STRUCTURE magazine | July 2015

July 2015 Wind/Seismic

A Joint Publication of NCSEA | CASE | SEI

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July 2015 edItorIal

7 Hold Your Breath for Just a Moment…

HIStorIC StruCtureS

44 the Warren truss

By Frank Griggs, Jr., D. Eng., P.E.

By David W. Mykins, P.E. rISk MaNaGeMeNt INFoCuS

9 Complicated + Complex = Wicked

58 Myth of Probabilistic Seismic Hazard analysis

By Praveen K. Malhotra, Ph.D., P.E.

By Jon A. Schmidt, P.E., SECB INSIGHtS StruCtural PerForMaNCe

11 Seismic Isolation – the Gold Standard of Seismic Protection

61 Ground Improvement for Building Support

By Damian R. Siebert, P.E. and Steven R. Kraemer, P.E.

By Mason Walters, S.E. ProFeSSIoNal ISSueS StruCtural deSIGN

16 determining the earthquake Shaking Force

By Roumen V. Mladjov, S.E., P.E.

62 two Generations: Why Young Member Groups (YMG) are Important

ProduCt WatCH

64 Fluid Viscous damping for Seismic energy Protection

By Shubin Ruan, Ph.D., P.E. and

Showalter, P.E., Michelle Kam-

Ben Eder

Biron, P.E., S.E., SECB and

51

By Charles Besjak, S.E., P.E.

Feature

StruCtural ForuM

By Andrew D. Sen, Charles W.

74 does a Name Have Value?

Roeder, Ph.D., Dawn E. Lehman,

By Barry Arnold, S.E., SECB

Ph.D. and Jeffrey W. Berman, Ph.D. eNGINeer’S NoteBook

37 transfer of Moments in Slab-Column Connections

By Jerod G. Johnson, Ph.D., S.E.

By Grace S. Kang, S.E. Since its 1972 inauguration, the 20-foot by 20-foot reinforced concrete Shaking Table, located at the Richmond Field Station of UC Berkeley and managed by the Pacific Earthquake Engineering Research (PEER) Center, has been very active and continues to introduce and employ innovative technology, and carry out cutting edge research related to the seismic behavior of structures and equipment.

SPotlIGHt

67 Innovation as Homage 27 How Big is that Beam? revisited

By Mehdi Rashti, S.E. The new 131,000 square-foot Center for Naval Aviation Technical Training Complex (CNATT) consists of three adjoining structures, with very diverse functions. The complex will house all aviation mechanics involved with the maintenance and repair of Huey and Cobra helicopters at the Marine Corps base at Camp Pendleton.

Due to the diverse programming needs of each activity and structural system incompatibility, the project team decided it was necessary to design three buildings so there could be seismic separations between them.

Pioneering Shaking table Continues to be Innovative

By Sofia Zamora, E.I.T. and

By Philip Line, P.E., John “Buddy”

StruCtural ForeNSICS

Center for Naval aviation technical training Complex

Feature

CodeS aNd StaNdardS

Jason Smart, P.E.

Feature

40

Michael Fillion, P.E., SECB

21 2015 Special design Provisions for Wind and Seismic

30

IN eVerY ISSue 8 Advertiser Index 55 Resource Guide (Concrete) 68 NCSEA News 70 SEI Structural Columns 72 CASE in Point

On the cover Numerous types of structures are now protected by seismic isolation in high seismic regions of the world, such as structures with unusual architecture. Read more in the Structural Performance article on page 11. Courtesy of Bruce Damonte.

STRUCTURE magazine

5

Seismic Construction Strong

By Larry Kahaner More than 8,600 people died in the earthquakes that hit Nepal in April and May. World leaders and the design and construction industries are once again focused on the devastating loss of life and property accompanying seismic events. Vendors continue to bring forward new technologies and innovations, as designers incorporate seismic resiliency into structures around the globe.

Publication of any article, image, or advertisement in STRUCTURE® magazine does not constitute endorsement by NCSEA, CASE, SEI, C 3 Ink, or the Editorial Board. Authors, contributors, and advertisers retain sole responsibility for the content of their submissions.

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Editorial

Hold new trends,Your new techniques Breath and current for industry Just issues a Moment…

By David W. Mykins, P.E., CASE Chair

I

remember when I was a kid, I remember riding in a car with my family and my grandma. As were nearing a local cemetery, she quietly said to me “You know, you need to hold your breath when we pass the cemetery. There are restless ghosts there that want to haunt you, and the only way to keep them away is to hold your breath”. That was pretty scary stuff. Needless to say, I did as I was told. I often feel the same need to hold my breath when I hear about a structural failure. What’s your response when you learn problems caused by the unfortunate errors of others?

Some seek the opportunity to capitalize on the incident and boost their own businesses at the expense of those involved. The Germans have a word for this; it’s called “schadenfreude” and it means taking joy from the misfortune of others. While this may seem to be good for your business in the short term, it may do long term and even permanent damage to your reputation if you are perceived to be taking unfair advantage of a dreadful situation.

A few will seek to help the parties involved by offering their services as forensic experts to determine the causes of the failure. This may mean establishing whether the standard of care was met by the design professional, whether there were standard construction safety measures that were not observed, or if there were some unforeseen forces or conditions contributing to the failure. The role of forensic expert may be best served by firms who are geographically removed from both the project site and from the offices of any involved parties. The physical separation allows forensic consultants to be objective and avoids the appearance of trying to take advantage of a competitor.

Most firms, however, will never be directly involved in the investigation of the failure. Most firms will simply be outside, interested – but hopefully not passive – observers. By this I mean: even if not directly involved, practicing structural engineers should take an interest in these types of stories and use the lessons learned from them to improve our own practices, and ultimately, the profession.

When an incident occurs, we shouldn’t simply read about it and think “better them than us” and move on. We should take the opportunity to turn inward and look at our own training, policies, procedures and quality control. Consider what steps can be implemented in our own practices to avoid a similar incident. Examine whether we have the right people in the right roles in our firm. Do we have adequate technical reviews of projects? This is especially true for projects that stretch our firm’s capabilities. Take a look at what types of tools, documents, references and resources your firm has, and whether they are still relevant to the way you are providing services. Chances are that you will find some holes and will STRUCTURAL need to make some changes or ENGINEERING additions to your practice policies INSTITUTE or procedures.

“You know, you need to hold your breath when we pass the cemetery.”

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As an example, one of the changes our firm made recently as a result of this type of self-examination was the decision to implement a program of internal peer reviews. Here’s how our process unfolded. We were considering ways in which we could improve the quality of our documents in light of a recent structural failure we had read about. We began to consider whether a peer review might have brought to light the problem that ultimately caused the failure. We had been peer reviewed by other firms in the past, and have also been peer reviewers. It had been our experience that, no matter how good a firm’s quality control procedures are, it is almost certain that the peer review process will bring up questions that improve the design. However, since peer reviews are not required in the jurisdictions where we typically practice, these experiences have been rare. So we decided to begin performing random peer reviews on ourselves.

This is just one example that we hope will work for us. If you are looking for some other ideas or helpful tools to organize your firm’s approach to managing the risks associated with structural engineering, you will find lots useful information on the CASE website (www.acec.org/CASE). CASE provides contracts, guidelines and practical administrative tools to help structural engineers in both business management and project management.

And next time you hear about an incident involving a structural failure, hold your breath for just a moment. Critically examine your own business practices and try to make at least one change to help improve the way that you provide services. Your grandmother will be relieved that you did.▪

STRUCTURE magazine

David W. Mykins, P.E., is the president and CEO of Stroud, Pence & Associates, a regional structural engineering firm headquartered in Virginia Beach, VA. He is the current chair of the CASE Executive Committee. He can be reached at dmykins@stroudpence.com.

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July 2015

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July 2015

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InFocus

Complicated new trends, new techniques + and Complex current industry= issues Wicked

By Jon A. Schmidt, P.E., SECB

H

uman language is inherently ambiguous. Most words have multiple meanings, which are subject to change from time to time and from place to place. Even the correct pronunciation of the same arrangement of letters can vary, and the only way to tell which is correct is by taking the context into account. “Did you read my column last time?” “Yes, I read it.”

Although spelled differently, the terms “complicated” and “complex” are often used as close synonyms in ordinary speaking and writing, so any distinction made between them is likely to be a highly technical one. Swedish authors Claes Andersson, Anton Törnberg, and Petter Törnberg attempt to thread this needle in a recent paper, “Societal Systems – Complex or Worse?” It appeared in the November 2014 issue of Futures (Vol. 63, pp. 145-157) and is available online at www.insiteproject.org/wp-content/uploads/2013/05/RP_2013.pdf.

Andersson et al. begin by acknowledging the close relationship between complicatedness and complexity. In fact, in order to differentiate them at all, we essentially have to define complexity as “what we intuitively think of as complexity, but minus complicatedness.”

One way to do this is advocated by Péter Érdi in his 2008 book, Complexity Explained: identify complicatedness as structural complexity, and complexity per se as dynamical complexity. Another is to associate complicatedness “with top-down organization, such as in engineering,” and complexity “with bottom-up self-organization – like the behavior of a school of fish or a crowd.”

Complicatedness is commonly addressed by means of various systems-based theories that account not only for the behavior of individual elements, but also the relations between them. Finite element analysis is an example familiar to structural engineers. The field of “complexity science” has emerged much more recently. While it is “highly multidisciplinary,” involving the collaboration of a wide variety of specialists, Andersson et al. point out that “it is not as methodologically diversified,” generally favoring formal and quantitative approaches, especially computer simulation grounded in non-linear dynamical systems theory. Its effectiveness is thus limited to “a specific class of systems that happens to be amenable to analysis using that particular toolbox.”

Difficulties arise when the proper domains of systems-based theories and complexity science are not carefully observed. In particular, there is “no reason why systems could not be both complicated and complex at the same time.” Andersson et al. refer to such systems as “wicked,” a term adapted from management science, where it was coined in the late 1960s for “a class of problems that failed to fit into the molds of the formal systems theoretical models that were being applied across the board at the time with considerable confidence.” As a result, wicked problems – such as “starvation, climate change, geopolitical conflicts, social disenfranchisement, and so on” – generally cannot be usefully defined apart from the proposal of a specific solution, which will often be only partial at best.

Wicked systems are similar, in that neither a systems-based theory nor complexity science is adequate for representing them – alone or even in combination. In particular, Andersson et al. assert that STRUCTURE magazine

several researchers who have attempted to apply complexity science to wicked systems have been unsuccessful because they failed to recognize that wickedness is not just a different type or higher level of complexity, but has the additional dimension of complicatedness. What makes the interaction of these two properties so intractable is how they “fuse into something quite unlike either quality in isolation.” In other words, wickedness is an emergent phenomenon: “the rules and entities are not only hard to uncover, they change as a result of the dynamics itself.”

Andersson et al. suggest that this renders wicked systems resistant to “just about any conceivable formal theorizing.” Utilizing Herbert Simon’s terminology, they note that such formalization requires three key idealizations: • “an internal environment where the dynamics that we study takes place”; • “an external environment that can be assumed to be static, or at least to be variable only in highly regular ways”; and • “The boundary between the internal and external environment … referred to as the interface.”

The resulting model “makes the world manageable” because “we declare our system as autonomous from external disturbance and we hide any complexity and complicatedness residing on lower levels.” We are then able to “study this internal environment during … the short run: a time scale that (i) is long enough ... for important dynamics to have time to happen and (ii) short enough that our assumptions about the interfaces remain valid.”

As Andersson et al. point out (citing Simon), “Engineered systems … are designed to fit into [the] above description … The parts of such a system can be improved independently, with respect to identifiable functions, as long as those functions in the system are retained … In fact, you can do anything to a component as long as you do not alter its interface.” This is why engineered systems are often extremely complicated, but not necessarily highly complex, and therefore not wicked; they are intentionally devised and constructed that way.

The two principal examples of wicked systems are societies and ecosystems. Interestingly, these directly correspond to two high-profile concepts among engineers today: resilience and sustainability. We typically are in a position to address such considerations only one project at a time, since that is the limit of what we can design to fit into the above description. However, it is clear that their actual scope is vastly greater. Is it sufficient for engineers to continue playing such a small but important part in human attempts to preserve these wicked systems? Or is there a larger role that we can and should embrace?▪ Jon A. Schmidt, P.E., SECB (chair@STRUCTUREmag.org), is an associate structural engineer at Burns & McDonnell in Kansas City, Missouri. He chairs the STRUCTURE magazine Editorial Board and the SEI Engineering Philosophy Committee, and shares occasional thoughts at twitter.com/JonAlanSchmidt.

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July 2015

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I

magine yourself inside a house that rests on a frozen, frictionless lake when a violent earthquake occurs. Apart from noticing some up-and-down vibration, how would you know the ground is shaking? The lack of a horizontal “connection” to the earth would allow the ice to shift horizontally without affecting the house. Neither you nor the house – nor even the cup of coffee on the table – would experience the earth’s horizontal shaking. This idealized scenario exemplifies the purest concept of seismic isolation, also known as base isolation – arguably the current “gold standard” of seismic protection.

Of course, the reality of seismic isolation is not quite so ideal, if only because most building sites have legal property boundaries. Even after a major earthquake they must occupy the same piece of real estate they started on. Such earthly considerations require: 1) limitations on how far a seismicallyisolated structure can be allowed to move during a design-level seismic event, and 2) a mechanism to restore the structure to its original footprint.

In essence, seismic isolation allows a structure to safely “dance” with the earth, rather than fighting it. In doing so, seismic isolation represents a departure from the widely-accepted conventional seismic design approach that essentially requires a structure to damage itself (potentially severely) to protect its occupants during a major earthquake. By definition, isolating a building from seismic shaking is the most effective way to protect not only a building, but its occupants, contents, and its function. The introduction of seismic isolation to earthquake engineering in the mid-1980s consequently resonated with the structural engineering profession’s growing interest in performance-based seismic design, which started to develop in the early 1990s.

In the thirty years since seismic isolation was first applied to a building in the U.S., extensive testing and research have been accomplished on the topic and numerous earthquakes have provided evidence of its effectiveness. However, adoption of seismic isolation has been relatively slow in the U.S. An excellent article published in STRUCTURE in March 2012 by Taylor and Aiken proposed likely reasons that U.S. adoption of the technology had not kept pace with that of other earthquake-prone countries. Primary among those reasons was added cost and design complexity. This article is intended to provide a brief introduction to seismic isolation and a status update on current applications of the technology, followed by descriptions of new developments that could herald wider application in the U.S.

Roots in the Past The idea of modern seismic isolation has its roots in vibration isolation, wherein resilient bearings are

used to protect buildings from disturbances due to vibrating machinery that predictably operate in pre-identified frequency ranges. In the case of seismic isolation, however, the vibrating “machinery” is the most unpredictable vibration source in existence – the earth itself. To date, nobody can accurately predict the most important data for designing against an earthquake: when it will occur, how it will shake, and how long it will last.

This lack of definitive design criteria precludes a precisely-targeted design of any seismic protection system – including seismic isolation. Rationally designing a seismic isolation system is therefore considerably more challenging than proportioning vibration isolators for motorized equipment.

Nonetheless, seismologists, working with geotechnical engineers and geologists, are capable of determining probable ranges of ground motion amplitude and predominant frequencies of seismic ground shaking for a given site. This information can be used by structural engineers as a rational basis to evaluate and design a robust seismic isolation system for a building or other structure. Evolving computational techniques afford the engineer practical ways to utilize such input to rationally project the most likely modes of dynamic behavior of an isolated building. The fact that the International Building Code (IBC) and other relevant codes require extensive project-specific “prototype” testing for each specific type of seismic isolation bearing to be used on a project bestows a level of behavioral reliability to seismic isolation not necessarily enjoyed by other systems intended to protect structures from earthquakes.

Since seismic isolation was first applied in the U.S. in the 1980s, the relatively small market for isolation bearings has not driven a proliferation of isolator types; in fact, the readily available domestic supply of isolators has consolidated into two distinct but competitive types, with much less significant participation from nondomestic manufacturers. However, these isolator types have been substantially improved since they were developed, and their engineering properties are now much better understood and applied in practice. The sizes, displacement capabilities, and load capacities of isolation bearings have increased dramatically as well.

Structural Performance performance issues relative to extreme events

Seismic Isolation – The Gold Standard of Seismic Protection Where is it Headed?

By Mason Walters, S.E.

Mason Walters, S.E., is a Senior Principal and the Technical Director at Forell/Elsesser Engineers. He is a member of the Structural Engineers Association of California (SEAOC), the Solar Seismic Committee, and the Structural Engineers Association of Northern California (SEAONC) Subcommittee on Protective Systems and the Subcommittee on Non-Building Structures. Mason may be reached at m.walters@forell.com.

How it works The currently applicable concept of seismic isolation works by replacing the direct, rigid connection of a structure to the earth beneath it with a set of horizontally flexible bearings that can allow the structure to remain relatively undisturbed even as the earth moves violently. A seismic

STRUCTURE magazine

11

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

Figure 1.

isolation system may be located at the base of a building or at a story within the superstructure (Figure 1). The resulting horizontal displacement between the structure and the earth effectively lengthens the fundamental period of horizontal seismic vibration, which can significantly reduce the horizontal base shear that a structure would experience. In addition, seismic energy may be dissipated through the provision of passive damping in order to provide some control over the amount of relative displacement that may occur (Figure 2).

Tools of the trade Currently available categories of seismic isolation bearings include multi-layer elastomeric devices, friction devices, and ball-bearingbased devices. Elastomeric seismic isolation bearings were originally modeled after elastomeric bridge bearings and consist of a vulcanized multi-layer “sandwich” of elastomer layers and thin steel plates. The two

most prevalent types in this category are the lead-rubber bearing (LRB), which contains a cylindrical shear-yielding lead core for energy dissipation, and the high-damping rubber bearing (HDR), which dissipates energy by amending natural rubber to contain filler materials such as carbon black. In locations where linear-elasticity suffices without the inclusion of damping, the nearly linear-elastic natural rubber bearing (NRB) may be used.

The most commonly used and well-known friction device is the friction pendulum (FP) isolator, which utilizes an articulated slider that moves horizontally on a spherical dishshaped surface. The spherical shape of the sliding surface determines the translational period of the isolation system, and forces it into slight vertical movement that creates a restoring force. The most recently developed “Triple Pendulum” version of this type of device, patented and manufactured by Earthquake Protection Systems, Inc., contains a compound articulated slider with multiple sliding surfaces to allow control of the sliding sequence and the resulting hysteresis curve.

Refer to Figure 3 for representative diagrams of the various isolator types named above.

Current Applications Although it is challenging to obtain accurate numbers on seismic isolation work, a recent estimate by Dynamic Isolation Systems put the approximate total of completed isolation projects at over 10,000 worldwide. This total, which is likely conservative, is heavily weighted towards Asia, particularly Japan, where vibrant “reminders” of potentially damaging seismicity occur frequently in densely populated areas. Table 1 provides a breakdown by nation. This points out a trend recognized among structural engineers who specialize in seismic isolation design: demand

Figure 3.

Figure 2.

STRUCTURE magazine

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July 2015

Table 1. Approximate numbers of seismically isolated projects worldwide – April, 2015.

Country US Japan Canada Chile Colombia Mexico New Zealand Peru Taiwan Turkey China Other Total

Total Projects (approx.)

500 7,800* 50 75 20 25 50 10 50 40 4,000 100 12,720

(List compiled by Dynamic Isolation Systems) *Japan total includes 3,300 major buildings and 4,500 single family homes.

for the technology jumps in the years immediately following a damaging earthquake.

This tendency was first noted following the 1989 Loma Prieta earthquake, and has been repeated after every subsequent major damaging event.

Numerous types of structures are now protected by seismic isolation on highly seismic regions of the world. Examples include buildings, bridges, viaducts, pipelines, offshore platforms, telecommunications facilities, water and fuel storage tanks, power transformers, computer floors – and the list goes on. Common isolated building types include mission-critical facilities such as hospitals and emergency response communication centers, historic and/or landmark public buildings, owner-occupied office buildings, and high content-value buildings such as museums.

Other less common examples include highend apartment buildings and high-value manufacturing facilities. Still more specialized

Figure 4. Courtesy of Bruce Damonte.

examples include a botanical storage facility containing rare/precious plant specimens, an isolated rooftop office building addition, structures with unusual architecture (Figure 4) and isolated floors (Figure 5) within buildings to support high-value contents.

Not surprisingly, seismic isolation currently retains a long-held reputation as a high firstcost approach for “high end” buildings. If one ignores its potential long-term benefit, seismic isolation can seem financially out of reach for less-critical projects.

This impression is compounded by exhaustive code-mandated evaluation, design, and review requirements that apply only to seismic isolation and other advanced seismic protective systems like passive damping.

Ironically, the original developers of modern seismic isolation conceived the approach as a way to simplify seismic design, as well as to enhance predictability and performance.

Special requirements in the current IBC and the American Society of Civil Engineers’ (ASCE) Seismic Evaluation and Retrofit of Existing Buildings (ASCE 41) that apply to seismic isolation but not conventional seismic systems include: • Site-specific ground motion studies • Design for two levels of earthquake input • Peer review with multiple reviewers These requirements, all of which are expensive and time-consuming, may be prudent to use for complex projects, but are generally recognized as unnecessary for simpler buildings.

In discussing cost-benefit comparison considerations between isolated and non-isolated buildings, R.L. Mayes, of Simpson Gumpertz & Heger, concludes that, considering the cost of earthquake insurance premiums, using base isolation without earthquake insurance can be a more cost-effective solution than a conventional fixed based structure with insurance when total cost of ownership is

considered, despite the first cost premium for base isolation.

Positive Developments There have been a handful of technical developments in recent years that can be expected to encourage more use of seismic isolation going forward. These developments include: 1) Recent acceptance of code revisions to simplify implementation of seismic isolation: The seismic isolation code provisions have remained largely unchanged since they were introduced as an appendix chapter to the Uniform Building Code (UBC) in 1991. In late 2014, ASCE Technical Committee 12 on Seismic Isolation & Damping under the Seismic Subcommittee of the SEI Codes & Standards Committee for ASCE/SEI 7 succeeded in obtaining approval for

Figure 5.

