SEI 3 2017

SEI Volume 27 | Number 3 | August 2017 Structural Engineering International

STRUCTURAL ENGINEERING INTERNATIONAL International Association for Bridge and Structural Engineering (IABSE)

Volume 27 | Number 3 | August 2017

Building for the future.

TWIN RIVER BRIDGE, CHONGQING, CHINA T.Y. Lin International (TYLI) designed the Dongshuimen Bridge across the Yangtze River and the Qianximen Bridge across the Jialing River – two iconic structures known collectively as the Twin River Bridge in Chongqing, China. Interconnected by a tunnel, the bridges allow seamless transport between the city’s southern district, downtown Chongqing, and the northern district. Both structures carry vehicular lanes and pedestrian paths on the upper decks and light rail on the lower decks.

TYLI’s innovative design resulted in two partially cable-supported girder bridges, an economical, structurally efficient concept that provides a highly aesthetic, transparent configuration. The simple, modern towers also dramatically express Chongqing’s dynamic combination of Photo: Yaping Lai

www.tylin.com

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IABSE

both historical and contemporary features.

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Structural Engineering International International Association for Bridge and Structural Engineering

Contents Nr. 3/2017 Editorial The Responsibility of the Bridge Designer; H. Beade-Pereda, UK

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Special Topic: Forensics in Structural Engineering Scientific Papers*

Front façade after the collapse, pg. 394

Collapsed building in final stage of construction, pg. 429

Introduction: Forensic Structural Engineering; A. M. Ivankovic´, Croatia; R.T.Ratay, USA An Overview of Forensic Structural Engineering; R. T. Ratay; USA Forensic Investigation on Cracking in Hanger-to-Girder Connections of Long-Span Suspension Bridges; Z. Liu, T. Guo, S. Pan, J. Liu; China Investigating Structural Stability of Historic Buildings: A Case Study of a Rampart Wall of Hilandar Monastery; N. Šekularac, J. Cˇ. Tovarovic´, J. Ivanovic´-Šekularac; Serbia Pathology, Assessment and Treatment of Structures Affected by Delayed Ettringite Formation; B. Godart; France Human Errors in the Design and Execution of Steel Structures—A Case Study; I. Tylek, K. Kuchta, A. Rawska-Skotniczny; Poland Rehabilitation and Lessons Learned from the Collapse of Viaduct 1 Located on the Caracas–La-Guaira Highway in Venezuela; L. B. Fargier Gabaldón; Venezuela Determination of Defect Factors for a Road Bridge Made of Prestressed Concrete; A. Madaj, W. Siekierski; Poland Forensic Investigations of Collapses of Buildings in “Aggregate”: Lesson Learnt from a Meaningful Case Study; F. Palmisano, P. Perilli; Italy Failure Identification: Procedural Causes and Corresponding Responsibilities; E. Brehm, R. Hertle; Germany

338 338 344 353 362 370 380 388 394 402

Technical Reports

View of the support area on the pier with two types of old bearings, pg. 442

Requisite Diversity of Expert Witness Testimony and Issues of Objectivity, Bias and Advocacy; J. B. Kardon; USA Problems Associated With Curtain Walls; J. Cˇ. Tovarovic´, N. Šekularac, J. Ivanovic´-Šekularac; Serbia Investigation of the Cause of Failure of the Omo River Bridge; G. Zerayohannes, E. Gebreyouhannes, A. Zekaria-Abdullahi; Ethiopia Damage Investigation of a Tunnel Subjected to an Unplanned Surface Load Through Non-Linear Analysis*; O. Arnau, C. Molins, A. Marí; Spain Investigation of the Collapse of a Building in the Final Stage of Its Construction; M. R. Grbic, N. Dzombic, M. T. Glisic; Serbia Improved Structural Health Monitoring of London’s Docklands Light Railway Bridges Using Digital Image Correlation; J. Winkler; Denmark, C. Hendy; England Assessment of Structural Steel’s Properties in a Railway Plate Girder Bridge*; J. Hołowaty; Poland

409 413 418 422 429 435 441

Technical Reports (General topics) Retrofitting of Three Historic Stone Arch Bridges over the Cetina River in Croatia; J. Radnic´, M. Smilovic´, N. Grgic´, F. Borovina, B. Borovina, A. Mlinar; Croatia Superstructure of the National Museum of African American History and Culture; R. Laberenne, G. J. P. Nordenson, E. Oswald, N. Krgovic; USA

