

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March 2016