STRUCTURE magazine

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July 2015

a number of major revisions for the 2016 edition of ASCE’s Minimum Design Loads for Buildings and Other Structures (ASCE 7) Chapter 17 requirements for the design, analysis, testing, and peer review of seismic isolation applications. These simplifications could significantly expedite the implementation of seismic isolation, as well as make it more economical.

2) New research and development: Late in 2014, researchers at Stanford University concluded a two-year long, $1.3 million NSF-funded research program to study two new types of seismic isolation bearings conceived for inexpensive application in light-frame residential construction. The validation testing for the project, called Seismically Isolated Unibody Residential Buildings for Enhanced Life-Cycle Performance, was conducted at the University of California, San Diego’s outdoor seismic shake table facility (Figures 6 and 7 , page 14 ).The tests shook a full-scale wood frame structure, first with re-centering dish-shaped sliding isolators, then again with flat sliding isolators, both comprised of high-strength plastic on galvanized sheet steel. Stanford representatives estimate that implementing such a system might add roughly $15,000 to a house that would otherwise cost $400,000. In the author’s opinion, retrofitting an existing home with seismic isolation should be expected to be more costly. continued on next page

Figure 6.

Figure 7.

3) Readily available design examples: The Structural Engineers Association of California (SEAOC) published a detailed design example for seismic isolation developed in accordance with the 2012 IBC. The example is contained in Volume 5 of the five-volume compilation comprising the 2012 IBC Structural/Seismic Design Manual, (SSDM) published in 2013. SEAOC plans to update the SSDM as the IBC is revised. The peer-reviewed example covers the use of both elastomeric and friction pendulum isolators.

4) Lower isolator prices: As the use of seismic isolation has gradually become more widespread, production of seismic isolation bearings has become more efficient, leading to gradually falling isolator prices – especially after accounting for inflation. Interestingly, the production leading to this efficiency is partly due to the proliferation in exported bearings for projects outside the U.S. A related improvement is that of production quality. One reason for this is that Japan, the largest single consumer of seismic isolation bearings by country, established a set of rigorous standards for isolator performance following the 1995 Kobe Earthquake. Japanese domestic and non-Japanese manufacturers had to improve their quality to match those standards in order to compete for Japanese business.

U.S. and non-Japanese isolation customers alike have consequently experienced a spin-off benefit from the improvement. Resulting increases in, for example, elastomeric isolator

shear strain capacity have allowed the use of smaller isolators for a given displacement demand, which have also helped reduce production costs.

5) Faster isolator procurement cycles: As more various sizes of isolators are produced, the number of different readily available isolator sizes has increased, reducing the need to manufacture new molds for both elastomeric bearings and ductile iron friction pendulum components.

Also, isolator manufacturers have honed their testing processes, and the pre-fabrication process is now more efficient due to a growing body of empirical testing of various isolator designs, and a streamlined testing process.

6) More publicity: Various seismically isolated structures have experienced actual earthquakes. Notably, the Japanese Red Cross Hospital in Ishinomiki City in Miyagi Prefecture, located approximately 75 miles from the epicenter of the M9.0 Tohoku Earthquake of March, 2011, was open for business immediately following the event thanks to its seismic isolation system and to the function of its emergency generators.

Accounts of this experience, including videos taken during and immediately after the earthquake, were described by various parties.

One example is an Oregon Public Broadcasting story by Ed Jahn.

7) Development of advanced “loss estimation” techniques: A 9-year long FEMA-funded research project by the Applied Technology Council (ATC) resulted in a probability-based seismic loss estimation methodology

STRUCTURE magazine

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July 2015

for buildings described in a twovolume FEMA document P-58-1, Seismic Performance Assessment of Buildings, complete with free software and introduced to the profession in September 2012. FEMA P-58-1 is tailored to equip structural engineers to rationally assess and communicate seismic risk using language and concepts common to the real estate and insurance industries. The FEMA P-58-1 procedure accounts for variation in demand parameters such as acceleration and interstory drift between different seismic structural systems in assessing potential earthquake-related losses. The lower structural demands associated with seismic isolation equate to lower risk of financial losses, injury, and downtime.

8) The Green Building movement: The Green Building movement may have largely overlooked the significant greenhouse gas contribution (or savings) of structural materials, but the advantage of providing higher seismic resilience as a sustainability measure has become obvious to most structural engineers in highly seismic areas. Seismic isolation – arguably the pinnacle of seismic resilience approaches for many types of building structures – is consequently gaining recognition for its potential role in seismic sustainability.

It may be that any of the above developments do not cause an immediate boom in demand for seismic isolation. However, taken together, they should ultimately help structural engineers move the needle toward seismic isolation’s original developers’ goal of simplifying the provision of highly reliable seismic performance, as well as making it more economically attractive.▪

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Structural DeSign design issues for structural engineers

A

new method has been developed for measuring the force of ground shaking during earthquakes at particular locations. The proposed Earthquake Shaking Force (EqSF ) rating is based on the maximum vector sum of the recorded ground accelerations in the three main directions scaled with the strong ground motion duration. An equation for calculating the Earthquake Force values is proposed. This new method has been used to analyze and compare more than 220 ground station recordings from 48 earthquakes in the United States and around the world. The results show that the new method can provide objective ranking of the ground shaking forces, and can help engineers in designing seismically-resistant structures.

Today geologists and engineers use Magnitude scales and the Modified Mercalli Intensity scale to measure and rank earthquakes. For a century the Modified Mercalli Scale has been used for evaluating the intensity of local ground shaking.

Different versions of this scale are still used, with its twelve-point range from I to XII, around the world. The Modified Mercalli Scale is based on the feelings and reactions of individuals and on observed damage to structures and underground facilities. This makes the scale subjective and inaccurate, because of the different individual interpretations (due to different sensitivity and reactions), the specific construction conditions of the country, the year of construction, or the level of building development.

Some improvement was achieved in California by modifying the Mercalli Intensity scale again, using Instrumental Intensity, in an attempt to correlate the intensity scale values with the peak ground accelerations and velocities. Instrumental Intensity is used by the TriNet system to produce ground shaking maps showing the peak ground accelerations and velocities by ground stations for a specific earthquake.

Today, California and many other states and countries have developed large nets of ground motion recording stations. During an earthquake, all necessary ground motion data are recorded and the Mercalli Intensity Scale is no longer providing reliable information. The new Earthquake Shaking Force method is a step forward in providing objective measurement of the ground shaking based on recorded ground motions during an earthquake.

Determining the Earthquake Shaking Force For Structural and Bridge Engineers

By Roumen V. Mladjov, S.E., P.E.

Roumen V. Mladjov, P.E., S.E., has more than 52 years in structural and bridge engineering, and in construction management.

He lives in San Francisco and his main interests are structural performance, efficiency and economy. He can be reached at rmladjov@gmail.com.

Measuring the Force of Ground Shaking The online version of this article contains additional data and detailed references. Please visit www.STRUCTUREmag.org.

This new EqSF approach is created for measuring and comparing the force of ground motions at local sites at a particular distance from an earthquake epicenter.

16 July 2015

Figure 1. Earthquake Shaking Force, 3D vector sum of ground accelerations.

The Earthquake Shaking Force is based on the well known Newton’s Second Law of Motion F = m* a, where the inertial force F, equals the mass m, multiplied by the acceleration a. Engineers are using the same basic principle to determine the lateral seismic forces for designing buildings and structures. The shaking force values are calculated using the instrumental readings for ground accelerations and the duration of strong ground motions.

Similar to the Richter Scale, the proposed method does not have a fixed top limit and the calculated values are rounded to the first decimal digit. Based on the highest recorded peak accelerations, we should expect a ground motion rarely to exceed EqSF level 13–16. As a reference for a recorded ground shaking with total acceleration (the space vector sum) equal to 1.0g and strong motion duration of 20 seconds the calculated EqSF value is 9.8.

The physical meaning of the proposed scale value is a force equal to the maximum vector sum of the ground accelerations in x, y and z directions (within an interval equal or less than 1.5 sec) in m/ sec/sec multiplied by one unit mass (m=1) (Figure 1). This theoretic value is corrected for vertical acceleration and duration. The whole calculated value is scaled in order to receive a range closer to the values used in different intensity scales.

The formula for calculating the Earthquake force of the local ground motion is: EqSF = 9.81{[Cahx^2+ Cahy^2 + (Cav/2)^2] (t/20)^2}^0.2 Equation 1 where, Cahx, Cahy and Cav are the corresponding horizontal and vertical accelerations (g) in x, y and z directions that provide the maximum vector sum, Cav/2 is the correction accounting for a relatively reduced impact of the vertical acceleration, t is the duration of strong ground motions in seconds, t/20 is a correction for the strong motion duration, ^0.2 is the scaling correction for the 3D vector value (in lieu of ^0.5), 9.81 is the acceleration of gravity in m/sec/sec m is the mass, taken as 1, which therefore does not appear in the equation.

All accelerations are taken from the three seismograms (two horizontals and one vertical) recorded at a ground station for a particular event. The time interval considering simultaneous action of the corresponding accelerations in three perpendicular directions is taken as 1.5 seconds. The duration of strong motion is the modified “bracketed duration” for the time interval between the first and last acceleration peaks greater than 0.1g. For the strong motion duration an upper limit of 75 seconds is used. An exception is made for seismograms with acceleration peaks smaller than 0.1 g that should result in t = 0 and EqSF = 0. The EqSF value for such earthquake readings is calculated with duration (t) equal to 0.5 seconds in order to allow comparing their ground shaking force with the forces at other locations. The original “bracketed duration”, proposed

by Page et al. and Bolt accounts for the intervals between 0.05 g peaks.

When accelerations are given in cm/sec/sec, the EqSF values may be calculated directly from: EqSF = 0.6237{[Cahx^2+ Cahy^2 + (Cav/2)^2](t/20)^2}^0.2 Equation 2

More than 40 earthquakes have been included in this study. The criterion for selecting these earthquakes was based on available information (seismograms), and on magnitudes near or more than 6.0. The main source of information was the website of the Consortium of Organizations for StrongMotion Observation Systems (COSMOS), listing recorded earthquake data from more than 500 earthquakes and 6,600 stations (by January 2015). From the selected strongest events, more than 150 recorded seismograms were analyzed. The summary of this analysis is presented in a Table online including data from El Centro, California in 1940 to South Napa, California in 2014. This table lists the magnitude, the focal depth, number of fatalities, number of stations with recorded data and the maximum calculated EqSF scale at a ground station for the event. One more piece of data is included – the average from the top five calculated EqSF (Average 5 EqSF). When only one station recording is available, the Average 5 EqSF is calculated as 80% of the single station result. The combination of Max EqSF and the Average 5 EqSF provides valuable comparative information for the shaking force of earthquakes. The use of Average 5 EqSF should be combined with engineering judgment. This information is more credible for earthquakes in California, Japan, or Taiwan, where the instrumentation net is well developed and recordings from multiple stations are usually available.

In Table 1 (page 18) are listed the recorded accelerations, strong motion durations and calculated EqSF values for representative stations for the earthquakes included in this study (extended table online). These data are structured additionally in a summary Table 2 (page 19) listing the stations by the calculated EqSF values and provided maximum horizontal, vertical and vector accelerations.

The maximum registered acceleration in one direction is at Tsukidate Station, Japan, on March 2011 – 2,700 cm/sec/sec (2.75g); during the same earthquake at this station are calculated also the maximum horizontal (vector sum) acceleration – 2,983 cm/sec/ sec (3.04g), and the maximum total (3-D) acceleration – 3,526 cm/sec/sec (3.59g).

The maximum vertical registered acceleration is at Tsukidate Station, Japan, on March 2011 – 1,880 cm/sec/sec (1.92g).

In Table 2 there are 21 stations listed with horizontal accelerations exceeding 1.0g, and 27 stations with total (3D) acceleration exceeding 1.0g.

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The longest strong ground motion duration, 68 sec., is recorded at Talca Station, Chile, in 2010, whereas the duration recorded at Hiroo Station, Japan, in 2003 is 65 sec., and the duration at Hokoto Station, Japan, in 2011, is 58 sec. The longest duration in California is at Maricopa Station, during the San Fernando earthquake in 1971 (31.5 sec.).

The highest EqSF value is calculated for the Tsukidate Station, Japan, in 2011 (23.4). The Tarzana Station has the highest EqSF value calculated in the US (13.5).

The first three places for the Average 5 EqSF are for Valparaiso, Chile (12.8), ChiChi, Taiwan (12.3), and Western Tottori, Japan (9.8). The highest Average 5 EqSF in California is for Northridge (9.4).

Frequency Influence on EqSF Determination Part of this study was an attempt to include in the EqSF scale the influence of the earthquake frequency (Hz or cycles in seconds) in addition to the acceleration and strong motion duration. Two optional criteria were studied – average frequency measured for “10-second bracket” and measured for the entire “strong motion duration”.

It was concluded that it is difficult to avoid some subjectivity in measuring the frequency.

The criteria were based on full cycles (cycles with vibrations registered on both sides of the neutral axis), but objective criteria could not be established. The studied two bracket durations give different results: the variation increases with the strong motion duration.

The tolerance in reliability of measurement of the frequency is estimated as 15 to 20%. There is almost no difference in frequency between recorded seismograms in two perpendicular directions. For several records (with peak ground acceleration varying between 226 and 1,019 cm/sec/sec or 450%), the impact

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The correction for vertical acceleration is introduced to account for its relatively smaller influence on structural damage. Reduction factors from 1 to 3 produce small differential results less than 7 percent; therefore the final selected reduction factor is 2.

The t/20 is introduced to account for the influence of ground motion duration on structural damage. Twenty (20) seconds were selected as basic duration based on the analysis of more than 100 earthquake records. The scaling correction (^0.2 in lieu of ^0.5) is to reduce the sharp increase in calculated degrees between smaller and greater acceleration values, and to scale the values closer to the traditionally used values in earthquake scales.

The above modifications were selected based on calculated EqSF values for some sites with available ground motion data from the Northridge, Loma Prieta, San Fernando, and Kobe earthquakes. The data used are from the Strong Motion Data Center on the Internet – homepage of the Department of Conservation, Division of Mines & Geology. Some data for the sites not listed on the web are based on presentations and publications following the Northridge Earthquake by Kircher (1994) and Somerville (1997). The data used was updated later based on the registered seismograms made available on the COSMOS website.

Study on Available Data from Earthquakes Recordings

Table 1. Earthquake Shaking Force.

CahX

CahY

CaV

t

cm/sec/sec

cm/sec/sec

cm/sec/sec

sec

EqSF

341.70

185.00

144.00

24.70

1148.00

1055.00

434.00

11.50

Gazli, Uzbekistan, May 17, 1976 M7.0 703.10

518.00

1331.70

582.50

434.90

11.00

Imperial Valley, CA, October 15, 1979 M 6.5 351.40

763.20

229.90

889.40

14.80

13.00

Valparaiso, Chile, March 3, 1985 M 7.8 656.00

707.00

437.00

696.00

849.00

53.80

393.00

54.30

Loma Prieta, CA, October 17, 1989 M 7.1 617.70

469.40

431.10

11.60

449.00

390.80

500.10

16.40

1468.30

1019.40

738.90

19.00

155.80

97.90

15.90

3.00

Cape Mendocino/Petrolia, April 25, 1992 M 7.0 Landers, CA, June 28, 1992 M 7.3 798.00

716.80

383.00

Northridge, CA, January 17, 1994 M 6.7 1744.50 922.70

739.50

912.00

535.30

593.00

1027.50 325.00

843.70

680.81

510.10

679.83

506.22

424.77

865.00

842.50

611.30

793.20

335.30

716.70

927.20

720.40

753.00

607.10

775.80

459.00

32.70

El Salvador, January 13, 2001 M 7.6 1155.00

574.00

342.00

Arequipa, Peru, June 26, 2001 Mw 8.4 304.11

264.87

176.58

East Coast of Honshu, Japan, May 26, 2003 M7.0 888.10

795.60

556.10

409.00

636.50

379.00

Hokkaido, Japan, September 25, 2003 M8.0 972.60

559.50

618.00

345.30

316.00

168.50

Chuetsu, Japan, October 23, 2004 M6.6 1307.91

1715.50

852.80

849.55

Chile, February 27, 2010 M8.8 639.61

686.70

598.41

922.14

663.50

1163.60

Capitola

Treasure Island

Average 5 EQF = 8.4

11.0 8.2

8.8

8.8

12.7

12.9

15.6

11.9

Average 5 EQF = 8.9 Lucerne Valley

Average 5 EQF = 9.4

Sepulveda, VA Hospital

Taichung TCU 129

32.90

12.5

Rikuzentakata, IWTH27

16.8

Hiroo, HKD120

44.00

65.00

51.50

8.2

13.1

12.2

932.00

14.00

1438.10

8.00

55.00

16.4

9.0

23.4

811.18

58.00

19.0

0.00

20.00

9.8

6.00

6.3

75.0

120.0

Tonkamachi, NIG021

19.9

Ojiya, NIG019

Average 5EQF= 13.5 Angol

Average 5EQF= 7.1

Average 5EQF= 7.2

Healthcote Valley Primary School Average 5EQF= 18.5 Tsukidate - MYG004 Hokota

Average 5EQF= 5.1

Carquinez Br. Array 1

PGA=1g with t=20 sec.

Last updated on March 28, 2015 - over 500 earthquakes reviewed with over 6600 station recordings.

STRUCTURE magazine

8.0

Seika, HKD097

Average 5EQF=9.99

Greendale

61.00

11.9

119.0

8.9

284.49

9.6

3.4

Tamayama, IWTH02

Conception San Pedro

9.50

7.3

148.0

Average 5EQF=12.37

18

July 2015

19.0

5.5

Arica Casa

Average 5EQF=11.8

1.1

42.0

12.5

Average 5 EQF = 8.9

15.0

3.8

Hakuta, Station SMNH01

Unidad de Salud, La Libertad

30.00

11.7

31.7

Hino, Station TTRH02

18.0

97.6

Chiayi CHY 080

Average 5 EQF = 9.8

33.0

7.9

1.2

Average EQF = 12.3

12.0

n/a

Takarazuka

JMA Kobe

8.4

n/a

12.3

Average 5 EQF = 8.7

km 6.0

1.8

Sylmar, Converter, Valve Group 7

14.2

313.92

0.00

7.7

2.4

Corralitos

53.50

961.38

981.00

7.3

San Isidro

Average 5 EQF = 6.5

568.98

564.40

South Napa, CA, August 24, 2014, M6.0 510.12

14.8

Llolleo

12.6

1879.93

1070.26

14.0

22.80

1268.49

1354.64

9.0

Average 5 EQF = 12.8

820.17

Tohoku, Japan, March 11, 2011 M9.0 2699.89

Bonds Corner, Hwys 115 & 98

Agrarias

11.6

Christchurch, New Zealand, February 21, 2011 M6.3 1426.70

8.3

7.2

23.00

Canterbury, New Zealand, September 3, 2010 M7.0 737.70

unknown

Average 5 EQF = 7.3

8.9

50.00

Depth

8.5

Karakyr, Uzbekistan

15.00

32.00

Note

8.2

8.1

Tarzana, Cedar Hill Nursery

35.00

Western Tottori, Japan, October 6, 2000 M 7.1

Pacoima Dam

13.5

15.00

Jiji, Taiwan, September 20, 1999 M 7.6

Average 5 EQF = 6.9

21.50

15.20

Kobe, Japan, January 17, 1995, M 6.9

9.5

El Centro ARRAY Station 9

Cape Mendocino/Petrolia

11.60

584.00

7.4

Average 5 EQF = 7.6*0.8 = 6.1 (single station)

12.3

24.00

Distance km

El Centro, CA, May 18, 1940 M 6.9 San Fernando, CA, February 9, 1971 M 6.6

Station

34.1

97.7 0.7

109.1

209.3 8.0

2.1

125.9

292.3 19.6

Assumed Vertical = 0.6 x CahX

Same as above

16.0 8.0 12.0 6.0

33.0

60.0 27.0 15.8 35.0 5.0

5.0

32.0 11.3

Table 2. Ground accelerations for earthquakes listed per EqSF values.

CahX

CahY

CaV

Scah (2-D)

SCah

SCa (3-D)

SCa (3-D)

cm/sec/sec

cm/sec/sec

cm/sec/sec

cm/sec/sec

g

cm/sec/sec

g

2699.90

1268.50

1879.90

2983.04

3.04

3525.99

972.60

618.00

316.00

1152.33

1.17

1194.88

1597.60 686.70 927.20 639.61

656.00

1185.90 922.14 753.00 598.41

437.00

758.40

1969.20

842.50

793.20

1744.50 1307.91

912.00

852.80

1468.30

1019.40

1155.00 798.00

720.40

1030.00 580.30

554.50 551.60

849.00

875.90

788.23

1.22

0.89

0.80

1

1.22

Japan, September 2003

Hiroo HKD 120

16.8

3

1.45

Western Tottori, Jap., 2000

1.18

Valparaiso, Chile, 1985

2.26

Northridge, CA, 1994

1.21 1.06

1.59

12.9

10

1763.68

1.97

Cape Mendocino/Petrolia, CA

Cape Mendocino/Petrolia

El Salvador

Unidad de Salud, La Libertad

Landers, CA

Lucerne Valley

1.31

1334.34

1.36

716.80

383.00

1072.66

1.09

1138.99

1.16

579.00

278.00

368.00 486.80

723.00

384.20

488.30 451.10

1181.58 643.45

665.50 735.69

0.96

1.20

0.66

0.68 0.75

862.98

932.00

992.19

1.01

1361.27

325.00

1841.28 947.90

1066.73

0.97

1.09

595.25

0.61

341.70

185.00

144.00

388.57

0.40

156.00

304.11

564.10

313.92

467.99

1.88

500.10

961.38

825.43

1618.42

584.00

167.00

749.43

1.59

593.00

539.55

1385.23

1559.14

1438.10

268.00

1047.96

434.00

1163.60 535.30

942.10

1934.18

1.80

1289.77

459.00

1.82

370.62

735.82

694.48

1088.34

0.48

0.38

0.75

0.71

1.11

2336.33

1113.36

1115.14

1766.71

1.07

Western Tottori, Japan

Unidad de Salud, La Paz

2.38

Christchurch, New Zealand

Heathcole Valley

1.13

Northridge, CA

Sylmar, Converter

San Fernando, CA

1.39

Darfield, New Zealand

1.14

The EqSF approach is based on objective measurements and provides useful data for professional engineers. This method provides reliable comparable data for the forces of local ground shaking that will complement the current information available after an earthquake. The EqSF takes into account all three components of the ground accelerations plus the strong motion duration. The EqSF values can be calculated for all local ground stations with recorded seismograms. The force at a particular location can be compared with the force at any other locations during the

9.6

9.5 9.0

8.9

8.8

18 19

20 21 22

23

8.0

26

7.6

28

0.42

Capitola

El Centro Differential

0.91

Nisqually, WA

Seattle, SDW

El Centro, CA, 1940

South Napa, CA, 2014

same seismic event or previous events. The EqSF values could be provided for different locations after an earthquake, much like temperatures and rainfall recordings. Depending on the relative locations and specific soil characteristics, EqSF values will differ even between parts of a large city. These results will reflect the different levels of ground shaking and therefore will provide a more realistic representation of the event.