449 454

IABSE News

Front Cover: Collapsed Omo River Bridge near Omarate. See article on pg. 418

IABSE Awards 2017 IABSE Conference Kuala Lumpur 2018 IABSE Denmark’s Structural Award 4th Young Engineers Colloquium Bochum 2017 - Report IABSE Conference Bath 2017 - Report 2017 IABSE Future of Design Conference NYC - Reflections Trip to China IABSE Book Vernissage: SED 5 – 3rd Edition Calendar of Events *Peer-reviewed papers

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Structural Engineering International International Association for Bridge and Structural Engineering SEI: Structural Engineering International (SEI) is the leading journal of structural engineering dealing with all types of structures and materials. Articles are written by practicing engineers and academia from around the world, and reflect the highest standards. All articles are reviewed by the SEI Editorial Board, with collaboration from over 250 experts and 28 correspondents. Scientific papers that have undergone a detailed peer-review by independent experts and demonstrate a significant contribution to the state of structural engineering knowledge are marked by the peer review stamp. To recognise contributions of the highest quality, an Outstanding Paper Award is presented each year. SEI is the quarterly journal of the International Association for Bridge and Structural Engineering (IABSE).

Aims and Scope: Structural Engineering International (SEI) publishes technical reports and scientific papers with a focus on structural engineering, to bridge the gap between professionals and academics. SEI is a journal with a mix of articles ranging from practical to scientific. SEI publishes articles on recent structures, developments in practice and research. Articles in SEI consider all types of structures, bridges, buildings, offshore structures, etc., constructed from all types of building materials: concrete, steel, composites, timber, fibre reinforced polymers, glass, etc. Articles in SEI also deal with the sustainability of structures and a broad range of structural engineering practice from design to construction to maintenance and monitoring, and finally to the re-use or demolition of structures.

SEI Editorial Board Ann Schumacher, dsp Ingenieure & Planer, Switzerland (Chair) Guido Morgenthal, Bauhaus-Universität Weimar, Germany (Vice Chair) Elsa de Sá Caetano, University of Porto, Portugal* Charles-Darwin Annan, Université Laval, Canada Flávio de Souza Barbosa, Federal University of Juiz de Fora, Brazil Stathis Bousias, University of Patras, Greece *Nominated

Andrea Frangi, ETH Zurich, Switzerland Dongzhou Huang, Atkins, United States Adam Lubell, Read Jones Christoffersen, Canada Anastasios Sextos, University of Bristol, United Kingdom Ana Mandic Ivankovic, University of Zagreb, Croatia Tina Vejrum, COWI, Denmark Scott Walbridge, University of Waterloo, Canada Dong Xu, Tongji University, China Takashi Yamaguchi, Osaka City University, Japan

SEI Advisory Board USA: John E. Breen; USA: Wai-Fah Chen; Japan: Yozo Fujino; Denmark: Niels J. Gimsing; UK: Peter R. Head; Switzerland: Manfred A. Hirt; UK: David A. Nethercot; Germany: Jörg Schlaich; Canada: Peter R. Taylor; France: Michel Virlogeux; The Netherlands: Joost C. Walraven; China: Hai-Fan Xiang.

SEI Correspondents Australia: D. Morris; Brazil: M. Schulz; Canada: J. Jiang; China: Z. Sun; Denmark: M. Braestrup; Egypt: F. Saad; Finland: A. Laaksonen; France: B. Godart; Germany: U. Kuhlmann; Greece: C. Gantes; Hungary: L. Dunai; India: V. Kumar; Italy: G. Bignotti, L. Ceriolo; Japan: K. Fujita; Korea: H.-K. Kim; Norway: L. Toverud; Poland: H. Zobel; Portugal: E. Caetano; Russia: S. V. Mozalev; South Africa: J. Wium; Spain: J. F. J. Alonso; Sweden: H. Sundquist; Thailand: E. Limsuwan; The Netherlands: A. de Boer; UK: T. Harris; USA: J. G. Burns, D. Frangopol.

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forces in motion

24.11.16 16:18

The Responsibility of the Bridge Designer Bridges link previously separate geographic areas, defying gravity and transforming the landscape. They bind communities, acting as connectors of people, inviting interaction and integration. Bridges are a paradigmatic case of human transformation of nature, symbols of union, progress and often innovation.

Bridges cross physical, cultural and spiritual barriers, frequently becoming landmarks or even icons. They are much more than a piece of infrastructure and the design of such emotional, prominent and long-lasting constructions should go beyond their main function as structures that link areas, and should always aspire to improve the quality of the built world.