The EqSF method is devised to measure and compare the force of ground motion but is not intended to measure earthquake damage.

However, the EqSF method could be used to calculate and rank all ground motions recorded at particular locations from previous earthquakes, and used to create maps as visual logs of EqSF values based on previously recorded seismograms. The accumulated data may be used in the future by local building departments, insurance and real estate companies for better assessing the potential seismic danger and may help creating microseismic zones. Comparing the performance of different structural lateral resisting systems July 2015

El Centro, Station 9

Carquinez Br.,Array 1

8.2

17

Joshua Tree

Calexico, Mexico, 2010

19

10.2

16

Landers, CA

El Centro, Station 6

0.81

1.15

10.6

Sepulveda, VA Hospital

414.39

STRUCTURE magazine

Greendale

11.0

14

8.0

7.7

7.4

6.4

6.0

24

25

27

29 30

31

during earthquakes resulting in identical or similar EqSF values may help adjust the response modification factors (R) used in the seismic design for buildings and structures.

Further analysis and research using the EqSF scale may provide improved understanding of earthquakes and the resulting propagation of ground motion and damaging forces.▪

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Conclusions

Pacoima Dam

11.6

Northridge, CA

Loma Prieta, CA, 1989

0.41

NOTE: The largest three values for accelerations in each direction, as for the vector accelerations are highlighted in yellow.

on EqSF values was found to be less than [(2.3/1.7)^2]^0.2 = 1.12 (or less than 12%).

Considering the element of subjectivity and the small influence of the frequency variation on EqSF, determination (less than 12% compared to a larger percentage of non-accuracy from 15% to 20%), the frequency was abandoned as having a minimal effect on the calculated EqSF values.

15

Mitsugi

0.79

1132.70

11.3

El Salvador

1.65

11

12

Waimea Fire Sration Taichung TCU 079

0.88

12.6

12.3

9

13

Taiwan

Kure, Japan

13.5

7

11.9

0.76

0.84

14.0

6

Hakuta, Station SMNH01

Hawaii, October 2006

777.45

894.71

Ojiya, NIG019

1.41

Imperial Valley, 1979

796.18

Tarzana

Japan, 2004

1.80

402.11

5

Chiayi CHY 080

Llolleo

14.2

4

Taiwan

1.39

1561.38

15.6

8

1361.11

820.17

Hino, Station TTRH02

16.4

13.6

1.18

2220.54

Angol

2

Shiogama, MYG012

1157.14

2.01

Chile, Feb. 2010

18.8

Japan, March 2011

716.70

1968.51

Hitachi, IBR003

Concepcion San Pedro

2.21

1027.50

Japan, March 2011

Chile, Feb. 2010

2168.81

390.80

510.12

1158.49

23.4

2.15

449.00

674.10

1044.48

Tsukidate, MYG004

2110.19

1703.60

500.31

1424.28

Japan, March 2011

2.35

500.80

147.00

256.00

1184.41

3.59

342.00

607.10

663.50

444.30

568.98

1194.45

1.17

2305.98

Rank

574.00

737.70

922.70

775.80

1149.74

2.03

EqSF

1787.48

1055.00

739.50

284.49

1989.64

Location

738.90

1148.00

1427.00

1165.70

Event

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T

he 2015 Edition of Special Design Provisions for Wind and Seismic (SDPWS) was approved as an American National Standard on September 8, 2014, with the designation ANSI/ AWC SDPWS-2015 (Figure 1). The 2015 SDPWS was developed by AWC’s Wood Design Standards Committee (WDSC) and contains provisions for design of wood members, fasteners, and assemblies to resist wind and seismic forces. Some of the more notable revisions include the following, which are explained in more detail below (see Table 1, page 22, for a summary of changes by Chapter): • Expanded applicability of the wall stud repetitive member factor to stiffness values (EI) for calculating stud out-ofplane deflection under wind load for stud spacing up to 24 inches on center • Added a new section on wind uplift force resisting systems • Added a new section on seismic anchorage of concrete/masonry structural walls to wood diaphragms consistent with ASCE 7-10 Minimum Design Loads for Buildings and Other Structures • Updated diaphragm flexibility terminology to coordinate with ASCE 7-10 • Added “envelope” analysis as an alternative to “semi-rigid” diaphragm analysis for horizontal distribution of shear to vertical resisting elements (i.e., shear walls) • Consolidated formerly separate provisions for design of open front structures and design of cantilevered diaphragms under Section 4.2.5.2 for design of open front structures with cantilevered diaphragms • Added minimum 3-inch nominal depth requirement for framing and blocking used in high load blocked diaphragms consistent with the International Building Code (IBC) • Updated provisions for distribution of shear to shear walls in a line to clearly address stiffness compatibility and to clarify basis of the familiar 2b s /h factor • Added a new shear wall strength reduction factor for high aspect ratio wood structural panel shear walls applicable for wind and seismic design • Added a new method to account for strength of high aspect ratio perforated shear wall segments • Added a new table for anchor bolt spacing along the bottom plate of wood structural panel shear walls designed to resist combined shear and wind uplift

Repetitive Member Factor Applied to Stiffness Section 3.1.1.1 was revised to permit application of the wall stud repetitive member factor

Codes and standards updates and discussions related to codes and standards

Figure 1. 2015 SDPWS is referenced in the 2015 International Building Code.

for stiffness, EI, in calculation of out-of-plane deflection of wall studs for stud spacing up to 24 inches on center. The values of the repetitive member factor remain unchanged, ranging from 1.5 for 2x4 studs to 1.15 for 2x12 studs, and are applicable only where certain specific conditions are met including use of blocked wood structural panel sheathing.

2015 Special Design Provisions for Wind and Seismic

Wind Uplift Force Resisting Systems New Section 3.4 addresses wind uplift resistance and is based on general requirements of structural design to provide load path for structural loads. It describes general design considerations for proportioning, designing, and detailing members and connections resisting wind uplift. For example, load path elements must have adequate strength and stiffness, and their design must also account for additional forces and deflections resulting from eccentricities in the uplift load path.

Anchorage of Concrete or Masonry Structural Walls to Wood Diaphragms New Section 4.1.5.1 addresses the anchorage of concrete or masonry structural walls to wood diaphragms for seismic forces consistent with provisions of ASCE 7-10 Section 12.11 and prior editions of the building code where they originally appeared. The requirements in SDPWS include those for continuous ties, use of the subdiaphragm concept, and prohibition of anchorage force transfer through wood framing that could induce cross grain bending and cross grain tension (Figure 2, page 23). continued next page

STRUCTURE magazine

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By Philip Line, P.E., John “Buddy” Showalter, P.E., Michelle Kam-Biron, P.E., S.E., SECB and Jason Smart, P.E.

Philip Line, P.E., is Director of Structural Engineering, John “Buddy” Showalter, P.E., is Vice President of Technology Transfer, Michelle Kam-Biron, P.E., S.E., SECB, is Director of Education, and Jason Smart, P.E., is Manager of Engineering Technology with the American Wood Council. Contact Mr. Showalter (bshowalter@awc.org) with questions.

Table 1. Summary of Changes in 2015 SDPWS Chapter 1 – No Changes Chapter 2 – General Design Requirements 1) Add section 2.1.3 Sizes for how WSP nominal thickness relates to Performance Category in PS1 and PS2.

2) Definition for “subdiaphragm” is added to coordinate with new Section 4.1.5.1.

3) Remove definitions for Flexible and Rigid Diaphragms to avoid conflicts with varying definitions in ASCE 7-10.

4) Add definition for “open front structure” to coordinate with section 4.2.5.2.

5) Add notation for L´ and W´ for cantilevered diaphragms. Coordinating changes are made in notation for L and W and include deletion of L c.

6) Revise definition for collectors to clarify use for transfer of diaphragm shear forces to shear walls.

Chapter 3 – Members and Connections 1) Revise 3.1.1.1 Wall Stud Bending Design Value Increase to permit wall stud repetitive member factor for stiffness and for studs spaced up to 24 inches o.c.

2) Revise values and footnotes in Tables 3.2.1 and 3.2.2 to reflect that 3-ply plywood is not commercially available for thicker panels.

3) Revise Table 3.2.2 to add a case for roof sheathing strength axis parallel to supports to address a common technique in panelized roof construction.

4) Add new section 3.4 and modify section 3.2.1 to address wind uplift force resisting systems.

Chapter 4 – Lateral Force-Resisting Systems 1) Add new section 4.1.5.1 to address seismic anchorage of concrete or masonry structural walls to wood diaphragms consistent with ASCE 7-10.

2) Revise section 4.2.5 Horizontal Distribution of Shear to use terms “idealized as flexible”, “idealized as rigid”, and “semi-rigid” consistent with ASCE 7-10.

3) Revise section 4.2.5.1 Torsional Irregularity to clarify requirements and improve consistency with ASCE 7-10.

4) Consolidate Open Front Structures and Cantilevered Diaphragms into Section 4.2.5.2 Open Front Structures to clarify requirements and improve consistency.

5) Revise section 4.2.7.1.2 High Load Blocked Diaphragms to add minimum 3” nominal depth of framing/blocking.

6) Replace diaphragm configuration figures in Tables 4.2A, 4.2B 4.2C to illustrate that diaphragm resistance is dependent on the direction of continuous panel joints with respect to loading direction as well as direction of framing members, but is independent of the panel orientation.

7) Revise column headings in Table 4.2C to clarify 6" nail spacing is for supported panel edges.

8) Add new section 4.3.2.3 Deflection of Structural Fiberboard Shear Walls for such walls with h/bs > 1.0.

9) Revise section 4.3.3.4 and add new section 4.3.3.4.1 to establish equal deflection as the general requirement for distribution of shear to shear walls in a line. An Exception permits distribution of shear in proportion to shear capacity when certain conditions are met.

10) Revise section heading 4.3.4 to add “and Capacity Adjustments” to reflect section content.

11) In section 4.3.4.3, adjust the length of each perforated shear wall segment with h/bs exceeding 2:1 by 2bs /h.

12) Add a new section 4.3.4.2 for strength adjustment for high aspect ratio walls.

13) Revise 4.3.5.1 Individual Full-Height Wall Segments to remove reference to shear wall line.

14) Revise 4.3.5.3 to clarify materials requirements for the perforated shear wall design method.

15) Add new section 4.3.6.1.1 Common Framing Member permitting (2) 2x framing members to replace a 3x framing member and reference from 4.3.7.1(4) WSP Shear Walls and 4.3.7.3(4) Particleboard Shear Walls.

16) Revise 4.4.1 Application to clarify that the walls are designed to resist wind uplift and not the sheathing only.

17) Revise section 4.4.1.2 Panels to address panels with the strength axis parallel or perpendicular to studs.

18) Revise 4.4.1.6(2) by permitting increased anchor bolt spacing in accordance with new Table 4.4.1.6 for wood structural panel shear walls designed to resist combined shear and wind uplift.

19) Add new 4.4.1.6(3) to provide a minimum end distance for anchor bolts used for wood structural panel shear walls designed to resist combined shear and wind uplift.

Appendix A – None References – Update References

STRUCTURE magazine

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July 2015

Diaphragm Flexibility Terminology In Section 4.2.5, diaphragm flexibility terminology was revised to utilize the terms “idealized as flexible,” “idealized as rigid,” and “semi-rigid” consistent with ASCE 7-10. The condition for which a wood diaphragm is permitted to be idealized as rigid (e.g. in-plane deflection of the diaphragm is less than or equal to two times the average deflection of adjoining vertical elements) remains unchanged from prior editions of SDPWS. The significance of “idealized” is to recognize that wood diaphragms always have some rigidity, and are neither truly flexible nor truly rigid but can be idealized as such where certain conditions are met.

These idealizations are employed to simplify structural analysis for distribution of horizontal diaphragm shear loads.

The use of semi-rigid diaphragm modeling for purposes of distribution of horizontal force is always permissible under ASCE 7. It is the method considered to most rationally account for actual distribution of horizontal diaphragm shear loads to vertical resisting elements; however, a semi-rigid diaphragm analysis requires significant calculation effort for all but the simplest box structures. An acceptable alternative to semi-rigid diaphragm analysis is the envelope analysis where distribution of horizontal diaphragm shear to each vertical resisting element is the larger of the shear forces resulting from analyses where the diaphragm is idealized as flexible and the diaphragm is idealized as rigid. While two separate analyses must be performed, one for diaphragm idealized flexible and one for diaphragm idealized as rigid, the envelope analysis provides a conservative alternative means of shear distribution and avoids calculation effort associated with semi-rigid diaphragm modeling. Specific recognition of the envelope analysis method is new in SDPWS Section 4.2.5.

Torsional Irregularity and Open Front Structures Revised provisions of 4.2.5.1 (Torsional Irregularity) and 4.2.5.2 (Open Front Structures) reflect efforts to clarify requirements that include use of terminology that is more consistent with ASCE 7. A coordinated hierarchy of requirements has been established for seismic design whereby open front structures with cantilevered diaphragms are subject to increased limitations on story drift and building configuration when compared to provisions for torsionally irregular structures that are not open front. Open front structures

must rely on diaphragm rigidity for distribution of forces to vertical elements of the seismic force resisting system by diaphragm rotation. Such structures are considered to be more vulnerable to torsional response than other box-type structure configurations due to reliance on the diaphragm for torsional force distribution to elements that are not optimally located at diaphragm edges. A structure with shear walls on three sides only (open front) is one simple form of an open front structure; however, open front structure requirements are applied to alternative forms employing cantilevered diaphragms (Figure 3).

Revised provisions of 4.2.5.2 (Open Front Structures) remove ambiguity from prior editions of SDPWS, primarily by consolidation of separate sets of provisions previously applicable to structure types described as either “open front” or “cantilevered diaphragm.”

Under new provisions of 4.2.5.2, open front structures with cantilevered diaphragms are subject to increased limitations relative to torsionally irregular structures that are not open front. For example, open front structures with cantilevered diaphragms are subject to the following design limitations: • for loading parallel to the open side, the maximum story drift at each edge of the structure shall not exceed the ASCE 7

Tie Force T Tie Force

Tie Force

Tie Force

Potential ccross-grain failure bending failu a) Appropriate wall anchor detail where anchor forces are transferred directly into diaphragm framing

b) Inappropriate wall anchor detail where anchor forces induce cross-grain bending in the wood ledger (not permitted)

Figure 2.

Plan Views

Shear Wall

Force

W’

Cantilevered Diaphragm

L’

W’ L’

Open Front

Cantilevered Diaphragms Shear Wall

Force

(a)

L’

Open Front

(b)

Figure 3. Excerpt from 2015 SDPWS new Figure 4A showing examples of open front structures. In this case, an example of a) a simple open front structure with walls on three sides only and b) a simple corridor wall structure.

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STRUCTURE magazine

23

July 2015

4/28/15 7:57 AM

V

δ

8'

Figure 4. 2015 SDPWS includes provisions to coordinate with ASCE 7-10.

δ

SW1

SW2

8'

2.3'

Figure 5. Distribution of shear to shear walls in a line to produce equal deflection in each shear wall is the underlying basis of the familiar 2bs /h factor.

allowable story drift regardless of whether a torsional irregularity is present • where a torsional irregularity is present, the L´/W´ ratio shall not exceed 0.67:1 for structures over one-story in height, and 1:1 for structures one-story in height • the cantilevered diaphragm length, L´, (normal to the open side) shall not exceed 35 feet.

An exception to Section 4.2.5.2 exempts small cantilevers having L´ of 6 feet or less as a practical approach to avoid unnecessarily triggering special open front provisions where cantilevers are small. Similarly, provisions of new Section 4.2.5.2.1 permit simplification of analysis by allowing the use of the idealized as rigid diaphragm assumption for relatively small one-story structures with diaphragm span not more than 25 feet and the L´/W´ ratio not more than 1. While 4.2.5.2 provides requirements specific to wood diaphragms in open front structures, these are in addition to and not a replacement of seismic design criteria of ASCE 7 (Figure 4).

of shear to shear walls in a line. While this concept is not new to SDPWS, the organization of requirements pertaining to distribution of shear to shear walls in a line is new.

New Section 4.3.3.4.1 states that “Shear distribution to individual shear walls in a shear wall line shall provide the same calculated deflection, δsw, in each shear wall.” At a given deflection, the force in each wall is determined by multiplying the wall stiffness times the deflection (commonly referred to as distribution based on relative stiffness or the equal deflection approach). A simplified approach permits distribution of shear in proportion to the nominal shear capacities of the individual full-height wall segments, provided that certain requirements are met. For wood structural panel shear walls, distribution of shear in proportion to the nominal shear capacity of each shear wall segment is permitted provided that the nominal shear capacity is adjusted by a factor of 2bs /h for wall segments with aspect ratios greater than 2:1. This factor is based on reduced stiffness observed from High Load Blocked testing and provides roughly similar results to the equal deflection calculation method Diaphragms for a reference wall line configuration comSection 4.2.7.1.2 on high load blocked prised of a 1:1 aspect ratio shear wall and a diaphragms now clarifies requirements for 3.5:1 aspect ratio shear wall, as depicted in minimum depth of framing members and Figure 5. Whether there is a strength benblocking consistent with similar provisions efit in one method over the other depends for stapled high load diaphragms in 2015 IBC on the specific wall configuration under Table 2306.2(2) footnote (e). Section 4.2.7.1.2 consideration. item 4 states: “The depth of framing members A common misunderstanding of the 2bs /h and blocking into which the nail penetrates factor was that it represented an actual reducshall be 3 inches nominal or greater.” tion in unit shear capacity for high aspect ratio shear walls. The actual strength reduction Distribution of Shear to Shear associated with high aspect ratio shear walls is less severe and addressed by new Section Walls in a Line 4.3.4.2. The 2bs /h factor accounts primarily Provisions of Section 4.3.3.4.1 contain the for stiffness compatibility of the high aspect equal deflection requirement for distribution ratio segment. Where 2bs /h is used to comply STRUCTURE magazine

24

July 2015

with load distribution requirements of Section 4.3.3.4.1, the strength reduction adjustments of 4.3.4.2 for high aspect ratio shear wall segments need not be applied.

Strength Adjustment Factor for High Aspect Ratio Walls As noted previously, Section 4.3.4.2 contains a new strength adjustment factor to account for the decreased unit shear capacity of high aspect ratio wood structural panel shear walls. The new factor, 1.25 – 0.125 h/bs, is applicable to shear walls with an aspect ratio greater than 2:1. As previously noted, where distribution of shear is based on the simplified alternative adjustment factor methods (e.g.

2bs /h for wood structural panels), further reduction of shear strength by the aspect ratio factors in 4.3.4.2 is not required.

Requirements of 4.3.4.2 are in addition to those in 4.3.3.4.1 to ensure deflection compatibility between shear walls in a line and, therefore, the smaller of the design capacities associated with requirements of 4.3.4.2 and 4.3.3.4 is to be used as the controlling design capacity for each individual shear wall.

High Aspect Ratio Perforated Shear Wall Adjustments Provisions for accounting for the strength contribution of high aspect ratio shear wall segments within a perforated shear wall have been revised. In prior editions of SDPWS, where a high aspect ratio perforated shear wall segment (e.g. h/bs >2:1) was considered in the calculated strength of the perforated shear wall, the shear capacity of the overall perforated shear wall required adjustment by the 2bs /h factor. The revised provisions of Section 4.3.4.3 allow the adjustment to apply only to the high aspect ratio perforated shear

wall segments, based on stiffness compatibility considerations, as opposed to the calculated strength of the overall perforated shear wall. Because the more severe 2bs /h factor is used, unit shear values of high aspect ratio shear wall segments within a perforated shear wall are not required to be adjusted by the aspect ratio factors of Section 4.3.4.2. While this revised method will generally permit increases in design strength of perfo- Figure 6. Excerpt of 2015 SDPWS new Table 4.4.1.6. rated shear walls incorporating high aspect ratio segments, there are cases where 4.4.1.6, anchor bolt spacing varies from 16 there is little change such as perforated shear to 48 inches on center, based on the nominal walls comprised entirely of identical high uplift capacity of the wood structural panel aspect ratio perforated shear wall segments. sheathing or siding.

Anchor Bolt Spacing for Combined Shear & Wind Uplift

Conclusion

The 2015 SDPWS is currently available as Section 4.4.1.6(2) regarding anchorage of a free download in electronic format (PDF) bottom plates and sill plates to resist combined as a non-printable read-only document, and uplift and shear was revised to permit determi- a printable electronic version is available for nation of anchor bolt spacing in accordance purchase (www.awc.org). Additional inforwith a new Table 4.4.1.6 (Figure 6) developed mation on SDPWS provisions is available in based on testing and analysis. Previously, an the SDPWS Commentary. The 2015 SDPWS anchor bolt spacing of 16 inches on center, Commentary is scheduled to be available in associated with the maximum wind uplift June 2015. The 2015 SDPWS represents the capacity, was prescribed. Using new Table state-of-the-art for design of wood members ADVERTISEMENT–For Advertiser Information, visit www.STRUCTUREmag.org

STRUCTURE magazine

25

July 2015

and connections to resist wind and seismic loads. Reference to the 2015 SDPWS in the 2015 IBC will make it a required design standard in those jurisdictions adopting the latest building code.▪ A note from the NCSEA Code Advisory Committee: NCSEA, through its member organizations (MOs), often contributes to the basis of code changes. SEAOC (the California MO of NCSEA) publishes articles that comprise the “Blue Book” a significant source of engineering consideration of code provisions and important earthquake engineering issues. For more information, please explore www.seaoc.org/bookstore.

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C

oncentrically braced frames (CBFs) resist large lateral forces due to wind and earthquake loading, and their ductility is largely derived from tension yielding and compressive buckling of the braces. Since 1990, AISC has focused on improving seismic resistance of CBFs by introducing detailing requirements for the connection, geometric limits of the brace, and capacity-designtype strength requirements for the gusset plate and the framing members.

In 1997, a requirement was introduced that the strength of the beam in a frame with braces in a chevron configuration must be able to withstand the post-brace buckling unbalanced forces, based on the assumption that the braces in tension yield and the braces in compression resist 30% of their critical buckling loads (AISC 2010).

However, many braced frames built prior to the mid-1990s do not meet this requirement and therefore are generally considered to be “weak” and substandard. A research project to address the performance and potential retrofit of these older braced frames is currently in progress.

The article How Big is that Beam? (STRUCTURE magazine, November 2014) addresses this beam strength issue. The author analyzed three two-story

frames using a nonlinear pushover analysis, shown in Figure 1a. The analyzed frames included a weak-beam chevron braced frame (Frame 1), a strong-beam chevron braced frame (Frame 2), and a multi-story X-braced frame (Frame 3); the results are repeated in this article as Figure 1b. Frames 2 and 3 were found to perform adequately, but Frame 1 exhibited a significant loss in strength and stiffness after brace buckling. As seen in the Figures, the predicted response of each braced frame system is characterized by sudden drops in strength, which is not characteristic of braced frame tests with adequate connections.

The 2014 article thereby concludes that a frame with an undersized beam results in poor frame performance. Further, the analysis suggested that Frame 2, with a strong beam and HSS braces meeting current AISC SCBF slenderness limits, may achieve an inelastic deformation of about 12 times the buckling deformation. The author did not provide dimensions for the frames reproduced in Figure 1a, but assuming a story height of 12 feet, this deformation corresponds to 4.0% to 4.8% average story drift.

This assumed story height

Structural ForenSicS investigating structures and their components

How Big is that Beam?

Revisited

(a) Frames From Prior Work

By Andrew D. Sen, Charles W. Roeder, Ph.D., Dawn E. Lehman, Ph.D. and Jeffrey W. Berman, Ph.D.

Andrew Sen is a doctoral student in structural engineering at the University of Washington. He can be reached at adsen@uw.edu.

(b) Pushover Curves from Prior Work

Charles Roeder, Ph.D., is a Professor of Civil Engineering at the University of Washington.

He can be reached at croeder@u.washington.edu.

Dawn Lehman, Ph.D., is a Professor of Civil and Environmental Engineering at the University of Washington. She can be reached at delehman@uw.edu.

Jeffrey W. Berman, Ph.D., is the Thomas & Marilyn Nielsen Associate Professor in the Department of Civil and Environmental Engineering at University of Washington. He can be reached at jwberman@myuw.net.

Figure 1. Frames and predicted response from prior work.

STRUCTURE magazine

27

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aligns with dimensions common in practice and corresponds to approximately 0.3% average story drift at brace buckling, which is typical of braced frames in experiments. While the observations from these analyses are logical based on the results presented, this article contends those results are dependent on numerous assumptions that mischaracterize CBFs. The consequences of these assumptions are clear in the experimental results described below, which vary drastically from the results in the previous work.

The authors have conducted tests that reflect the three categories of braced frames considered in the November article, and these experiments are discussed here. These tested frames include a chevron braced frame with a weak beam (left column of Figure 2), a chevron braced frame with a strong beam (center column of Figure 2; Sen 2014), and a multistory X-braced frame (right column of Figure 2; Lumpkin 2009). Figures 2b and 2c present the experimental cyclic and backbone curves of the three frames. The braces in all three frames were identical (HSS5×5×3/8).

While the member sizes of these frames are different than those shown in Figure 1, the relative strengths of the various members are similar. The Figure 2. Comparison of Weak-Beam Chevron, Strong-Beam Chevron, and Multi-Story X Frames. frames were loaded quasistatically under a fully reversed increasing amplitude The curves in Figure 2b show that there ultimate strengths of the chevron braced cyclic protocol. The base shear is normalized by was little difference in the strength and frames were lower, although the multi-story the base shear force corresponding to buckling, stiffness of the weak- and strong-beam X frame had heavier columns, so some addiVbb . Equation 1 shows Vbb as a function of the chevron frames. This is not intuitive, tional strength would be expected. buckling capacity of the brace, Pcr , and the because the first story braces did not yield Another significant point of comparison brace angle, θ. in tension while the second story braces is that the weak-beam chevron frame had did. However, frame action provided more approximately the same deformation capacVbb = 2Pcr cos θ Equation 1 lateral resistance on the first story than the ity as the multi-story X frame prior to brace second story. The chevron braced frame is fracture. Deformation capacity is discussed an indeterminate system. While the truss in terms of drift range here, because local CADRE Pro 6 for Windows assumption is suitable for initial design, it cupping deformation at the brace mid-span is not valid for seismic evaluation due to in compression typically precipitates fracSolves virtually any type of structure for the complex inelastic behavior of CBFs. ture of the brace in tension; hence, both internal loads, stresses, displacements, System capacity is dependent on the plastic directions of loading influence the ultimate and natural modes. Easy to use modeling tools including import from CAD. Much mechanism, which is incomplete in models response. The weak-beam chevron frame more than just FEA. Provides complete that neglect beam-to-column connection reached a story drift range of 4.12% prior structural validation with advanced resistance and consequently lateral resis- to brace fracture; this is comparable to the features for stability, buckling, vibration, tance due to frame action. This modeling maximum story drift range of the multishock and seismic analyses. simplification is pronounced in a weak- story X-braced frame, which was 4.38%.

CADRE Analytic beam frame, where the plastic mechanism Thus, the weak beam did not impact system Tel: 425-392-4309 is yielding of the beam and buckling of the drift capacity significantly. Finally, it must be www.cadreanalytic.com brace. For a strong-beam frame, the plastic noted that none of the HSS braces achieved mechanism is yielding of the brace in ten- the large story drifts predicted for Frames sion and buckling of the brace. Compared 1 and 2 in the prior work. HSS braces of to the multi-story X-braced frame, the this slenderness fracture at maximum story STRUCTURE magazine

28

July 2015

Acknowledgements This material is based upon work supported by the National Science Foundation Network for Earthquake Engineering Simulation under Grant No. CMMI-1208002, Collaborative Developments for Seismic Rehabilitation of Vulnerable Braced Frames and Graduate Research Fellowship under Grant No. DGE-1256082. Any opinion, findings, and conclusions or recommendations expressed in this material are those of the authors and do not necessarily reflect the

views of the National Science Foundation.

Additional support was also provided by the American Institute of Steel Construction in the form of steel donations.

The online version of this article contains detailed references. Please visit www.STRUCTUREmag.org.

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STRUCTURE magazine

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July 2015

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drifts in the order of 2.5%, with a maximum drift range of 4.5% to 5%. This has been documented in hundreds of prior tests (e.g., Tremblay 2002 and Fell et al.

2009). Other brace cross sections, such as wide flanges, may achieve larger drift levels, but 4.0% average story drifts in tension are not achieved with HSS braces subjected to cyclic loading, as in earthquakes, since this implies much larger drift range (e.g., 8.0% if demand is balanced). Pushover analysis fails to capture the cyclic deterioration of the brace and is therefore unsuitable for predicting the deformation capacity of the system.

A number of factors likely contribute to the discrepancies observed between the pushover analysis presented in the prior work and the experimental results described here. Braced frames are commonly modeled as trusses, but it is clear from the test results that considerable resistance is derived from moments and shear forces that develop in the beams and columns. Also, while neither the monotonic nor cyclic protocols represent real earthquake deformation demands, cyclic protocols simulate the load reversals that an earthquake may induce in a structure. This enables phenomena such as strain hardening and low-cycle fatigue to occur, which are important response characteristics. Finally, springs or concentrated hinges that are used in some nonlinear structural analysis software may not be suitable for capturing the material and geometric nonlinearities that develop in braced frames; fiber-based or shell elements are required. Numerical results should always be interpreted with caution, and the limitations of software and input parameters (e.g., plastic hinge definitions) should not be overlooked.

Much research still needs to be conducted to determine the viability of weak beams in older and modern chevron braced frames, but these preliminary results suggest that good performance may be achieved using shallower, lighter beams than permitted by the current Seismic Provisions. This is in stark contrast to the results of simplified analyses of the How Big is That Beam? article, but the actual post-buckling behavior of braced frames is complex and difficult to capture without employing more robust analysis techniques.▪

Center for naval aviation teChniCal training Complex Integrating 3 Ds in 3D is Key to Project Success

By Mehdi Rashti, S.E.

Figure 1. CNATT Helicopter Maintenance Hangar at night.

I

ntegrating 3D modeling software programs for the three “Ds” for this project – Ram for design, Revit for drafting and Tekla for detailing – proved to be the key for success for this fast track design-build project at a major Marine Corps base in California.

The project consists of three adjoining structures, with very diverse functions, which form the new 131,000 square-foot Center for Naval Aviation Technical Training Complex (CNATT). The primary function of the complex is to house all aviation mechanics involved with the maintenance and repair of Huey and Cobra helicopters at the Marine Corps base at Camp Pendleton. This includes administrative activities, training, and the actual maintenance and repair of the helicopters in an open-bay hangar area.

Aesthetically, one structure, rather than three, would have served the purpose. However, due to the diverse programming needs of each activity and structural system incompatibility, the team decided it was necessary to design three buildings so there could be seismic separations between them. The tall hangar building with a braced frame lateral load resisting system is one building (Figure 1); the four-story administration building with masonry shear walls is the second building; and a three-story classroom masonry shear wall building with an open space and column-free auditorium on the first floor is the third building. The team individually modeled and analyzed each building separately to manage the electronic file sizes. This also allowed several individuals to work on the same model at the same time through the work-sharing feature of Revit.

The need to quickly double the size of its helicopter maintenance and repair training facilities was “mission critical” for the Marines in order to meet the rapidly growing demand for Huey and Cobra helicopters used in global combat operations. Therefore, the Marine Corps put the project on a fast track schedule and selected design-build as the STRUCTURE magazine

method of project delivery. The Corps also wanted the project to be a showcase design to reflect the vision of CNATT as “the preeminent leader in aviation maintenance training.”

These goals presented significant challenges to the designers and builders. The accelerated schedule meant the team needed to start work immediately based on the parameters of the project outlined in the Request for Proposals (RFP). Another critical challenge was figuring out how to shorten the length of time required to fabricate a 320-foot span truss for the hangar building so the entire project could be designed and built within the allotted time frame. It also meant the design-build team had to resolve several potential problems during the planning and design phases of the project rather than address them later during the construction phase, as is normally the case. For instance, the team needed to overcome the problem of limited site access so that several heavy-duty cranes could be positioned to lift the hangar door truss. Another important issue to resolve involved the sequencing of field-assembled members for storage and erection.

Use of BIM The team relied on 3D Building Information Modeling (BIM) software to help speed up the design timeframe and improve the quality control process. In order to achieve these goals, the team developed the following system. First, the engineers used the architectural Revit model to add structural members without any particular attention paid to the sizing of the members. The second step required them to export the Revit model to RAM/Risa for structural member sizing.

Once they finished this task, the engineers then brought the analyzed model back to Revit for preparation of construction documents before finally exporting the documents to Tekla for steel detailing.

30

July 2015

Figure 2. Helicopter Maintenance Hangar.

During the initial phase of the project, the structural team proposed a system of open web steel joists for the framing of the hangar roof system. As the work progressed, they realized they needed larger openings for the mechanical, electrical, and plumbing (MEP) lines to pass through the truss web members in the hangar building. This changed the direction of the hangar roof framing system from an open web steel joist type of truss to a custom-designed roof truss with fewer webbing members.

Using Revit to model the actual truss member sizes and Tekla for the connection plates for the hangar roof truss connections allowed for detailed coordination among all the disciplines. Similarly, by modeling all of the masonry wall, roof, and floor-framing members in the administration and classroom buildings, it was possible to coordinate the exact location of MEP penetrations through the

shear walls to avoid conflicts during construction. To facilitate this process, the team held weekly BIM coordination meetings among all the stakeholders, including the steel fabricator, to expedite the critical decision-making on the changes needed during the design work schedule. These meetings and coordination efforts saved money by avoiding conflicts between the various systems during construction.

The result was the development of “No-Fly Zones” which are designated areas that are off-limits for contractors to position their pipes, duct, or other penetrations in the masonry shear walls in the administration and classroom buildings. Basically, the “No-Fly Zones” ensured that the inadvertent placement of unwanted holes or penetrations in critical sections of the walls would not affect the major components of the wall such as the chord bars, control joints, and beam pockets.

Design Challenges Hangar The anchor feature of the aviation training complex is a high-bay hangar space to house nine helicopters. The basic structural system design for the hangar portion of the building includes a corrugated metal roof deck spanning between steel purlins (Figure 2). Monoslope custom steel trusses (transverse trusses), spaced at 25-foot intervals, support the steel purlins (Figure 2b, page 33). A three-dimensional space truss at the front of the hangar, coupled with lattice-type “box” columns, support the header truss along the front of the hangar (Figure 2a). The 3D space truss, which is 10 feet wide and 16 feet deep, spans 320 feet along the front of the hangar, providing a clear, unobstructed entry for normal operations of helicopters. A single truss supported by individual columns provides support at the rear continued on page 33

Figure 2a. Box truss.

STRUCTURE magazine

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July 2015

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Figure 2b. Transverse truss.

Figure 2c. Single-ply truss.

of the hangar (Figure 2c). The transverse trusses cantilever out 12 feet on the rear side of the hangar to reach the outline of the adjoining administration building and to provide enclosure between the two buildings. Masonry exterior walls enclose the hangar door pockets.

A system of horizontal trusses in the plane of the bottom chord of the transverse trusses provides diaphragm action for distribution of lateral loads to the supporting brace frames. Two independent, 5-ton bridge cranes covering the entire hangar bay provide weight handling in the operation of the facility. Rail attachments to the bottom chord of the transverse trusses support these cranes.

The hangar space required clear, unobstructed entry to allow for the normal operations of the helicopters. In order to accommodate a 29-foot clear hook height in the hangar bay, the structural engineers had to evaluate the ”box” truss for various shapes and sizes to minimize the tonnage. This was particularly challenging because they were dealing with very limited depth available for the “box” truss at the

front of the hangar. The designers also analyzed the transverse trusses supporting the bridge crane rails, resulting in the addition of strategic strengthening members within the truss to limit the deflection to L/600 for proper operation of the crane rollers. The final design was a 320 foot-long span “box” truss with a weight of 250 tons.

Administration Building Adjacent to the hangar building is a four-story reinforced concrete masonry administration building which forms one leg of the “L”-shaped structure. This building has a composite floor system, interior steel columns, and exterior bearing reinforced masonry shear walls. The second-floor elevation is 16 feet and the subsequent floorto-floor height is set at 14 feet. The biggest challenge for the design team for this building was to get natural lighting into the upper-level classrooms in order to meet the LEED natural day lighting requirements. This meant the team had to design the roof of the hangar and

Figure 3. Column-free, multi-purpose auditorium.

STRUCTURE magazine

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July 2015

Figure 3a. Construction photo inset of 5-foot deep tapered steel girder.

Figure 3b. Construction photo inset of 8-foot deep concrete transfer girder.

the height of the administration building in such a way as to allow for natural light to penetrate the upper-level classrooms.

girder. Holes in the 5-foot deep tapered steel girder accommodated passage of the utilities from one side of the auditorium to the other.

Classroom Building The classroom building, called The Applied Instructional Facility (AIF), forms the other leg of the “L” shaped building. It is separated from both the hangar and the administration building. The AIF building includes electronic classrooms for up to 400 students, aviation training shop laboratories, and a column-free auditorium at the first floor with a seating capacity of 250.

Interior steel columns and the exterior masonry walls form the support of the structure at the upper levels. The interior steel columns and the exterior masonry walls terminate at the second floor to allow for a large column-free multi-purpose auditorium (Figure 3). In order to transfer loads, the engineers designed an 8-foot deep concrete transfer girder spanning 80 feet for the support of the 3-story exterior masonry wall above the auditorium (Figure 3b) and a 5-foot deep tapered steel girder to support the upper level interior steel columns (Figure 3a). The designers limited the deflection of the concrete and steel girders to L/600 to avoid cracking in the masonry walls and excessive defection in the steel

Steel Fabrication and Construction The designers, contractors, and fabricators needed to make an important decision about the means and methods of construction because they were dealing with two difficult field issues. The first challenge involved access to the site, which was very limited. The second challenge involved the difficulty of positioning the cranes required to lift the main entry “box” truss of the hangar. In the end, the team decided it was more desirable to field-bolt the box truss rather than shop-weld it, because field-bolting would provide more flexibility for fabrication of smaller pieces and assembly in the field.

To achieve some level of economy, the smaller transverse trusses were fabricated into two pieces in the shop and field assembled into a single truss in the field, and then lifted into position as a single piece oneby-one using a single 200-ton crane (Figure 4 ). For the box truss, the fabricator shop-welded connections for the top and bottom “ladder” chords into five segments. These chord segments, as well as all the webbing members, were field-bolted for the entire truss assembly to be put together and lifted into position using two 200-ton cranes. The decision to design and build the entire truss on site enabled the team to trim nearly 12 weeks from the original design/construction schedule.

Conclusion The design and construction process is constantly evolving. 3D modeling available through BIM has only been around for a few years. In a few more years, designers and builders most likely will simultaneously access cloud-based information in real time. At this point in time, however, the most valuable lesson learned by this structural engineering team is a time-honored, practical lesson. And that is, the best way to avoid any field coordination problems during construction is to identify and deal with critical design and construction issues immediately at project start.▪ Mehdi Rashti, S.E., is the Chief Executive Officer and founding principal of the SMR Consulting Group in San Diego, CA, the parent company of SMR-ISD Consulting Structural Engineers, Inc. He is a member of the Structural Engineers Association of California (SEAOC).

Photos by Pablo Mason, courtesy of Harper Construction.

Figure 4. Field-fabricated truss for hangar.

STRUCTURE magazine

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July 2015

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Figure 1. Assumed distribution of shear stress.

I

t is a well-understood concept and an inevitable law of statics that loads must be transferred between beams and columns.

This is an idea that is not foreign to engineers. Since our first classes in structural analysis, we have been developing our expertise at analyzing and designing beam-column intersections.

The idea of balancing the sum of forces at such locations is one that cannot be disputed. However, there are some vagaries with this concept when considering flat plate and flat slab systems, which by definition have no beam-column joints. Hence, moment transfer becomes a more complex issue.

Some engineers choose to ignore it, and they may even be rationally justified in doing so. This certainly simplifies the analysis, requiring only a two-way punching shear calculation. When does an engineer need to consider the transfer of moments between flat plates/slabs and columns?

This is a difficult question, one for which the past several versions of ACI 318 – including ACI 318-14 – do not provide a direct answer.

Section 8.4.2.3.1 states: “If gravity load, wind, earthquake, or other effects cause a transfer of moment between the slab and column, a fraction of Msc, the factored slab moment resisted by the column at a joint, shall be transferred by flexure in accordance with 8.4.2.3.2 through 8.4.2.3.5.”

Later on, section 8.4.3.2.6 states: “The fraction of Msc not calculated to be resisted by flexure shall be assumed to be resisted by eccentricity of shear in accordance with 8.4.4.2.”

Sections 8.4.2.3.2 through 8.4.2.3.5 address the development and proportioning of the Msc force as applied flexural and shear forces originating from the factored slab moment. The distribution of shear stress at the critical perimeter might then follow a pattern as depicted in Figure 1.

A valid question at this point might be when transfer of moments does not occur. Of course, there is no such thing as a perfectly balanced load; eccentricities and similar phenomena are always present. Moments will never perfectly balance from one side of a column to another, regardless

of the precise geometries. Codes even prescribe such things as ‘accidental’ eccentricity that must be included in design due to unbalanced moments that cannot be explicitly accounted for.

The language in codes (and concrete textbooks) seems to indicate that there is a point at which flexural and shear loads due to transfer moments from slabs to columns must be considered. However, no reference clearly indicates what that point is. To be conservative, we could always account for it. There is nothing wrong with this, except for the extra time that it requires in an often lean atmosphere of budgets and schedules. Leading researchers have stated that this transfer of moments should be considered whenever it becomes significant to the analysis – just a more complex and equally vague reiteration of what ACI 318 already states! Others have indicated that at an unbalanced column moment in excess of 10% from one side of the slab to the other is an appropriate trigger for requiring this level of analysis.

As ACI 318 is currently written, it is contingent upon us as design engineers to exercise appropriate judgment in determining whether the transfer of moments between the slab and the column is significant. In some cases, such as edge/corner columns or slabs with inconsistent bay sizes, there will clearly be a greater likelihood for significant transfer moments at this interface. For interior columns with consistent bay spacing, continuity of the slab and compatible stiffnesses between adjacent bays may predicate a transfer of flexure small enough to be ignored.

This may be true especially for concrete shear wall structures or projects in regions of lower seismicity (Seismic Design Category A, B, or C), where the effects of deformation compatibility issues per ASCE 7-10 section 12.12.5 are not significant. However, we must make such a determination on a case-by-case basis.▪

STRUCTURE magazine

Transfer of Moments in Slab-Column Connections

37

By Jerod G. Johnson, Ph.D., S.E.

Jerod G. Johnson, Ph.D., S.E. (jjohnson@reaveley.com), is a principal with Reaveley Engineers + Associates in Salt Lake City, Utah.

A similar article was published in the Structural Engineers Association-Utah (SEAU) Monthly Newsletter (January 2012). Content is reprinted with permission.