Héctor Beade-Pereda IABSE Prize Recipient 2016 IABSE BG Executive Committee Associate, Bridge Designer, Knight Architects, UK

Even in the early stages of their history, bridges transcended their mere functional role to become objects with aesthetic ambitions. For a long time, the aim of minimising construction materials, meant the former aspiration was frequently combined with an external expression of the structural behaviour, resulting in a clean and legible appearance. Nowadays, far less emphasis is put on the appearance or material optimisation, instead replaced by minimised effort and time of both designers and contractors. In addition to this, excessive or unnecessary solutions are often conceived and developed, encouraged by the disappearance of technical calculation and construction limitations.

Every single bridge is unique and significant. During their centennial lives, most of them will be crossed by countless people, whose experience will be determined by the quality of the work done by those involved in their design and construction. For this reason, ease and speed in design and construction should rarely be considered the central tenets of a bridge project. Bridges should be designed for their users (in most cases they are publicly funded), not for the designer’s own benefit, selfishly trying to fill gaps in our portfolios or, at the other end, to simplify our existence by just repeating an existing bridge (designed by us or by others). Self-criticism, identifying when the problem exceeds our skills (or could be better with additional skills) and understanding when we need additional learning or other’s input or involvement is also fundamental.

Bridge design is a thrilling challenge that comes with great responsibility. The responsibility of designing structures that should have at least a 100-year service life, a time period that is too long for bad design. Good design is not just the designer’s responsibility, but for every party involved in the process: this includes those who identify the crossing need and, those who decide the procurement route and time and economic allocation, and those who construct the bridge, among many others.

Good bridge design requires creativity, sensitivity, honesty and effort. Good design shouldn’t be expensive; on the contrary, it is likely that it will offer more for the same or less than a bad design. If a bridge is unnecessarily expensive, the design cannot be considered a good one. What is frequently too expensive is not investing in design and suffering with a bridge that is improvable for generations to come.

Beauty is the result of what is right, says a Japanese aphorism. The ideal bridge should be not only functional, structurally coherent, aesthetically pleasing, durable and sustainable, but also respectful of the history and culture of the site, appropriate to the environment and of a suitable scale to its context. It should also meet any specific boundary conditions (hydraulic, seismic, navigational or wind and impact related), be concerned about detail and be financially viable. A solution that satisfactorily harmonizes all these design requirements need not be a mere pipe dream. It is in our hands. It is our responsibility and IABSE is the perfect vehicle for bridge designers to help each other to achieve this goal.

Héctor Beade-Pereda Structural Engineering International Nr. 3/2017

Editorial

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Introduction: Forensic Structural Engineering This Special Issue on Forensic Structural Engineering begins with an overview of the professional practice through origins, types and causes of failures, outlining the process of forensic investigation and revealing forms of dispute resolutions and the role of an expert consultant/witness. Papers coming from Europe as well as from China, USA, Venezuela, Mexico and Ethiopia are mainly scientifically oriented, which shows that research within forensics engineering is very popular, necessary and significant worldwide. In addition, several examples of forensic investigations are presented in order to reveal causes of collapses or deteriorations and to prescribe measures to avoid similar adverse structural performance.

Scientific contributions begin with the paper predicting the fatigue life of hanger-to-hanger connections in suspension bridges. A methodology for detecting undesirable deformations, in compliance with the conservation requirements, is developed through the

case study of a rampart wall in a historic monastery.

A specific phenomenon, delayed ettringite formation, is described from pathology and assessment to final treatment solutions classified according to their purpose. Further on, in a case study of a steel structure, human errors are revealed as major cause of structural failures. Investigation of the collapse of a historic concrete arch bridge resulting from the reactivation of an ancient landslide offers a lesson for the assessment of soil–structure interaction in similar situations. The approach for determination of defect factors is presented through the example of faulty design of prestressing in a concrete bridge. Forensic investigation of the collapse of a set of adjacent buildings aims to explore the general criteria for similar cases in historical city centres.

To prevent failures and avoid the loss of human life, an optimised reliability management system is proposed in the final scientific paper.