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PIONEERING PIONEERING PIONEERING SHAKING SHAKING SHAKINGTABLE TABLE TABLE CONTINUES CONTINUES CONTINUES TO TO TO BE BE BE INNOVATIVE INNOV INNOVATIVE ATIVE

By Grace S. Kang, S.E.

I

n 1968, civil engineering faculty at the University of California, Berkeley (UC) proposed to construct a 100-foot by 100-foot welded steel shaking table, weighing 2.2 million pounds. It was to be hydraulically powered by 60 to 72 actuators, allowing operation in three translational degrees of freedom, with an acceleration capacity of ⅔ g. The proposed shaking table was designed to allow researchers to experimentally test and identify the earthquake behavior of large-scale structures weighing up to 4 million pounds.

However, technical uncertainties and economic constraints related to then-available servo-valves, hydraulic actuators and control systems, the desired operational displacement limits of the table, and the ability to limit overturning rotation and flexure of the table, resulted in the design and construction of a prototype shaking table that would serve as a test bed to assess and improve shaking table technology.

The as-built, 20-foot by 20-foot reinforced concrete shaking table was officially dedicated in 1972, shortly after the 1971 Sylmar (San Fernando) California earthquake (M6.6). Since its 1972 inauguration, the Shaking Table, located at the Richmond Field Station of UC Berkeley and managed by the Pacific Earthquake Engineering Research (PEER) Center, has been very active and continues to introduce and employ innovative technology, and carry out cutting edge research related to the seismic behavior of structures and equipment. Two images of the table during construction are shown in Figures 1 and 2.

Figure 1. Installation of actuators.

Largest in the United States The PEER-UC Berkeley Shaking Table was the first modern shaking table and remains the largest six-degree-of-freedom (6 DOF) Shaking Table in the United States. Although the Shaking Table was originally configured to produce only one horizontal and one vertical component of motion, currently it is able to subject test specimens to three translational components of motion: vertical and two horizontal; plus three rotational components: pitch, roll and yaw. These 6 DOF can be programmed to reproduce any waveform within the force, velocity, displacement and frequency capabilities of the Shaking Table system.

The Shaking Table can subject structures to horizontal accelerations of up to 1.5g. Models up to 150,000 pounds have been successfully tested.

The Shaking Table is constructed of heavily reinforced concrete, utilizing both mild-steel reinforcement and post-tensioning tendons.

Structurally, the Shaking Table may be considered as a one-foot thick diaphragm stiffened by central transverse ribs that extend below the table’s bottom surface. The eight hydraulic actuators that drive X and Y motions, along with yaw, are attached between the Shaking Table foundation and the table’s transverse ribs. The four vertical actuators which control pitch, roll and vertical displacement, are bolted between the foundation and the Shaking Table. The test structure is attached to the table by post-tensioning rods, inserted through a 3-foot by 3-foot matrix of 2⅝-inch conduits, penetrating the Shaking Table surface. The length of the actuator assemblies, 8 feet-8 inches in the vertical direction and 10 feet-6 inches in the horizontal direction, serve to effectively de-couple the degrees of freedom motion. The high performance capabilities of the actuators, along with corrective commands provided by the sophisticated controller, complete the de-coupling. The Shaking Table itself weighs 100,000 pounds.

STRUCTURE magazine

Figure 2. Table top installation.

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Figure 3. Testing of reinforced concrete shearwall building model.

Testing and its Impact on Structural Engineering The PEER-UC Berkeley Shaking Table provides the engineering community and industry the unique capability of subjecting large form factor, high mass test structures to large amplitude, simultaneous earthquake input signals in both horizontal directions (X & Y) and vertically (Z). The unique XYZ capability of the Shaking Table STRUCTURE magazine

allows the academic community to characterize the performance of a given structure using realistic ground motions.

This shaking table ushered in a period of more complex test specimens – specimens graduated from planar models to three-dimensional models where the dynamic response (accelerations and drifts) of the assemblies are monitored in multiple directions. Testing of wood framed buildings, particularly with soft-stories, and testing of reinforced concrete model structures has been conducted. Successful

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experimental testing of a wide range of engineered assemblies has also taken place on the Shaking Table, including large-scale reinforced concrete frame and wall structures, multi-story steel frame structures, various types of braced frame structures, masonry buildings, tubular steel off-shore drilling platforms, bridge and dam models, piping systems, industrial and commercial racking systems, seismic dampers, soil-structure interaction of pile supported systems, and various base isolation systems for buildings, bridges and industrial storage tanks.

Figure 3 (page 41) shows a shear wall building model during testing.

The table was pivotal in the proof-of-concept testing of energy dissipation devices, which were not readily accepted by the structural engineering profession before the 1980s. The first test was of a steel energy-dissipating device in 1977, followed in the 1980s by rubber bearing isolators, high-damping rubber, and lead core rubber bearing isolators. These isolators were mounted on 3, 4, 5, and 9-story steel frame building models, and this testing was essential in demonstrating that the concept of energy dissipation could be beneficial to the seismic performance of the superstructure. This tested technology is designed and installed in numerous retrofits of critical-facility and landmark buildings, as well as new hospital and government buildings. Recent testing has been conducted with the devices on elevated building floors.

Friction dampers, polymeric dampers, and visco-elastic dampers were also tested in the late 1980s in building models. These devices have been designed and installed in many buildings in the United States and Canada. More recent research supported by industry led to testing of piping systems using energy dissipation support devices to reduce damage by limiting drift.

Shake table testing of steel and concrete building models led to the development of physical data and hysteretic models for cyclic behavior.

Testing also contributed to the identification of dynamic effects on structural systems such as soft-story behavior, and the dynamic amplification of shear forces. Analytical models that had been developed could be calibrated with empirical testing data. Research modeling and testing that simulated the collapse of older buildings led to the development of acceptance criteria that was a predecessor to ASCE 41-13 Seismic Evaluation and Retrofit of Existing Buildings.

Seismic Qualification and Other Testing Capabilities Additionally, the Shaking Table has been used in the seismic qualification testing of many specialized pieces of commercial and industrial equipment installed globally. The PEER-UC Berkeley Lab is accredited by the International Accreditation Service to perform panel testing in accordance with ASTM E2126, and beam-column testing and steel frame testing in accordance ANSI/AISC 341 Chapter K. The lab is also accredited to perform both AC156 and IEEE693 test protocols utilizing the Shaking Table. The simultaneous XYZ motion and high capacities of the Shaking Table allows industrial clients the capability to complete a given AC156 or IEEE693 code-compliant qualification test at the highest code required test amplitude, in a single testing configuration.

A 10-ton bridge crane services the Shaking Table lab, and the recent installation of a 19-foot by 27-foot rolling skylight installed directly above the Shaking Table allows for the testing of unlimited height structures. The unlimited testing height capability has been employed on numerous occasions to perform testing that historically has not been able to be completed due to vertical height limitations.

Recent seismic tests have included stacked wine-barrel racks (following the 2014 South Napa Earthquake), high-voltage electrical porcelain insulators, laboratory refrigerators, mechanical units, and switchgear that are components of critical function facilities.

STRUCTURE magazine

Figure 4. Hybrid testing of tuned mass damper (TMD) system. Load cells and isolators are located under corners of the blocks under the yellow frame. The TMD is assumed to be on top of a tall building.

Advanced Computational Enhancements – Hybrid Testing PEER is again upgrading its Shaking Table. In the past, shaking tables have simply been used to reproduce as accurately as possible predetermined motions. PEER researchers have been at the forefront of developing smart or hybrid shaking table technologies, and these are now being implemented on the shaking table.

In this case, the Shaking Table platform is considered to be a part of the overall system to be tested. The platform rests on a part of the overall structure, which is represented by a finite element model.

Thus, the physical part of the structure supported on the shaking table platform is shaken by, but interacts with, the numerical portion of the model. Ground motions are specified at base of finite element model. In these hybrid applications, the Shaking Table movement can represent the top surface of a foundation (for soil-structure interaction studies), bridge (for vehicle-structure interaction studies), or building systems (i.e., for investigating the effects of reactive mass or sloshing dampers on structure response). This approach permits study of specimens that are too large in their entirety to be tested experimentally on a shaking table. The approach is also an efficient means of performing parameter studies where the properties of the analytical model supporting the table are changed to represent different properties of the supporting soil or structure.

Hybrid Testing Overcomes Size Limitations of Test Specimens Hybrid Shake Table tests provide means to dynamically test subassemblies of large systems in full-scale or near full-scale that could otherwise not be tested on a shake table due to size, weight or strength limitations imposed by the simulator platform. In general, it is beneficial to perform hybrid shake table tests instead of traditional shake table tests whenever the dynamics of the test specimen significantly affects the response of the supporting structure or soil and, therefore, alters the required input to the shake table as testing progresses.

One example application where it is advantageous to employ the Hybrid Shake Table test method is for the experimental testing of tuned mass damper (TMD) systems on tall buildings. On a traditional shake table it would be necessary to test the entire building including the TMD to correctly capture the dynamic interaction of the TMD

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with the building response. Employing the Hybrid Shake Table The Shaking Table is used by researchers, industry, and design method, it is possible to model the tall building numerically in a finite professionals worldwide. For more information about the Shaking element program and only test the TMD specimen experimentally. Table capabilities, conducting research using the Hybrid shaking tables are a subset of real time hybrid simulation. table, or testing for seismic qualifications, visit the Implementation requires very precise control of the shaking table as PEER website at http://peer.berkeley.edu or contact well as the capability to carry out the necessary finite element analyses peer_center@berkeley.edu. in real time. To accomplish this, the PEER-developed open-source FRamework for Experimental Setup and COntrol (OpenFRESCO) Grace S. Kang, S.E., has been in private practice for over 25 and the open-source System for Earthquake Engineering Simulation years specializing in structural seismic design, and is Director (OpenSEES) have been installed in a set of multi-core high performance of Communications at PEER. She can be reached at computers and linked to the shaking table’s controller by means of a real g.kang@berkeley.edu. time shared memory network. During a hybrid simulation, OpenSees is All graphics courtesy of NISEE-PEER. used to compute the absolute displacements (and/or accelerations and/ or velocities) at the location of the shaking table platform in the numerical model for each step in the response history analysis, and these results are used to control the ® shake table platform in real time. At the same time, forces at the base of the test specimen are measured in real time and fed back into the numerical model at the virtual interface between the numerical model and Nearly two decades ago Hardy Frames created the first steel shear wall Shaking Table platform. This means that the system and revolutionized the residential building industry. Today the name numerical analysis needs to be performed in Hardy Frame¨ remains the most trusted name in seismic and wind solutions. real time to drive the shake table platform on the fly and produce accurate displacement, velocity and acceleration histories at Now as part of MiTek’s complete range of structural products, the Hardy the interface between the numerical model Frame® Shear Wall System, along with USP Structural Connectors¨ and and the experimental specimen. the Z4 Tie-Down System, can offer you even stronger integrated solutions.

The advantage of this approach is that the dynamic behavior of both the numerically modeled structure and the experimental Contact us today at hardyframe.com/solutions or 800.754.3030 and let us specimen are being captured correctly. create the right solution for you.

Furthermore, only the TMD specimen needs to be tested on the shake table instead of the entire tall building. Specimens can therefore be tested at full- or large-scale, and much more economically. If the test specimen on the shake table is nondestructive, such as a response modification device (TMD, seismic isolation, or damper), the Hardy Frame® Hybrid Shake Table test method also allows Special Moment for performing a wide range of building Frame parameter studies by simply changing the finite element model of the numerical porHardy Frame® tion of the structure before conducting HFX Panel another test. See Figure 4 for the testing of a TMD. Hence, the PEER-UC Berkeley Shaking Table, with its newly developed advance hybrid simulation capabilities, provides a unique opportunity to test large and complex systems, where test specimens dynamically interact with supporting structures, foundations or soils that would not be possible otherwise. Future upgrades include the ability to incorporate auxiliary actuators connected to different points within a test specimen or between the specimen and fixed points located on or ©2015 MiTek, All Rights Reserved off of the platform.

HARDY FRAME IS NOW AN EVEN STRONGER CHOICE.

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STRUCTURE magazine

Historic structures significant structures of the past

B

ridge historians and early textbooks generally call a truss with alternating compression and tension diagonals a Warren; however, sometimes it is called an equilateral truss since all panel lengths and diagonals are of equal length creating a series of equilateral triangles. When the panel lengths are shorter than the equal length diagonals, it was sometimes called an isosceles or isometric truss.

Figure 1. Commonly accepted Warren Truss.

The Warren Truss

By Frank Griggs, Jr., Dist. M. ASCE, D. Eng., P.E., P.L.S.

When the span length increases and the height of the truss necessarily increases, the long compression members in the top chord need bracing to minimize buckling in the vertical direction.

In this case, verticals are placed from the lower chord panel points up to the mid point of the chord member directly above. In addition, the deck structure stringers get longer requiring either heavier members or the addition of verticals from the top chord panel points dropping down to shorten panel lengths.

Figure 2. Warren Truss with verticals to support top chord and deck structure.

Dr. Griggs specializes in the restoration of historic bridges, having restored many 19 th Century cast and wrought iron bridges. He was formerly Director of Historic Bridge Programs for Clough, Harbour & Associates LLP in Albany, NY, and is now an independent Consulting Engineer.

Dr. Griggs can be reached at fgriggs@nycap.rr.com.

Neither of these truss styles are what James Warren and Willoughby Monzani patented in 1848 in England. They based their patent on similar trusses that were built in France by Alfred H. Neville and a patent that was granted in England to William Nash in 1839 on a similar design.

Warren and Monzani were well known English

engineers, and their design was for a truss that could be used as a deck or a through truss. They used cast iron for the top chord, and diagonals and wrought iron bars and links for the lower chord members. The top chord cast iron members were connected through cast iron junction blocks, and the cast iron diagonals and lower chord wrought iron members were connected with pins. The title of the patent application was Construction of Bridges and Aqueducts and was issued on August 15, 1848 with Patent #12,242. Their profile was rectangular. Even though Squire Whipple in the United States had published the method of determining loads in truss members under uniform and varying loads, this method had not made its way to England. It wasn’t until 1850 that W. B. Blood developed a method of analyzing triangular trusses, as Whipple had.

Warren and Monzani’s patent stated, The specification of this invention exhibits four different modes of building bridges, which it is stated may, with some slight modifications, be applied to the construction of aqueducts and roofing.

1) The bridge is built with cast iron side bands, rods, or plates, inclined towards each other, and combined so as to form a series of Vandykes [V shapes]. They are bolted at top to horizontal compression rods, and at bottom to horizontal tension rods, and carry a roadway at top or at bottom, or at both.

2) Or the bridge may be built of cast iron side angular frames (placed with the apices [point] downwards), which have their bases bolted together, end to end, and their apices bolted to horizontal rods.

3) Or, instead of the preceding modes of longitudinal construction, hollow castiron transverse frames may be employed, which are inclined, and bolted together at top, and are similarly attached at bottom to horizontal rods, bars, or plates.

Figure 3. Warren and Monzani patent drawing showing deck at both levels.

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4) Or wrought iron tie rods may be bolted at top to compression rods, and at bottom held together by the side of wooden girders, and the structure strengthened by means of stay rods.

The angles of the plates are regulated by longitudinal screw rods and nuts.

It is clear they didn’t size their members or give details as to the load, either tension or compression in their diagonals. He didn’t even think of his web members as triangles but only connected VanDykes (V’s) between a compression member on top and a tension member on the bottom. They had four claims as follows, 1) The mode of constructing bridges, aqueducts, or roofing with iron rods, bars, or plates, inclined towards each other, and connected together at top by compression band, and at bottom by tension band, so as to carry a roadway at top or bottom, or at both.

2) The mode of constructing bridges with cast iron angular frames bolted together at their bases, and having their apices bolted to horizontal compression rods.

3) The mode of constructing bridges with transverse hollow cast iron

Figure 4. Newark Dyke Bridge, cast iron A-frame on pier.

frames inclined towards each other, and bolted together at top and at bottom to horizontal plates.

4) The mode of constructing bridges with wrought iron rods inclined towards each other, and attached at top and bottom, as described.

It appears their only claim to originality was in the use of triangles with top compression chords and lower tension chords. The first major bridge, built by Joseph Cubitt in 1852 roughly to the patent, was the Newark Dyke Railroad Bridge of the Great Northern

Railroad. In it he used alternating cast iron compression and tension diagonals with cast iron upper chords and wrought iron links for the lower chord. At the middle panel he had opposing cast iron members.

The bridge crossed the Dyke on a sharp angle, requiring a span of 240 feet 6 inches.

Cubitt said the Warren design was brought to him by C. H. Wild. He wrote, Each girder consists of a top tube, or strut of cast-iron, and a bottom tie of wrought-iron links, connected together by alternate diagonal struts and ties of cast and wrought iron respectively, dividing the whole length into a series of equilateral triangles, of 18 feet 6 inches length of side.

These girders rest on the apices of castiron A-frames, placed on the masonry of the abutments (Figure 4). Each pair is connected by a horizontal bracing at the top and the bottom, leaving a clear width of 13 feet for the passage of the trains…

The trusses are so arranged, that all compressive strains are taken by the cast-iron, and all tensile strains by the wrought iron; the strains, in all cases, in the direction of the length are of the respective parts, and all cross strain is avoided. The parts are so proportioned, that when loaded with

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combination. But they are merely trusses with parallel chords and diagonals, or rather, oblique members, with only one series of obliques, and without verticals, except to concentrate weight upon the obliques from intermediate points along the upper or lower chord, according as the girder is loaded at such upper or lower chord.

Whipple did not think there was anything new with what was being called a Warren Truss. In fact, in his 1846/47 book he wrote of trusses without verticals. He called this a “cancelled truss which dispenses with vertical pieces, except perhaps at the ends, or at the first bearing points from the ends.” He found, in fact, that a truss, his trapezoidal without verticals, used 8% less iron.

Figure 5. End View of one of the two parallel Newark Dyke Spans. Note massiveness of cast iron members as well as the verticals to support the deck at mid panel points.

Figure 8. Whipple Plan 1846 but bridge historians call it a Double Warren Truss.

a weight equal to one ton per foot run, which considerably exceeds the weight of a train of the heaviest locomotive engines in use on the Great Northern, or on any narrow-gauge line, no tensile, or compressive strain on any part, exceeds five tons per square inch of section.

It is clear that by 1852 Wild had taken the Warren configuration and, applying Blood’s method of analysis, calculated the load in each member so that it could be proportioned properly. An end view of the bridge showed, however, the hugeness of the members which was typical of English and European bridge design at the time. Over time, the bridge style was converted to all wrought iron with built up riveted members.

In the United States, the Warren/Wild/Cubit design was known by our engineers. Many of them subscribed to the Proceedings of the Institution of Civil Engineers where Cubitt had published his article. Prior to 1848, Whipple had designed and built similar trusses on the New York and Erie Railroad and discussed them in his 1846/47 book on bridges. He included the plan shown in Figure 6.

Not only did he design this span, he built several for the New York and Erie Railroad

Several trusses that were patented in the United States incorporated the alternating tension and compression diagonals associated with the Warren Truss. The first was a wood and iron rectangular truss by A. D. Briggs in 1858 (#20,987) followed by Alber Fink in 1867 (#62,714) with a combination wood and iron trapezoidal truss with equilateral triangles with verticals dropping down to support the deck at mid panel points. He

Figure 6. Whipple plan for a bridge similar to the Warren Plan with inclined end posts.

in 1848, the same year Warren got his patent in England.

In an article in Appleton’s Magazine and Engineers Journal in January, 1851, he described some of his New York and Erie bridges and wrote, These were wrought-iron skeleton girders upon the triangular plan, such as have since been called Warren girders, and by some regarded as a newly invented

Figure 7. Whipple’s Brandywine Creek Bridge, New York and Erie Railroad, 1848.

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wrote, “I adopt the triangular, system of bracing between the two chords, both because this system best avoids evils arising from the unequal expansion of a wrought-iron bottom and wooden top chord, and because it is the system of bracing having the least amount of material for equal strength with other systems.” In the same year, J. Dutton Steele (#63,666) received a patent for an Isometric Truss. He had been building them since 1863 and called it an isometric plan, as the diagonals were of equal length with a shorter panel length. He had Charles Macdonald write a long report comparing all of the standard bridge designs, including the Pratt, Howe, Whipple and Warren trusses. Macdonald concluded the only cost savings in a truss bridge are in the web members, as the top and bottom chord requirements were the same for most bridges. For a standard bridge span length of 165 feet, he determined the Howe trusses needs 54% more iron in the web and the Pratt needs 31% more iron than the Isometrical truss. He then compares the Isometrical truss with Linville’s double intersection truss and determines the isometrical uses 19% less iron in the web. He presents the results of a study by C. Shaler Smith in 1865 where he compared the Fink, Bollman, Triangular (Warren) and Murphy Trusses.

Figure 9. J. Dutton Steele patent drawing for an isometric plan.

Smith determined the Triangular and Murphy were more efficient than either the Fink or Bollman trusses for both through and deck trusses. His conclusion stated the Isometrical Truss required less iron in the web system than any other trusses considered. In addition, he found the Isometrical Truss was, especially in wood, much easier to adjust in the event of wood shrinkage.

In 1872, Whipple, in an article in the Transactions ASCE entitled “On Truss Bridge Building,” wrote that he had objections to Macdonald’s pamphlet and how he used the Whipple Double Intersection truss in his comparison, stating: “Now, Mr. Macdonald represents what he designates as the ‘Whipple Truss,’ with diagonals inclining only 30° from the vertical. I desire here

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Figure 10. Little Juniata Bridge, Pennsylvania RR, cast and wrought iron with verticals, Pony Truss ~1870.

to enter my emphatic protest against the imputation of ever having tolerated any such practice.” He then got into the Isometric Truss (and the Warren style), writing: But what of the Isometric? The name, at least, as applied to bridge trusses, is new, and euphonic [pleasing to the ear] withal.