Through technical reports, readers are introduced to the concepts of objectivity, bias and advocacy, emphasising the importance of diversity of opinion in the resolution of disputes. Further on, examples of problems associated with curtain walls in buildings, launching methods in bridge construction, unplanned surface loading in a segmental tunnel lining and the necessity of following construction standards and good practices are discussed. In addition, digital image correlation is noted as a promising tool to collect objective measures of structural performance and assessment of structural steel necessary in the upgrade of old railway bridges. Few more contributions on forensic issues will be presented in the forthcoming issues of SEI. Ana Mandi c Ivankovi c, Chief reviewer, SEI Editorial Board Robert Ratay, Guest reviewer

An Overview of Forensic Structural Engineering Robert T. Ratay, PhD, P.E., Consulting Structural Eng., Adjunct Prof., Department of Civil Engineering, Columbia University, New York, New York, USA. Contact: robert@robertratay.com DOI: 10.2749/101686617X14881932437176

Abstract Forensic structural engineering has, in some countries, become and, in some others, is evolving into a designated field of professional practice of determining the causes of structural failures and identifying the parties responsible. The practice involves engineering investigations, rendering opinions and, if required, giving expert testimony in judicial proceedings. Whether they occur during construction or during their service lives, failures of constructed facilities are almost always followed by engineering investigations and resolution of claims. As the findings inevitably create claims of damage and often result in disputes and legal entanglements, the forensic structural engineer operates in an adversarial environment and therefore needs not only to be able to perform the necessary investigations but also to have an understanding of the ins and outs of the practice of forensic engineering—which this paper intends to enhance.

This paper is an overview of the professional practice of forensic structural engineering and includes discussions of: what structural failure is, origins and causes of failures, types of cases of failures, importance of codes and standards, the standard of care for design professionals and duty to perform for constructors, brief outline of the process of forensic investigation, with references to detailed publications, meaning and importance of “reasonable degree of engineering certainty,” forms of dispute resoluPeer-reviewed by international ex- tions, who is and what the role of an perts and accepted for publication expert consultant/witness is, a brief look

by SEI Editorial Board at how the practices differ in various IABSE countries, and selected reference Paper received: May 15, 2016 literature.

Paper accepted: November 10, 2016 338

Review Paper

Keywords: forensic engineering; structure; failure; investigation; expert.

Introduction Increasing numbers of structural failures make the news throughout the world, and their investigation has become an active, important and lucrative field of practice. Some countries have established practices; others approach it in an ad-hoc manner.

This paper is a composite, update and expansion of past publications of the author in IABSE SEI 2007,1 STRUCTURE 20092 and ICE Forensic Engineering 20093 with the purpose of informing the novice and updating the experienced structural engineers on what the professional practice of Forensic Structural Engineering is all about today.

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Forensic Structural Engineering is the professional practice of investigating and determining the causes of failures of constructed facilities, as well as rendering opinions and, if required, giving testimony in judicial proceedings.

As a designation of a field of professional practice, the term Forensic Engineering appears to be used most consistently in the United States, where its application in the investigation of failures, and especially in the role and activities of the expert consultant/witness, follow well-accepted practices both in the engineering and in the legal arenas.4,5 Indeed, it appears to have been in the United States where it was first recognized as a specialty practice by the American Society of Civil Engineers, driven by the need to deal effectively and credibly with the investigation of significant structural failures and by the necessity of competent technical support in the litigation of the resulting damage claims. To be clear, despite the widespread use of the title Forensic Engineer, to date, none of the 50 State Boards of Licensure—that defines who qualifies to practice engineering and that issues Professional Engineer licenses—includes the designation Forensic Engineer.

IABSE’s Working Group on Forensic Structural Engineering, formed in 2012, has conducted a survey of the existence and practice of forensic engineering in the IABSE member countries. Selected findings have been presented,6,7 and a comprehensive report is in preparation.

One of the survey’s findings is that the designation “Forensic Engineer,” or its equivalent in a country’s language, is known and used in more than a third, not known nor used in less than a third, and the responses are not clear in the rest of the of the surveyed countries.

The purpose of this paper is to acquaint architects, engineers, constructors and building officials as well as attorneys, insurers and owners of constructed facilities with the engineering practice and some rudimentary legal aspects of forensic engineering, as it is practiced primarily in the United States, in order to determine the causes of and indicate the responsibilities for the failure of a structure.

The author has made some effort to explore the engineering and legal practices dealing with failures in a number of countries, and despite the fact that he has performed forensic investigations and lectured in countries outside

as well as in states inside the United States, he cannot help viewing the subject from his position in the United States, where for decades he has been practicing structural design, condition assessment and investigation of failures as an expert consultant/witness.

What Is Structural Failure?

Structural failure may be defined by various words; it is “nonconformance with design expectations” or “unacceptable difference between intended and actual performance.” It ranges from hardly visible minor defects to catastrophic collapses.