This is a truss with parallel chords without vertical members in the web: one of the general types discussed and compared in my publication of 1847 with reference to Fig. A., page 14…

I am not aware that there had existed any examples of the parallel chord truss without verticals, prior to their construction by me over 20 years ago, with the important exception of the plank lattice bridge. This was first known to me under the name of “Town’s Lattice Bridge,’” and it was a very cheap and serviceable bridge when properly constructed…

But somehow it occurred to me…that a plan in which every member of the web system should do something in the way of advancing the weight toward the abutments, might possess advantages over one having vertical members merely to transfer the action of weight directly from chord to chord without advancing it at all horizontally…

The Trapezoidal truss, with and without verticals, though depending upon combinations so old that “the memory of man” (especially the present generation) “runneth not to the contrary” still, perhaps, owes something to me for economical form and proportions…

These gentlemen [Macdonald and Merrill] are pleased to term ‘The Whipple Truss;’ and considering that the Isometric and the Post [with inclined posts] trusses are merely modifications (and not very

Figure 11. Bell’s Bridge, Delaware, Lackawanna & Western RR 1872, Double Warren or Whipple.

favorable modifications either) of a type of truss first used and thoroughly discussed by me.

It is clear that Whipple believed that the Warren or Isometric trusses were simply extensions of trusses he wrote about in the 1840s, and built in the 1840s and 1850s. In an article on the Pratt Truss (STRUCTURE, May 2015), a case was made that the trusses called Howe and Pratt should really be called Whipple Trusses. A similar case is made here that the Warren Truss should really be called a Whipple Truss. The reasons are that Warren, when he developed his truss, did not know how to size his members nor could he distinguish between tension and compression in his web members. He never designed or built a truss with an inclined end post nor a truss with verticals. A truss as he patented it was never built. Whipple on the other hand had analyzed, designed and built trusses with various web members and inclined end posts prior to Warren’s patent.

It is probably too late to change what most people call the various trusses, but it should be at least recognized that most of the truss patterns used in the late 19th and 20th century had their origins in the United States and Squire Whipple between 1841 and the 1880s. What were called the Warren trusses were built in the thousands as short span pony trusses with no verticals, longer spans with verticals, even longer spans with double intersections, and still longer spans with subdivided panels. They were originally built with cast and wrought iron members with pins and, later, with wrought iron members and cast iron joints with pins, and later fully riveted in steel. Polygonal top chords were also added in many trusses to extend the span length. J. A. L. Waddell used the pattern in many of his lift spans after the turn of the century. Several examples of the bridge style are shown in Figures 10, 11 and 12.▪

Figure 12. Warren, Isometric, Truss, Polygonal top chord, with verticals, all riveted steel bridge for BNSF Railroad over Verdigris River, Oklahoma~1960.

STRUCTURE magazine

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SEISMIC CONSTRUCTION STRONG New Products, Innovations on Tap

By Larry Kahaner

T

he earthquakes that hit Nepal in April and May have once again focused the construction industry and world leaders on the devastating loss of life and property involved in seismic events. More than 8,600 people died in the two earthquakes that destroyed more than a half-million homes, leaving many residents exposed to the upcoming monsoon rains.

Many governments are responding to seismic threats. “The market for energy absorption technologies in seismic applications is expanding as global government requirements are changing to protect critical assets, economic loss and loss of life,” says Greg Herman, Sr. Project Manager at ITT Enidine Inc. (www.enidine.com) in Orchard Park, New York. “Today, new builds and structural retrofits must consider energy absorption devices to mitigate these issues. Seismic protection is no longer an afterthought. Energy absorption technologies are used on an increasing basis to alleviate wind and seismic concerns.”

In discussing the company’s products, he notes: “Fluid Viscous Dampers (FVD’s) are devices that operate by converting kinetic energy into heat, typically over multiple actuations. FVD’s are velocity dependent devices; they do not provide damping force or stiffness to a structure unless subjected to external excitation. FVD technologies use the flow of internal fluids and custom orifice geometry to obtain linear and non-linear damping properties. While applying FVD’s as energy dissipation devices, a structure will increase its critical damping ratio and reduce the dynamic magnification factor, thereby reducing the dynamic response of the structural system in the event of an earthquake.”

Herman adds: “There are different designs and technologies for energy absorption as it relates to seismic damping applications. Selecting the appropriate technology is critical to providing the proper protection of the structure. These selections are based on the application, as there are different requirements for bridge, building (in frame vs. base) and tuned mass damping systems.”

He highlights these differences: • Bridges and base isolation components are often exposed to wind, thermal and traffic excitation on a daily basis. FVD’s for these applications must be able to withstand harsh conditions and sustain a long lifespan.

• For buildings, the excitation frequency is significantly less than for bridges. FVD’s for these applications should be designed to endure long periods of static installation with features that promote extended life span.

• For tuned mass damping systems, a linear damper is often recommended, as these applications require high energy dissipation in tight spaces and require a relatively high amount of seal travel over the design life.

When it comes to cost, Herman says that FVD’s can be a price-effective solution. “In reviewing the total cost of ownership, adding FVD’s for energy dissipation can often be a more cost effective solution, especially when coupled with a supplier that represents reduced risk for any associated elements and integration into a project. Because FVD’s STRUCTURE magazine

have the ability to reduce the stress and displacement of a structure, they require less material, reducing the customer’s total cost.”

At Simpson Strong-Tie (www.strongtie.com) based in Pleasanton, California, Marketing Communications Project Manager Elizabeth Rajs is seeing an increasing demand for light-frame, multi-story wood frame buildings. “There are complicated design challenges specifically associated with multi-story buildings that must withstand seismic activity or wind events. Our anchor Tiedown System for Shearwall Overturning Restraint and our Uplift Restraint System for Roofs address those challenges, and have become a popular continuous rod tiedown solution for these types of construction projects. Simpson Strong-Tie has everything engineers need to design the safest building possible with materials specifically suited for the application.”

The company has several new offerings, Rajs explains: • The Simpson Strong-Tie Strong-Rod Systems Design Guide (F-L-SRS15 ) is now available in print or for download at www.strongtie.com/srs. This new document guides designers through the specific challenges and solutions available to address multi-story light frame construction shear wall overturning restraint as well as roof wind uplift restraint using Simpson Strong-Tie Strong-Rod Systems. Three methods of specification are explained.

• The Simpson Strong-Tie RTUD5 and RTUD6 ratcheting take-up devices are new additions to their cost-effective line of products for ⅝-inch and ¾-inch diameter threaded rod, respectively. Once installed on top of the BPRTUD5-6 bearing plate, a series of internal threaded wedges enable the device to ratchet down the rod as the wood structure shrinks but engage the rod in the reverse direction when under tensile loading, ensuring continuous engagement is maintained enabling the rod system to perform as designed. Unlike similar products, the RTUD can be attached using nails or screws.

• Simpson Strong-Tie has thoroughly researched and tested practical solutions for concrete podium slab anchorage to provide designers with additional options when designing light-frame structures over concrete podium slabs. The use of the special detailing of anchor reinforcement shown in ACI 318, Anchorage Provisions, will allow the full tensile capacities of the anchors to be developed in shallow concrete podium slabs. For assistance, designers can visit www.strongtie.com/srs for suggested anchorage-to-podium slab details, slab design requirements and Shallow Podium Slab Anchor Kit product information.

• For other foundation anchorage conditions, design tools, such as the Simpson Strong-Tie Anchor Designer Software, are available to help designers navigate the complex anchorage provisions contained in the ACI 318 reference design standard. Anchor products, including the Pre-Assembled Anchor Bolt (PAB), are also available to simplify specification.

Rajs concludes: “Because no two buildings are alike, each project is optimally designed to the designer’s individual specifications. continued on next page

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July 2015

Run-assembly elevation drawings and load tables are provided to the designer for approval.” (See ad on page 35.)

“Although Taylor Devices (www.taylordevices.com)has been designing and manufacturing special dampers and motion control products for approximately 60 years, these types of devices have only been used in structures beginning in the mid 1990s,” says Alan Klembczyk, the North Tonawanda, New York company’s Vice President, Sales & Engineering.”Since that time, we have adapted more and more special spring and damping devices into structures that were previously qualified for classified military and aerospace applications.”

Klembczyk says that Taylor Devices has more than 600 structural projects worldwide with these special devices, some of which include a new combination of damping and spring components in either a

parallel or series arrangement. “Typical applications include new tall buildings, new medium-sized buildings, retrofits, large bridges, pedestrian bridges and various other structures. The great feature of these Taylor damping products is that due to its output proportional to input velocity, it is the only structural product in use today that functions out-of-phase with the structural response (allowing reduction of stress) and acceleration, along with the reduction of deflection that is provided by all energy dissipation products.”

He says that since the company has developed specialized components with non-linear behavior, structural engineers can now engage the added benefits that an optimized product can offer. “Our devices can incorporate linear or non-linear damping exponents, linear or non-linear spring components, pre-loaded mechanisms for zero drift during small inputs, high cycle-life components with no fatigue failures over decades of use, friction-free devices, force limiting devices, shock transmission units YOU BUILD IT. and any combination of these attributes.

WE’LL PROTECT IT. Our ‘bag of tricks’ in solving structural and dynamic problems has grown for more than six decades,” says Klembczyk.

As for topical projects, Klembczyk notes that the company has recently delivered Stand firm. Don’t settle for less than the seismic protection more than 100 non-linear dampers for of Taylor Fluid Viscous Dampers. As a world leader in the new San Diego Central Courthouse the science of shock isolation, we are the team you Project. These dampers were modeled and want between your structure and the undeniable forces analyzed by Skidmore, Owings & Merrill of nature. Others agree. Taylor Fluid Viscous Dampers LLP (SOM) as having two distinct Force vs. Velocity regions; one being linear for are currently providing earthquake, wind, and motion low velocities experienced during wind protection on more than 650 buildings and bridges. events and one being non-linear for the From the historic Los Angeles City Hall to Mexico’s higher velocity range that occurs during Torre Mayor and the new Shin-Yokohama High-speed seismic events. Wind tunnel testing at Train Station in Japan, owners, architects, engineers, RWDI in Guelph, Ontario, provided and contractors trust the proven the site specific wind environment and technology of Taylor Devices’ analysis at SOM verified performance Fluid Viscous Dampers. during both these wind and seismic inputs. “With this additional level of analysis performed, our devices enabled an optimized solution for both wind and seismic events,” he says.

“Business has been very good,” Klembczyk adds. “Our sales and product lines are continuing to grow as more engineers are willing to take a close look at what benefits our products can offer.

A relatively simple structural analysis can demonstrate the benefits of adding our devices to their structures. Oftentimes, it is a simple matter of taking Taylor Devices’ Fluid Viscous Dampers give you the seismic protection that first step for engineers you need and the architectural freedom you want. to realize what they may not have considered before.”▪

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SEISMIC PROTECTION FROM TAYLOR DEVICES

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Risk ManageMent

risk management topics for structural engineers

Myth of Probabilistic Seismic Hazard Analysis

By Praveen K. Malhotra, Ph.D., P.E. Table 1 lists the mean µ, standard deviation σ, and coefficient of variation COV of 50-year maximum PGA based on the PSHA by the USGS in 1996, 2002, 2008 and 2014. Note that the uncertainty (COV) in ground motion prediction increased every time the PSHA was repeated. This suggests that the USGS was less certain of its predictions in 2014 than in 1996. There are three possible explanations for this surprising observation: • As computers became faster and more powerful, it became possible for the USGS to consider more sources of uncertainty in the PSHA. This is somewhat ironic because computer resources should be used to decrease rather than increase uncertainty.

• The number of opinions regarding the geological and seismological models increased over the years due to greater number of researchers investigating the seismic hazard. The uncertainty increases when more and more opinions are incorporated into the PSHA. • Seismic hazard analysis is in its early stages and geoscientists are still discovering more and more that they do not know.

Whatever the causes of rising uncertainty, this trend should be reversed to show progress in the science of ground motion prediction. Of course, uncertainty should not be hidden; it should only be reduced through a scientific process. Note in Table 1 that the mean value of maximum 50-year PGA has reduced since 1996. But is this reduction significant enough to warrant a change in design ground motions?

6 The answer to that question is provided by the effect-size, defined as a change 4 in the mean value as Probability Density

T

he purpose of science is to explain and predict. Uncertainty in scientific predictions degrades their value. Uncertainty should not be hidden, but it should be reduced over time to show progress in science.

Predictions of future ground motions are needed for mitigation and insurance of seismic risk. The United States Geological Survey (USGS) is responsible for predicting future ground motions throughout the U.S. and territories. The USGS performs a probabilistic seismic hazard analysis (PSHA) by combining the uncertainties in time, location and size of earthquakes (geological models) with those in the ground motions for specific earthquakes (seismological models). The PSHA is performed out of necessity – rather than choice – because uncertainties in geological and seismological models are too high to be ignored at present time. Geological and seismological models are regularly updated and the PSHA repeated by the USGS. For earthquake insurance, annual estimates of ground motions are needed because insurance policies are renewed every year. For seismic design, ground motion estimates are needed for the remaining life (or ownership) of the structure, say 50 years. During a 50-year time span, a site can be shaken by different earthquakes. The maximum ground chaking in 50 years is uncertain. Figure 1 shows the probability density function of 50-year maximum horizontal ground acceleration (PGA) for a ‘firm-rock’ site in Los Angeles, based on the 2014 PSHA by the USGS. According to Figure 1, any acceleration is possible in Los Angeles during the next 50 years. The expected (mean) value of 50-year maximum PGA is µ = 0.208 g (g = 9.81 m/s2) and its standard deviation is σ = 0.21g. The coefficient of variation (a common measure of uncertainty) is COV = 0.21/0.208 = 1.01.

Table 1. Mean µ, standard deviation σ, and coefficient of variation COV of 50-year maximum PGA for Los Angeles based on the PSHA by the USGS in 1996, 2002, 2008 and 2014.

Year

µ (g)

σ (g)

COV

1996

0.25

0.188

0.75

2002

0.258

0.2

0.78

2008

0.234

0.221

0.95

2014

0.208

0.21

1.01

STRUCTURE magazine

a fraction of the standard deviation. Table 2 lists the effect-size of change in ground motion predictions by the USGS between successive PSHA. The effect-size of 0.2 to 0.3 is considered small, ~0.5 is considered medium, and > 0.8 is considered large. Therefore, the effect-size of change in ground motion predictions by the USGS is too small to justify a change in design ground motions. In other words, the uncertainty in ground motion predictions is so high that a slight reduction in the ground motion estimates from 1996 to 2014 cannot be taken seriously.

In Los Angeles, the design acceleration for a building is 0.64 g according to ASCE 7-10, whereas the expected maximum acceleration in any given year is only 0.0129 g according to the USGS. The design acceleration is 50 times the expected acceleration in any given year, and still there is 1/5000 chance per year of collapse of a code-designed building and much higher chance of property damage. In Boston, where the uncertainty is higher than Los Angeles, the design acceleration is 75 times the expected maximum acceleration in any given year. High ratio between the design and expected maximum accelerations makes risk mitigation less attractive compared to insurance. But, insurance without mitigation is costly. Also, insurance cannot reverse the loss of life, loss of business, or the damage to the environment. Due to high uncertainty, voluntary mitigation gets pushed from one year to the next. There are significant social and economic reasons to reduce uncertainty in ground motion predictions, and yet there is no active discussion on reducing uncertainty.

2

0

0

0.2

0.4

0.6

0.8

Peak Horizontal Ground Acceleration, PGA (g)

1

Figure 1. Probability density function of 50-year maximum PGA for a ‘firmrock’ site in Los Angeles, CA, based on the 2014 PSHA by the USGS.

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Table 2. Effect-size of change in estimates of 50-year maximum PGA in Los Angeles between successive PSHA by the USGS.

Interval

Effect-Size of Change

1996-2002

0.04

2002-2008

0.11

2008-2014

0.12

Some suggestions for more meaningful application of PSHA are as follows: • The PSHA should be considered as a tool for incorporating uncertainty; it should not be considered as an excuse for using models with high uncertainty. Models with high uncertainty increase the costs of mitigating and insuring the risk.

• The focus in seismic hazard analysis should shift from considering uncertainty to reducing uncertainty. In the seismic hazard community, there are more discussions on considering uncertainty, and practically no discussions on reducing uncertainty.

This needs to change. The real progress in the science of ground motion prediction can only be made by reducing uncertainty.

• Uncertainty in geological and seismological models should be reduced by gradually replacing empirical models with physics-based models, and by limiting the number of opinions considered in the PSHA. Empirical models, by their very nature, are uncertain and there are several opinions regarding empirical models. For example, at present time, there are about 300 ground motion prediction (attenuation) models in the published literature.

• The design ground motions should not be revised unless a statistically significant change occurs in the predicted ground motions. The uncertainty in ground motion prediction is so high that a slight change in ground motion estimate does not justify a change in the design ground motions.

• Site-specific PSHA should not be used for the sole purpose of reducing design ground motions. Since there are many opinions regarding geological and seismological models, a combination of opinions can be easily found to obtain a lower estimate of the design ground motion. This practice should be stopped, because it provides a biased assessment of the risk.

• Same predictions of ground motions should be used for seismic design and insurance of all types of structures.

There is no convincing argument for separate PSHA for buildings, bridges, dams and nuclear power plants. Design ground motions should be different for different types of structures, based on the consequence of damage; but the underlying assessment of the hazard should be the same.

• Seismic hazard models should not be made more complex unless they reduce uncertainty. It does not make sense for ground motion prediction models to become more complex and less certain at the same time. Complexity without certainty is self-serving.▪ Praveen K. Malhotra, Ph.D., P.E., is a Principal at StrongMotions Inc. in the Boston Area. He can be reached at praveen.malhotra@strongmotions.com.

The online version of this article has detailed references. Please visit www.STRUCTUREmag.org.

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Ground Improvement for Building Support What is it and Who is Responsible?

By Damian R. Siebert, P.E. and Steven R. Kraemer, P.E.

M

any types of ground improvement (GI) are used in modern geotechnical construction. For the purpose of this article, GI refers to a rigid or semi-rigid cylindrical inclusion (GI element) installed through otherwise unsuitable bearing soils into underlying suitable strata to enable the use of spread footing foundations for building support. Due to favorable economics compared to viable foundation alternatives, use of GI is accelerating in the United States.

Commonly, GI elements are formed of crushed stone, grouted crushed stone, grout or concrete. They are typically designed and installed by specialty subcontractors using a variety of analysis and design approaches, installation methods and equipment.

Historically, GI has been used to limit footing settlements to tolerable levels by transforming the unsuitable strata into a stiffer “composite” media consisting of lowcapacity GI elements and the “improved” soil. GI was not typically relied upon to increase bearing capacity significantly or provide the factor of safety against this mode of failure. Recently, in addition to settlement control, GI elements are now also being used in situations where they must provide most if not all of the margin of safety against a bearing failure.

As applications of GI are being pushed to new limits, the need for deep understanding of installation conditions, behavior and adequate quality assurance becomes more critical. The adequacy of GI in bearing capacity applications cannot be assumed simply because of its successful legacy for settlement control. In such challenging applications, the Engineer(s)-of-Record (EOR) and project team must ensure that the GI provides comparable bearing capacity, settlement control, resiliency during earthquake and other loading conditions, quality assurance and overall performance as other “conventional” foundation systems.

Experience is proving that this is easier said than done.

Due to the nature of GI elements and commonly used installation methods, quality assurance is more difficult than for other

types of foundation systems. Each step of the process – including the decision to use GI, specifications, GI type selection and design, reviews, element/system testing and on-site monitoring – is vital to success.

Adding further complication is the fact that there is confusion in the industry over what various element types are called or consist of, as they are sometimes referred to generically even though the elements and their installation methods can differ significantly.

Some are proprietary, making it difficult to know important details.

The already fuzzy lines of responsibility for GI among the Geotechnical Engineer, Structural Engineer, Architect – especially if the Structural Engineer is retained through the Architect – General Contractor and specialty GI subcontractor become further blurred. Adding to the complexity is the fact that most current building codes in the U.S. do not address use of GI; the authors are only aware of one where GI is mentioned: the Massachusetts State Building Code, and its use only requires notification to the Building Official. Typically, design – including element type, installation method, configuration, depth and capacity – of a GI system is performed by a registered Professional Engineer retained by the subcontractor using building loading information provided by the Structural Engineer and information on site subsurface conditions provided by the Owner’s Geotechnical Engineer. The project may have a GI specification as part of the construction contract documents, prepared by the Owner’s Geotechnical Engineer.

Sometimes, structural redesign of the footings may also result.

Each team member has contractual and (potentially) legal obligations in a design/ construction project. When GI is used, the Owner’s Geotechnical Engineer may or may not have recommended its use; GI is sometimes proposed by an Owner or contractor as a value engineering alternative. Details of the GI design are provided by the subcontractor’s designer through a P.E.-stamped submittal. The submittal is typically reviewed by the Geotechnical and Structural Engineers. The use of GI,

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the subcontractor’s GI design and reviews by the Owner’s Engineers impact and intertwine with the Engineer-of-Record (EOR) responsibility for the building foundations in potentially complex ways, which can depend on applicable codes, project specifications, contracts and actions by the involved parties.

Often, the Structural Engineer and Owner rely upon the Geotechnical Engineer to be sure that “everything is okay,” despite the fact that the Structural Engineer is the EOR for the structural aspects of the building including the foundations. By submitting a stamped design for GI, the specialty subcontractor’s designer takes primary responsibility for foundation performance, often contractually through the General Contractor. Because of his/ her unique understanding and role in the project, the Geotechnical Engineer also has a professional responsibility in the process, including advising the Client and the project team of the special issues and potential risks involved.

The very important technical issues and responsibilities associated with the use of GI on building foundation projects are too complex to be addressed in detail in this short article. The key point is that use of GI presents special challenges that are the result of their inherent complexities, and that are becoming more acute each day due to the very rapid evolution of the systems, their application, and the contractors installing them. It is important for project teams to understand the roles and responsibilities of each team member in the process. In addition to the potential benefits, projectspecific challenges and risks must also be identified, and provisions must be put in place to mitigate them.▪

Damian R. Siebert, P.E. (DSiebert@ haleyaldrich.com), is Lead Underground Engineer and Steven R. Kraemer, P.E. (SKraemer@haleyaldrich.com), is Senior Vice President at Haley & Aldrich, Inc. in Boston, Massachusetts.

Professional issues

issues affecting the structural engineering profession

Two Generations: Why Young Member Groups (YMG) are Important This is the second of a two part series that draws attention to Young Member Groups. The first article “Rainbows Only Come After Rain” by Ellen (Chuan-Hua) Kuo PE, LEED AP.BD+C was published in the February, 2015 issue of STRUCTURE.

Why are Young Member Groups Important?

A Young Members’ Perspective The extent to which industries train market entrants and have established recruiting procedures varies from industry to industry. It has been my experience that the U.S. structural engineering industry is one of the professional fields that offers the least amount of such a structured guidance for newcomers.

Some companies do defy this regular practice by providing excellent training/mentoring programs, but they are the exception rather than the rule.

Paradoxically, structural engineering is a highly technical career where a lot of the learning must take place on the job. This is because the structural behavior of the whole does not occur in a vacuum, rather it is greatly influenced by the unique characteristics of individual components. An engineer in training must learn to apply theoretical concepts and models to describe an observed behavior and estimate future behavior. Such training can only occur while solving real engineering problems. It is this unique feature of our industry that makes the development of standardized training very difficult; the development of an engineer occurs while solving case by case conditions.

This is why young professionals benefit greatly through an opportunity to consult more senior and experienced engineers and, when provided, the opportunity to bounce ideas and approaches with their young peers.

This is why YMG’s are so important: They provide young engineers with a platform to enhance their professional development while having a support network. In this network, young professionals are not afraid to fail, rather they want to learn from their mistakes and try again.

Young engineers enter an unusual environment where making mistakes carries significant liability. Building codes provide vetted principles to guide engineers in the analysis and design of elements unfamiliar to them. The codes are a compilation of the experiences and studies of professionals and academics intersecting with the needs of the contracting industry, the engineering

community and the public we serve. This makes the standards referenced by the codes (and the code itself) at times difficult to understand. YMGs offer young professionals the opportunity to discuss their interpretations and learn from each other’s experiences.

Additionally, YMG’s introduce young members to the importance of professional involvement in our field. Structural Engineers must become familiar with the liability of our practice, its interpretation by the law and how our practice can improve through active collaboration with local authorities. The building codes and standards are advanced by volunteer leaders of our industry. As the next generation of leaders, we must recognize that the more involved our industry becomes with general society, the more updated and relevant the codes we must comply with will become.

Involved young members can accelerate the incorporation of new technologies and find ways to make processes more agile in order to allow future engineers to participate more actively in the shaping of the plans for a project.

For me, working with Ellen Kuo in starting the SEAMASS YMG and becoming actively involved has been a mind expanding experience. It has been inspiring to understand how the ideas and concepts of a few have become standards and legislation; although to a lesser extent, my involvement has allowed me to impact positively on the local young member community. I see a lot of potential in the ability of other young members to further our profession by coordinating the young member perspectives and transforming them into action items for the improvement of the entire community. I believe general society can benefit from a more active and integrated participation of structural engineers, and I hope the YMG’s will become gateways to launch brilliant young structural engineers into active participation.

Sofia Zamora, E.I.T., is a Structural Designer at CBI Consulting Inc. Sofia co-founded the SEAMASS YMG and leads Professional Networking efforts at the NCSEA YMG Support Committee.

She may be reached at szamora@cbiconsultinginc.com.

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Making the Connection An Experienced Engineers’ Perspective Close to 60 years ago, a small group of structural engineers practicing in the Boston area started having diner meetings at Joseph’s Restaurant on Newbury Street in Boston to discuss business matters. In 1958, the group became the Boston Association of Structural Engineers, a firm based membership made up of business owners. The group started having regular business meetings at MIT, and many of its members became involved with the advisory committees to the Massachusetts State Building Code.

In 1996, BASE became a charter member organization of NCSEA. Two BASE members, Craig Barnes and Emil Troup have served as President of NCSEA in its early years.

Realizing that BASE membership allows for one business owner from each firm, in 2006 they created SEAMASS to be the member organization of NCSEA enabling all structural engineers in Massachusetts to gain access to opportunities to improve the practice of structural engineering and to foster proactive communication among structural engineers. In an effort to kick start SEAMASS, all BASE members agreed that, as part of their membership, they would be committed to also being SEAMASS members.

Instantly, SEAMASS had close to 60 senior level structural engineers as its membership. Unlike SEAoNY who started their organization with mostly young engineers, SEAMASS did not have a strong young member base. At a recent NCSEA conference, it became apparent that many MO’s were facing the same problem; of course, except for SEAoNY, who claimed that they would probably have to create a OMG (old members group).

As with many MO’s, there is the initial struggle with finding themselves and becoming established. In the midst of this, how would SEAMASS attract younger members? Not everything happens overnight. I think the first big step for SEAMASS was when directors started asking younger members from their firms to help volunteer at SEAMASS events. A few years ago at a past SEAMASS annual meeting, the young members showed up in force, many helping to run the event.

experienced engineers who became my mentors. Whenever I got stuck on something, I had my favorite mentor to reach out to. From the time I started out as a member of BASE, there were a handful of senior members who took me under their wings, in a way that mimicked the sincerity of one’s parent. Their interest gave me the opportunity to become involved and grow within my profession. By having a ‘place’ for young members to gather, they too will have the opportunity to share in the same experience.

Like professional sports teams with farm teams, we should have the same system in our profession where it becomes a natural progression from university student to practicing engineers.

One of the benefits of our YMG and their opportunity to network has been in giving both the employer and potential employee a place to meet. In my experience, I have been able to direct firms looking for new staff to potential YMG candidates and, on the other hand, have been able to recommend young members to the right firms. Going both ways, it helps when people know of each other, making the employment process less risky.

What a great place to share ideas, senior engineers with their experience and rules of thumb and young members fresh from the Universities with the latest technology. For me, I think it is exciting and the newer young members bring a lot of energy to our MO.

Every generation is different. The ability to network, newer and older, gives one the opportunity to better understand each other. Through the process of establishing a YMG, many new relationships have been formed. With our YMG, we have been able to reach out to each other both ways; I recently found myself reaching out to a young member to gain their perspective on an issue. In the end, we both benefitted.

What has been most rewarding for me is to see how our YMG has taken the ball and has run with it. They are a great group and have been a benefit to our MO. They check in with us once and a while, but in the end, it is their group and it is obvious they know how to run it. I recently received an invite for an event the YMG has put together. I was impressed!

Another interesting event organized solely by the YMG! Here in Massachusetts, our YMG is an asset and we appreciate them. If you haven’t started a young member group, having one would be well worth your while.▪ Michael Fillion, P.E., SECB, is the owner of Fillion Group, Inc., Structural Engineering.

He is a member of the Structural Advisory Committee to Massachusetts State Building Code and a member of the NCSEA Existing Building Committee. He may be reached at Mrf.structure@verizon.net.

Rendering courtesy of HEWITT

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The sense I got at that meeting was that these younger members were enthusiastic and really wanted to be part of the organization. This was an opportunity for the young members to network and become recognized.

Shortly after, while at the NCSEA conference in Atlanta, YMG chair Heather Anastasia from Florida won everyone’s hearts in her address at the delegate meeting. Heather shared her story as a little girl building tree houses in her yard to her experience of being in the work force as a young engineer. I sense most MO’s left Atlanta excited about the prospect of starting their own YMG. The networking really paid off; by the time we landed back in Boston, we had already speculated that Sofia Zamora & Ellen Kuo would be ideal candidates to start our YMG. Within a week of accepting their new roles as YMG co-chairs, Sofia & Ellen reached out to Heather who was very helpful in getting the Young Member Group (YMG) started. In short time, SEAMASS officially had started an energetic YMG. Although they had some guidance from senior members at the start, we encouraged them to be independent, giving them as much responsibility amongst themselves. I believe this enabled them to feel that they had a stake in the success of their MO. Members of the YMG regularly sit in on our monthly board meetings. The young members organize teach-ins, field trips and social networking events. Some of the YMG events are open to the general membership, giving younger and more senior level engineers a chance to net-work.

The SEAMASS hub is Boston, where there are over 100 universities. Of these, there are a handful of universities that offer structural engineering curricula. The YMG has been connecting with the local universities to be a resource to their students. Having strong ties with Northeastern University, the YMG has established an annual distinguished lecture during Engineers week. At the event, engineering students and staff from the local universities, along with practicing structural engineers, meet to socialize and attend the distinguished lecture. This past year, Ben Schafer lectured on The Birth of the Skyscraper preceded by a SEAMASS presentation on snow loads and the performance of buildings in Massachusetts, 2015. The idea was to expose students to what practicing structural engineers encounter in the work place. The YMG is also working on a mentorship program which they hope will continue to grow.

Recently, SEAMASS members were able to sit in on students’ thesis projects, giving guidance and consultation.

I can remember when I started out as a young engineer and am grateful to the

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Product Watch

updates on emerging technologies, products and services

Fluid Viscous Damping for Seismic Energy Protection

By Shubin Ruan, Ph.D., P.E. and Ben Eder

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he Pacific Rim, or The Ring of Fire, is commonly known as the area bordering the Pacific Ocean that attracts a high volume of seismic and volcanic activity from shifting plate tectonics. Over the past 100 years, 90 percent of earthquakes and 75 percent of volcanic eruptions have occurred in the Pac Rim, making it one of the most active and destructive areas in the world. With the recent influx of high-magnitude earthquakes over the past few years, we are reminded of the importance of seismic protection and its value in safeguarding some of our world’s most precious structural assets.

In creating a new structure during the design phase, it’s important to account for all possible causes of structural collapse. In order to adequately protect against seismic activity, energy dissipation devices such as fluid viscous dampers (FVDs) are often applied to mitigate structural sway in bridges and buildings.

When properly specified and designed, fluid viscous dampers serve as a structure’s primary defense during a seismic event – preventing catastrophic damage and costly repairs.

Purpose and Practice FVDs are often used as an energy dissipation device on structures built in seismic hotspots.

A structure that utilizes fluid viscous dampers will increase its critical damping ratio and reduce the dynamic magnification factor. This reduction will reduce the dynamic response of a structural system in the event of an earthquake or other external excitation.

A fluid viscous damper is a device whose resultant force is proportional to its relative velocity (linear FVD) or some power of velocity (non-linear FVD). FVDs operate by converting kinetic energy into heat, typically over multiple cycles (tension and compression). This technology utilizes silicone fluid and custom orifice geometry to obtain linear (α = 1.0) and non-linear (α ≠ 1.0) fluid viscous damping performance. FVD performance is characterized by the Constitutive Law (Figure 1). Because of the inherent unpredictability of the magnitudes of earthquakes, engineers may want to consider a lower velocity exponent (α < 0.5). As the damping alpha decreases, the damping efficiency increases. For example, a FVD with an α = 1.0 has a damping efficiency

Figure 2. Critical applications require high-performance damping devices.

of 78.5%, while a FVD with an α = 0.1 has a damping efficiency of 97.1%.

Most importantly, a low damping alpha provides more consistent damping force over a larger range of input velocities (Figure 2).

For example, a FVD with α = 1.0 would be providing 2.000(X) the design damping force if the actual velocity input is 2.0(X) the design velocity. But, for a FVD with α = 0.1 if the actual velocity input is 2.0X the design velocity the FVD is still providing nearly the design damping force (actual 1.072X design force).

Application Use When considering seismic support options for a building or bridge, it’s important to understand the advantages of using FVD devices over other types of energy dissipaters.

Constitutive Law: F = CV α F: Damping Force [kN] V: Relative Velocity [m/sec] C: Damping Coefficient [kN/(m/sec)α] α: Velocity Exponent Figure 1. Constitutive Law chart.

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First and foremost, FVDs are applied to add damping to a structure, reducing structural response during periods of seismic excitation.

From a lateral perspective, FVDs can reduce displacement and story drift, which lessens the stress placed on structural members. FVDs have the ability to reduce floor accelerations and, therefore, the inter-story shear and base shear/overturning moment.

In addition, FVDs are typically easy to install, operate and monitor in comparison to other devices. Design features such as internal fluid reservoirs and innovative sealing technologies extend damper design life.

Monitoring options such as fluid indicators and cycle counting are part of a comprehensive health-assessment program that can help extend damper life and reduce risk of leaving a structure unprotected without the required damping in the case of a seismic event.

Finally, adding FVDs to a structure can often be a cost-effective solution for a long-term preventive maintenance plan.

When considering the costs associated with structural damage, the initial purchase price is only a small portion of the damper’s total value, reducing the customer’s total installed costs.

CONSTRUCTION CEMENT

FA S T ER STRONGER MORE DURABLE 3000 PSI IN 1 HOUR Figure 3. Customizing your FVD device to specific application requirements will extend the life of your entire structure.

There are several areas to consider when selecting an FVD device for your unique application. To ensure proper installation and performance, design engineers must take the application requirements into account before choosing a device.

In some applications, specific requirements may warrant a combination of devices in order to achieve optimal protection. It’s important to remember that there is a wide array of technologies that are available for any application, and the goal is truly to find the most effective solution. For proper performance and desired response, specification of design features, testing requirements and documentation are all critical components during the design phase of a new structure.

For instance, a major manufacturer of seismic protection systems was recently asked to design a customized solution for a new oil rig structure. After analyzing the specifications, the team decided to use a tuned mass damping system with laminated rubber bearings and viscous dampers to help carry the static load, provide proper lateral stiffness and optimum critical damping ratio,. These combined technologies helped to improve the total system robustness, allowing for a more reliable structure and extended service life.

Extensive Testing Prior To Development Is Key Because critical applications require highperformance devices, extensive testing is

important for any damping technology being applied to a structure to ensure accuracy, repeatability and longevity.

As technology evolves, it is becoming imperative that structural engineers consult FVD manufacturers early and often to ensure achievement of the best available solution for any application.

What to look for when testing damping performance: • Consistent and symmetrical damping force over multiple cycles • No dead-band in performance between tension and compression cycles • Test data that matches data for expected performance outcomes • Data that matches application requirements • Cycle outcomes that are repeatable and measurable When designing a structure to withstand the impact of seismic energy, it’s important to consider utilizing additional support systems to provide adequate protection. Modern FVDs and other energy dissipation devices can extend the life of your structure, ensure safety for occupants, and foster a more reliable operation.▪ Shubin Ruan, Ph.D., P.E., is a project engineering manager for ITT Enidine.

Shubin may be reached at shubin.ruan@itt.com.

Ben Eder is the Americas Infrastructure Sales Manager for ITT Enidine Inc. Ben may be reached at ben.eder@itt.com.

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award winners and outstanding projects

Spotlight

Innovation as Homage Chhatrapati Shivaji International Airport – Terminal 2

By Charles Besjak, P.E., S.E., AIA Skidmore, Owings & Merrill LLP was an Outstanding Award Winner for the Chhatrapati Shivaji International Airport – Terminal 2 project in the 2014 NCSEA Annual Excellence in Structural Engineering awards program (Category – International Structures over $100M).

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s recently as 15 years ago, only 6 million travelers used Chhatrapati Shivaji International Airport each year. Today the facility can accommodate more than 40 million passengers annually, and the new SOM-designed Terminal 2 building was created precisely to handle this greatly expanded scale of operation. The 4.4 million-square-foot building combines international and domestic passenger services under one roof, optimizing terminal operations and reducing passenger walking distances. In the building’s X-shaped plan, moreover, concourses radiate from a central processing core so they can “swing” between domestic and international service 24 hours a day.

The Terminal 2 scheme is as elegant as it is efficient – a point exemplified by its headhouse roof. Beautifully molded and punctuated by skylights, the roof spans four base structures whose programming includes roadside dropoff, check-in, and security and immigration.

This 753,475-square-foot surface is supported by only 30 composite mega-columns. It is one of the world’s largest roofs without an expansion joint.

To achieve it, the mega-columns are spaced 210 feet in one direction and 112 feet in the perpendicular direction, and support a system of orthogonal 13-foot-deep steel trusses akin to a two-way flat slab. Truss depths increase near the mega-columns to create “column pod” areas, and the truss system cantilevers as much as 131 feet from the nearest megacolumn at the roof edge.

Moreover, SOM separated the megacolumn-and-roof system from the four base structures, which comprise stacks of pouredconcrete bays. Pierced openings in these stacks allow the mega-columns to ascend unimpeded, so that the headhouse roof responds independently to loads that most notably include expansion and contraction caused by temperature variation; thermal gradient was applied to the steel in the structural analysis model, and it informed the design and fabrication of individual roof members.

In addition to impacting response to loads, SOM’s Courtesy of Robert Polidori | Mumbai International Airport Pvt. Ltd. disaggregation of the base concrete bays from the headhouse roof struc- conditions, as well as the inclusion of corture helps to produce the unique aesthetic of ners, curves, and entrance vestibules, all Terminal 2. The facility’s monumental spaces mandated a very precise design of cable preand mushrooming columns evoke the airy tension. And owing to the shape and scale of pavilions and interior courtyards of traditional the headhouse roof and cable wall, a windIndian architecture. tunnel study was carried out to determine Similar to the monumentally sized head- cladding pressures and structural roof loads. house roof, Terminal 2 claims the title of These unusual efforts also supported the having the longest and largest cable wall in regionalist aesthetic. Because they are installed the world. The facility’s two separate cable with insulated glass units, the cable walls wall systems total over 3,280 feet in length create a diaphanous perimeter that makes and 118,403 square feet in area.

Terminal 2’s pavilion-like appearance highly The cable walls employ unidirectional cables legible from afar. that span two vertical levels of the termiThe overall engineering solution yielded a nal structure. At the uppermost departures variety of other positive outcomes. Among floor, for example, unidirectional cables others, it is a tailor-made response to site conwere necessitated by the cable roof ’s com- straints. The X-shaped plan molds Terminal plete envelopment of the terminal headhouse, 2 around the existing terminal, which had to which precluded horizontal anchorage remain operational during construction, and points. Here, expansion joints installed at it incorporates modular designs to accommothe intersection of the cable wall and four date rapid phased construction. Moreover, the base structures allows individual segments of glazed cable wall works in tandem with the the wall to move independently. skylit headhouse roof and the piercings in the On the east and west sides of Terminal 2, mean- concrete base structures to maximize daylight while, the cable wall reaches to the cantilevered penetration and reduce energy consumption. edge of the headhouse roof. Because the roof Finally, the mega-columns served as primary experiences large vertical deflections, a backup hoist mechanisms so that the headhouse roof system prevents a loss of cable pre-tension due could be constructed without the use of tower to roof deflections. In this system, composite cranes; individual column pods were concolumns cantilevering from concrete base struc- structed atop one peripheral mega-column tures are interconnected with structural steel and then transported to their final location girders at the top and bottom of the cable wall; via temporary erection trusses. the unidirectional cables span between these Yet, of all these benefits, visual identity takes girders, and the top of the back-up system is priority. CSIA bridges historic central Mumbai supported laterally by the headhouse roof with and its burgeoning peripheral neighborhoods, dampers at composite column locations. and Terminal 2 weaves the airport into the In addition to its unique size and length, wider community in both form and function. the CSIA cable walls include a number of Empowered by engineering, the building’s features rarely pursued elsewhere. To pre- architectural expression has captured the herivent collision between the curved portion of tage and vision of contemporary Mumbai.▪ the wall and absolute corners, for example, Director Charles Besjak, P.E., S.E., AIA, corner cables are reinforced with horizonis in charge of structural engineering for tal stiffener plates connected between the SOM’s New York office. He can be reached cable and the corner column of the backup at Charles.besjak@som.com. system. Variations in height and anchoring

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Online Young Member Trivia Night a Huge Success

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News form the National Council of Structural Engineers Associations

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NCSEA Young Member Groups converged for the first online, nationwide Trivia Night on May 21. 11 teams with 60 participants met online in a battle of wits.

The Trivia Night was led by Sofia Zamora, EIT, of CBI Consulting, and Scott Lefton, EIT, of IES, members of the NCSEA Young Member Group Support Committee, and was designed to promote interaction and connection among Young Member Groups in NCSEA Member Organizations.

Through Google Hangouts and Google Forms, with online timers, teams gathered at locations with screens and internet access. The event was divided into East and West Coast to account for time zone differences, and each team location was led by a moderator. Each session began with an introduction of the Young Member Group Support Committee and an invitation to attend the NCSEA Structural Engineering Summit in Las Vegas. Over the course of the next hour or so, the teams competed, answering questions on sport history, natural disasters, structural engineering design theory, facts about the U.S., structural engineering history, and a tie breaker on food.

Winning teams were a team from Washington, DC, led by Brian Petruzzi of Weidlinger Associates from the East Coast, and the West Coast team from Colorado, led by Jeremy Crandall of Structural Consultants and Katie Courtright. Each winning team received a cash award.

The teams got into the spirit of competition, using the chat box to post comments. The Michigan team’s extensive use of the chat box to banter with the other teams made for a fun East Coast event.

Online & nationwide, the event featured 11 teams: East Coast: Massachussetts (2 teams) North Florida Washington, DC (2 teams) Michigan Connecticut

West Coast: Colorado Montana - Bozeman Montana - Great Falls Oregon

Feedback from the participants was overwhelmingly positive, and plans are underway for another Trivia Night and other events to connect Young Member Groups.

East Coast winners from Washington, DC.

The team from Michigan.

NCSEA Webinars More detailed information on the webinars and a registration link can be found at www.ncsea.com.

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July 21, 2015 ACI Development of a Building Code for Repair of Existing Concrete Structures

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July 14, 2015 The Most Common Errors in Wind Design & How to Avoid Them

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NCSEA News

The two teams from Massachussetts.

Diamond Reviewed

Keith Kesner, Ph.D., P.E., S.E., Chair, ACI Committee 562

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Non-CalOES courses award 1.5 hours of continuing education. Approved for CE credit in all 50 States through the NCSEA Diamond Review Program. Time: 10:00 AM Pacific, 11:00 AM Mountain, 12:00 PM Central, 1:00 PM Eastern. NCSEA offers three options for registrations to NCSEA webinars: Ala Carte, Flex-Plan, and Yearly Subscription. Visit www.ncsea.com for more information or call 312-649-4600.

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A reservation link for the NCSEA group rate of $180 plus $15 resort fee, which includes complimentary valet and self parking, airport shuttle, in-room internet and Spa & Fitness Center access, is available at www.ncsea.com.

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Whether you like gambling, entertainment, shopping or outdoor activities, the Red Rock area of Las Vegas has you covered!

The host hotel, Red Rock Resort, features restaurants, lounges, a movie theatre and bowling alley, and a walkway to Downtown Summerlin, with shops, dining and entertainment options.

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Saturday, October 3 8:00 – 12:00 NCSEA Annual Business Meeting 12:30 – 2:00 NCSEA Board of Directors Meeting

STRUCTU

More than just a conference hotel!

News from the National Council of Structural Engineers Associations

Thursday, October 1 7:00 Delegate Interaction Meeting 8:00 Welcome & Introduction 8:15 – 9:30 Keynote, Ashraf Habibullah, SE, President and CEO, Computers & Structures, Inc. 9:45 – 11:00 Basis for ASCE 7 Seismic Design Maps, Ron Hamburger, P.E., S.E., SECB, Senior Principal, Simpson Gumpertz & Heger 11:00 – 12:00 Building Rating, Retrofit Ordinances, and Community Resilience, Panel from Structural Engineers Association of California 1:00 – 2:15 A. The ASCE 7-16 Tsunami Loads Design Standard, Gary Chock, S.E., President, Martin & Chock; Chair, ASCE 7 Tsunami Loads & Effects Subcommittee B. Effective Communication: Tips for Improving Your Skills, Kirsten Zeydel, S.E., President, ZO Consulting & Annie Kao, P.E., Field Engineer, Simpson Strong-Tie 2:45 – 3:45 A. Wood & Cold-Formed Light Steel Frame Construction - Deficiency in the IBC Special Inspections, Kirk Harman, P.E., S.E., SECB, President, The Harman Group B. Find the Lost Dollars: 6 Steps to Improve Profits, June Jewell, CPA, AEC Business Solutions 4:00 – 5:00 A. Changes to Wind Loading in ASCE 7-16, Don Scott, S.E., PCS Structural Solutions; Chair, NCSEA Code Advisory Committee Wind Engineering Subcommittee B. BIM and Structural Engineering, Desiree Mackey, P.E., S.E., BIM Manager, Martin/Martin 6:30 – 8:30 Welcome Reception on Trade Show Floor 8:30 – 11:30 (Red) Rock ‘n Bowl event

Friday, October 2 8:00 – 10:00 Member Organization Reports 8:00 – 10:00 Vendor Product Presentations 10:15 – 11:30 A. Lateral Design of Buildings with Sloped Diaphragms, Steven Call, P.E., S.E., Call Engineering B. Working with Multiple Generations, Panel Discussion with the NCSEA Young Member Group Support Committee 1:00 – 2:30 A. Lateral Analysis: Right Way/Wrong Way with Software, Sam Rubenzer, P.E., S.E., Structural Engineer, FORSE Consulting B. Quality Assurance for Structural Engineering Firms, Cliff Schwinger, P.E., SECB, Vice President & Quality Assurance Manager, The Harman Group 2:45 – 3:45 A. Concrete & CMU Elements in Bending + Compression, John Tawresey, S.E., retired, KPFF Consulting Engineers B. The Decline of Engineering Judgement, Jon Schmidt, P.E., SECB, Associate Structural Engineer, Burns & McDonnell 4:00 – 5:00 A. Problem Solving for Repairing Wood Structures, Kimberlee McKitish, P.E., Nutec Group B. Business Ownership Transfer, Craig Barnes, P.E., SECB, Founding Principal, CBI Consulting 6:00 – 7:00 Awards Reception (formal attire encouraged, but not required) 7:00 – 10:00 NCSEA Banquet & Awards Presentation, featuring the NCSEA Excellence in Structural Engineering Awards and the NCSEA Special Awards

NCSEA News

Wednesday, September 30 8:00 – 5:00 Committee Meetings 8:00 – 12:00 NCSEA Board of Directors Meeting 5:30 – 6:30 Young Engineer Reception 6:30 – 8:30 SECB Reception

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Structural Columns

The Newsletter of the Structural Engineering Institute of ASCE

Electrical Transmission & Substation Structures Conference 2015 September 27 – October 1, 2015, Branson, Missouri Grid Modernization – Technical Challenges & Innovative Solutions Earn up to 19 PDHs. The ASCE/SEI Electrical Transmission & Substation Structures Conference is recognized as the must-attend conference that focuses specifically on transmission line and substation structures and foundation construction issues. This event – for utilities, suppliers, contractors, and consultants – offers an ideal setting for learning and networking.

Technical Program Sunday, September 27, 2015 Pre-Conference Workshop – Panel Discussion of Storm Handling, Resiliency and Security Issues Monday, September 28, 2015 10:15 a.m. – 12:00 p.m. Technical Session 1 – Structural Analysis 1 1:30 – 3:15 p.m. Technical Session 2 – Special Design Considerations 4:00 – 5:20 p.m. Technical Session 3 – Managing Aging Infrastructure

Tuesday, September 29, 2015 8:10 – 9:30 a.m. Technical Session 4 – Structural Analysis 2 10:15 a.m. – 12:00 p.m. Technical Session 5 – Case Studies 1:30 – 3:15 p.m. Technical Session 6 – Foundations 4:00 – 5:20 p.m. Technical Session 7 – Substation Design Issues Wednesday, September 30, 2015 8:10 – 9:30 a.m. Technical Session 8 – Construction Challenges 10:15 a.m. – 12:00 p.m. Technical Session 9 – Line Design 1:30 – 2:50 p.m. Technical Session 10 – Rerating and Upgrading 3:35 – 4:55 p.m. Technical Session 11 – Codes and Standards Thursday, October 1, 2015 Technical Demonstrations Day Visit the conference website at www.etsconference.org for complete information and to register.

RegistRation open – Second ATC-SEI Conference Improving the Seismic Performance of Existing Buildings and Other Structures December 10 –12, 2015, Hyatt Regency San Francisco The Second ATC-SEI Conference on Improving the Seismic Performance of Existing Buildings and Other Structures will provide a forum for the presentation and exchange of new information on the seismic evaluation and seismic rehabilitation of existing buildings, including case studies, new discoveries, innovative use of new technologies and materials, implementation issues, needed improvements to existing standards and methods, and socio-economic issues. The goal is to provide an invaluable opportunity to advance the understanding of the tools, techniques, and innovations available to assist the attendees in meeting the challenges of seismic evaluation and rehabilitation.

Save the Date

Errata SEI posts up-to-date errata information for our publications at www.asce.org/SEI. Click on “Publications” on our menu, and select “Errata.” If you have any errata that you would like to submit, please email it to Jon Esslinger at jesslinger@asce.org.

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Technical Program Each day of the conference will start with presentations by two keynote speakers.

Thursday, December 10, 2015 Four concurrent tracks on the following subjects: • Resilience/Mitigation Programs • Concrete Structures/Nonlinear Analysis • Historic Structures/Emergency Technologies • Standards and Guidelines Friday, December 11, 2015 Four concurrent tracks on the following subjects: • Managing Risk • Nonstructural/Wood Soft Story • Implementation of Case Studies • Innovative Solutions for Retrofit Saturday, December 12, 2015 Six concurrent sessions on the following subjects: • Napa Earthquake • Resilience Framework • Steel Structures Retrofit • Case Studies • Future Directions Visit the conference website at www.atc-sei.org for complete details and to register.

July 2015

Sign up for any ASCE live webinar to be held in June, July, or August – either 60 or 90 minutes – and get a special individual member rate of $99, applied automatically when registering. Take advantage of convenient, efficient training that provides practical knowledge and earns PDHs. This offer does not apply to site/group webinar registrations, on-demand webinars, or P.E. exam review courses and cannot be combined with other offers. Visit the ASCE Continuing Education website at www.asce.org/continuing _education for more information.

SEI Election Announcement July 31, 2015 Deadline structural dynamic systems, seismic protection, earthquake engineering, structural control, system identification, structural health monitoring and applied nanotechnology. NSF has awarded him the prestigious faculty early CAREER award for innovative research in Adaptive Stiffness Structures (1998).

He and his coworkers were awarded the American Society of Civil Engineers (ASCE) Moisseiff award in 2015, for their 2014 Journal of Structural Engineering on Adaptive Negative Stiffness Structural Systems. His editorial activities include service as the Managing Editor of the Journal of Structural Engineering [ASCE International journal] (2011 – present), Editor of the Structural Control and Health Monitoring [Wiley International Journal] (2008 – present) and Editor-in-chief [North America] of Structural Monitoring and Maintenance [Techno-press International Journal] (2014 – present).

ASCE Structural Engineering Institute (SEI) elected him as an inaugural fellow in 2012 for his service to the structural engineering community. He served as the chair/vice-chair/secretary/ member (2006 to 2012) of ASCE, SEI, Technical Activities Division Executive Committee (TAD-ExCom). He was the founding chair of ASCE structural health monitoring committee (2004 – 2006), ASCE-Engineering Mechanics Institute, and chair of the structural control committee (1998 – 2002), ASCE Structural Engineering Institute. A native of India, he obtained his B.S.C.E from Bangalore University (1980) and M.S. from Indian Institute of Science (1982). Then, serving as a structural design engineer (1982 – 1986) for Tata Consulting Engineers, India, designed thermal and nuclear power stations, before arriving in the United States for his doctoral studies. For further details visit his website www.satishnagarajaiah.rice.edu.

Full Name: _____________________________________Member’s ASCE/SEI ID No:________________ (Please print) Date:______________ Signature: _______________________________________________________________

Return postmarked no later than July 31, 2015 to: SEI Board Election, 1801 Alexander Bell Dr., Reston VA 20191.

SEI 2015 Board of Governors Election Official Ballot

Technical Activities Division

q Satish Nagarajaiah q Write-in vote:_______________________________

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July 2015

The Newsletter of the Structural Engineering Institute of ASCE

The current Board of Governors positions on the Structural Engineering Institute Board of Governors are: representatives from each of the four Divisions (Business and Professional, Codes and Standards, Local Activities, and Technical Activities), one appointed by the ASCE Board of Direction, and the most immediate and available Past President of the SEI Board. The representatives from the Divisions each serve a four-year term.

In accordance with the Structural Engineering Institute Bylaws, this year SEI is conducting an election for a Technical Activities Division representative on the Board of Governors. The TAD Executive Committee has nominated Satish Nagarajaiah as its candidate. In accordance with the SEI Bylaws, each ballot provides a space for a write-in vote. If you are a member of SEI please complete and mail your ballot to the address provided.

Either vote for the named candidate OR provide a write-in candidate. Because we must confirm SEI membership, only signed ballots will be accepted.

Satish Nagarajaiah, Ph.D., F.SEI, M.ASCE, is a Professor of Civil and Environmental Engineering at Rice University in Houston, Texas.

He is affiliated with the Mechanical Engineering Department, Material Science and Nano-Engineering Department at Rice. He obtained his Ph.D. (1987 – 1990) from State University of New York at Buffalo, where he was a post-doctoral researcher before he started his academic career in 1993. His teaching and research interests are in the areas of structural engineering,

Structural Columns

Take Any of ASCE’s Live Webinars this Summer for $99 Each

CASE in Point

The Newsletter of the Council of American Structural Engineers

New Guideline Now Available!!

National Practice Guideline on Project Risk Management and Business Risk Management is intended to assist structural engineering companies in the management of risk associated with projects and to provide commentary regarding the management of risk associated with business practices. The guideline is organized in two sections that correspond with these two areas of risk, namely Project Risk Management and Business Practices Risk Management. The goal of the guideline is to educate and inform structural engineers about risk issues so that the risks they face in their practices can be effectively mitigated, thus making structural engineering firms more successful.

Developed by the CASE Guidelines Committee, this document is available at www.acec.org/coalitions/coalition-publications/.

CASE Risk Management Tools Available Foundation 1 Culture: Create a Culture for Managing Risks & Preventing Claims

Foundation 2 Prevention & Proactivity: Act with Preventive Techniques…Don’t Just React

Tool 1-1 Create a Culture for Managing Risks and Preventing Claims

Tool 2-1 A Risk Evaluation Checklist Don’t overlook anything! A sample itemized list of things you should look for when evaluating a prospective project.

Inject into your firm a culture of risk management. This is the first and most comprehensive tool offered on risk management in the engineering industry. It includes a video, a story board and role playing guide to involve your staff in the risk management discussion. If you want to start your firm personnel on the path to good risk management habits, this is where you start.

Tool 1-2 Developing a Culture of Quality

Tool 2-2 Interview Guide and Template Getting “the right people on the bus” is one of the most important things we can do to mitigate risk management, and yet we never learn about interviewing skills in school. It is the second tool related to the Second Foundation of Risk Management, Prevention and Proactivity. The tool will help your firm conduct higher quality interviews and standardize the process among all your staff.

Culture is hard to define in an organization but it is a key part of what gives a design firm character. CASE Tool No. 1-2, Developing a Culture of Quality was developed to identify ways to drive quality into a firm’s culture. It is recognized that every firm will develop its own approach to developing a culture of quality, but following these 10 key areas offer a substantial starting point. The tool includes an attached white paper and PowerPoint presentation that can be customized to facilitate the overall discussion.

Tool 2-4 Project Risk Management Plans This plan will walk you through the methodology for managing your project risks, along with a few common project risks and templates on how to record and track them.

You can purchase these and the other Risk Management Tools at www.acec.org/coalitions/coalition-publications/.

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Tool 2-3 Employee Evaluation Templates This tool is intended to assist the structural engineering office in the task of evaluating employee performance. The evaluations provide a method to assess employee performance and serve as an integral part of the company’s risk management program.

Tool 2-5 Insurance Management: Minimize Your Professional Liability Premium This tool is designed as a guide to help you provide critical additional information to the underwriter to differentiate your firm from the pack.

July 2015

CASE Announces the 2015 Scholarship Winner

The CASE Summer Planning Meeting is scheduled for August 6 –7 th in Chicago, IL. The night of August 6th will feature a roundtable discussion on risk management and business practice topics. Firms can develop business relationships leading to growth and success through the CASE roundtables held twice each year. Topics in the past have included the Business of BIM, using social media within your firm, Peer Review and Special Inspections. Attendees to this session will earn 2.0 PDHs. If you are interested in attending the roundtable/meeting, please contact CASE Executive Director Heather Talbert at htalbert@acec.org.

The CASE scholarship, administered by the ACEC College of Fellows, is awarded every year to a deserving student seeking a Bachelor’s degree, at minimum, in an ABET-accredited engineering program. Since 2009, the CASE Scholarship program has given $15,000 to engineering students to help pave their way to a bright future in structural engineering. CASE strives to attract the best and brightest to the structural engineering profession and educational support is the best way we can ensure the future of our profession. The 2015 winner, Shawna Peterson, will graduate May 2016 with a Master’s Degree in Structural Engineering from Stanford University.

Since its inception in 1995, the American Council of Engineering Companies’ prestigious Senior Executives Institute (SEI) has attracted public and private sector engineers and architects from firms of all sizes, locations and practice specialties. Executives – and up-and-coming executives – continue to be attracted by the Institute’s intense, highly interactive, energetic, exploratory, and challenging learning opportunities.

In the course of five separate five-day sessions over an 18-month timeframe, participants acquire new high-level skills and insights that facilitate adaptability and foster innovative systems thinking to meet the challenges of a changed A/E/C business environment.

The next SEI Class 21 meets in Washington, D.C. in September 2015 for its first session. Registration for remaining slots is available.

Executives with at least five years’ experience managing professional design programs, departments, or firms are invited to register for this unique leadership-building opportunity. As always, course size is limited, allowing faculty to give personal attention, feedback, and coaching to every participant about their skills in management, communications, and leadership.

SEI graduates say that a major benefit of the SEI experience is the relationships they build with each other during the program.

Participants learn that they are not alone in the challenges they face both personally and professionally, and every SEI class has graduated to an ongoing alumni group that meets to continue the lifelong learning process and provide support.

For more information, visit http://sei.acec.org/ or contact Deirdre McKenna, 202-682-4328, or dmckenna@acec.org.

Wanted

Engineers to Lead, Direct, and Get Involved with CASE Committees!

If you’re looking for ways to expand and strengthen your business skillset, look no further than serving on one (or more!)

CASE Committees. Join us to sharpen your leadership skills – promote your talent and expertise – to help guide CASE programs, services, and publications.

We have a committee ready for your service: • Risk Management Toolkit Committee: Develops and maintains documents such as business practices manuals and policies for engineers under CASE’s Ten Foundations for Risk Management.

Follow ACEC Coalitions on Twitter – @ACECCoalitions.

STRUCTURE magazine

Expectations and Requirements To apply, you should • be a current member of the Council of American Structural Engineers (CASE) • be able to attend the groups’ two face-to-face meetings per year: August, February (hotel, travel partially reimbursed) • be available to engage with the working group via email and conference call • have some specific experience and/or expertise to contribute to the group Please submit the following information to htalbert@acec.org • Letter of interest • Brief bio (no more than 2 paragraphs) Thank you for your interest in contributing to your professional association!

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CASE is a part of the American Council of Engineering Companies

A/E Industry’s Premier Leadership-Building Institute Filling Fast for September Class – FOUR SLOTS LEFT!!

CASE in Point

CASE Summer Planning Meeting – Save the Date

Structural Forum

opinions on topics of current importance to structural engineers

Does a Name Have Value?

By Barry Arnold, S.E., SECB

Y

our name is probably your greatest asset. Contained in a name are a person’s or company’s brand, identity, and reputation. There was a time when agreements were made based only on a person’s name. Does a name have value? Absolutely! A name is so important that each year, people and companies spend millions to clear their name, protect their name, and promote their name.

The following are a few specific examples.

A Name is Used to Market a Product Fishermen in the south pacific catch a fish named the Patagonia Toothfish. Despite its delicious taste, it never found a broad market because of its unsavory name. In 1977, the name was changed to Chilean Sea Bass, and it became an immediate culinary sensation.

A great amount of money and effort is spent annually by corporations to research product names to enhance marketability.

A Name Provides Clarity When a person suffers from acute abdominal pain, they seek the assistance of a gastrointestinal specialist. Although a podiatrist or rheumatologist can offer advice, they will likely not provide an accurate diagnosis and remedy. A name reduces confusion, aiding the person seeking a specific type of services.

A Name Increases Clout & Credibility Thirty years ago, geotechnical engineers were often referred to as “dirt guys” or “soils engineers,” which undermined (no pun intended) the important service that they provide. They undertook a grassroots effort to enhance their reputation by promoting the name “geotechnical engineer.” It took time, but they now receive the respect that they deserve.

A Name Confers Distinction and Acknowledges Achievements When someone at a project meeting introduced himself as an architect and an engineer, he was questioned about his credentials. Upon discovering that he was a registered architect and had completed a BS in civil engineering, but was not licensed as an engineer, he was told not to

say or imply that he is an engineer unless and until he met all of the corresponding education, experience, and examination requirements.

This particular incident notwithstanding, the effort to protect the engineer title was fought and lost many years ago. We now have a plethora of people calling themselves engineers who are not engineers at all. The fight to protect the PE title is now underway.

Although the name “Professional Engineer” is protected, the abbreviation PE is not, allowing contractors to call members of their construction team “project engineers” and advertise them as “PEs.” “Structural Engineer” is a protected title only in those states that have passed legislation to that effect. Elsewhere, that title can be used by anyone – even those who have no particular skills or knowledge of structural design.

Instead, engineers are left to decide for themselves which projects they will pursue, guided only by a code of ethics. The assumption is that engineers will “perform services only in the areas of their competence” and “hold paramount the safety, health, and welfare of the public.” This approach has worked, to a degree, but has received mixed reviews as to its success and value.

That may seem good enough for us, but our clients – whom I define as persons or companies that put their money and livelihood at risk by relying on the expertise of the structural engineer – feel differently. In the absence of action by the engineering profession, many public and private clients have assessed the situation and provided their own remedy, taking matters into their own hands to protect their investments. Relying on government entities, professional organizations, and ethical individuals to ensure that only qualified individuals are allowed to practice engineering and use certain titles is a gamble that they do not have to take.

In a world where one can get a PhD online for any profession at next to no cost, clients have reason to be skeptical when it comes to believing everything that they read or hear in response to a Request for Proposals (RFP).

Even though minimum credentials and

qualifications of a structural engineer may not be legally defined in the state where the project is located, they nevertheless publish RFPs requiring licensed structural engineers. Some have even gone so far as to establish their own criteria regarding education, experience, and examination, and they summarily reject proposals that do not meet their standards.

Protecting the Structural Engineer title has been a hot topic for many years, such that NCSEA, SEI, and CASE have individually published position statements supporting structural licensure –a key part of which is to define and protect the title. These organizations have joined together in a collaborative effort and formed the Structural Engineering Licensure Coalition (SELC) to promote structural licensure.

The name Structural Engineer is worth protecting because it can be used to market, provides clarity, adds clout and credibility, and is a title of distinction – it is an honor to possess it. Looking forward, the structural engineering profession should be at the forefront of the effort to protect the Structural Engineer title in every state.

To support this effort, my advice to all those practicing structural engineering is to take the 16-hour NCEES Structural exam. It does not matter what organizations you join, what your state SEA is doing, or how your legislators feel about structural licensure – take the exam. It does not matter if your company will support or reimburse you – take the exam. Passing the exam is an accomplishment that you can take with you anywhere and note proudly on any resume or proposal. Becoming a licensed Structural Engineer is a personal decision and one that will have long-term benefits throughout your professional career.▪ Barry Arnold, S.E., SECB (barrya@arwengineers.com), is a Vice President at ARW Engineers in Ogden, Utah. He is a Past President of the Structural Engineers Association of Utah (SEAU), serves as the SEAU Delegate to NCSEA, and is the current President of NCSEA, as well as a member of the NCSEA Structural Licensure Committee.

Structural Forum is intended to stimulate thoughtful dialogue and debate among structural engineers and other participants in the design and construction process. Any opinions expressed in Structural Forum are those of the author(s) and do not necessarily reflect the views of NCSEA, CASE, SEI, C 3 Ink, or the STRUCTURE® magazine Editorial Board.

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July 2015