Any of the following, listed in order of seriousness, qualify for structural failure: • • • • • • • •

Unacceptable aesthetic appearance Unreasonable maintenance needs Excessive deformation Signs of distress Local failure Extensive failure but no collapse Partial collapse Total failure (collapse)

There also are cases with no readily visible signs. These can be suspected design errors, construction defects and hidden deterioration that present risks of potential failures but may not be recognized without structural analysis or field probing and testing.

Causes of Failure Things can go wrong anywhere along the planning–design–construction– service lives of a constructed facility.

These cases and some of what can go wrong in them include the following: • Planning: system concept • Design: concept, calculations, drafting • Design–construction interface: shop drawing detailing, drafting, review and approval • Construction: erection, inspection, accident • In service: misuse, overload, alteration, adjacent construction • Deterioration: lack of maintenance Failures are always the result of human action, which lie in one or more of the following: • Negligence: failure to properly analyze or detail the design or the disregard of codes and standards • Incompetence: failure to understand engineering principles or respect

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• • • • •

the technical limitations of materials or systems Oversight: failure to follow design documents and safe construction practices Greed: shortcuts; intentional disregard of industry requirements and safe practices Disorganization: failure to establish a clear organization and define responsibilities of parties Miscommunication: failure to establish and maintain lines of communication among the parties.

Misuse, abuse, neglect: using the facility for purposes beyond its design intent or foregoing preventive maintenance.

The forensic engineer needs to have an understanding of the design– construction process, including the various methods of project delivery, in order to recognize where along the process a particular error was likely to occur.

Types of Cases The types of cases of failures that forensic engineering expert consultants/witnesses encounter can be categorized into three groups: • When a “condition” surfaced: This is when some structural deficiency, such as possible overstressing, cracking, excessive deformation, objectionable vibrations, etc., is observed.

• When a failure or collapse occurred: This is when part or all of the structure has failed or collapsed.

• When dispute resolution or litigation is already in progress: This is when the damage had already occurred sometime in the past, and the legal activities, such as preparation for trial in court, are underway.

The expert consultant’s role and scope of work in each of the three types of cases is addressed in the Forensic Investigation Process section.

Importance of Codes, Standards and Regulations In forensic engineering work, design codes and standards provide the baseline for the design requirements that were in effect at the time when the design was done, and they define the minimum level of performance that the failed structure should have met.

Satisfying the governing code is a minimum requirement. Copies, especially Review Paper

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old copies, of design codes, standards and manuals are valuable references on a forensic engineer’s bookshelf.

During construction in the United States, inspectors of the US Government Office of Safety and Health Administration (OSHA) make unannounced visits to construction sites, issue violations when warranted and impose fines for violations of the socalled OSHA Regulations. By act of Congress, OSHA is to investigate all construction failures where fatalities occur. OSHA’s report opines on the innocence or culpability of the contractor(s) at the site; it does not address the designer’s responsibilities.

The OSHA Regulations and the OSHA report on the particular incident are significant documents for the forensic engineer to review in his or her engineering investigation and for opining on the responsibilities.

Standard of Care and Duty to Perform It is important to recognize that engineers, as other professionals, do not guarantee perfection. (What is perfection anyway?) So, how can engineers be judged for their errors?

In the United States, engineers and other professionals have to meet a “standard of care” (in some languages, the phrase in English translation is “expectable quality” or something similar). For an authoritative discussion of the standard of care, see Refs. [8,9]. Some countries use other approaches to judging engineers for their errors, such as the Netherlands where the engineer is held liable for “culpable faults”—well explained in Section 3 of Ref. [10].

The undertaking of a design project by an engineer implies that he or she possesses skill and ability sufficient to enable him or her to perform the required services at least ordinarily and reasonably well and that he or she will exercise and apply to the project his or her skill, ability and judgment reasonably and without neglect. This is a promise, and this is all a client can expect from a professional, be it an architect, engineer, physician or other professional.

Since the engineer, when providing professional services, is exercising judgment gained from learning, background and experience and usually provides 340

Review Paper

those services in conditions where certain unknown or uncontrollable factors are common, some level of error in those services has to be acceptable.

The designer does not guarantee perfection, only that he or she exercised ordinary care.

A legal definition of the standard of care in the United States is the Book of Approved Jury Instructions:11 “In performing professional services for a client, a professional has the duty to have that degree of learning and skill ordinarily possessed by reputable professionals, practicing in the same or similar locality and under similar circumstances. It is his or her further duty to use the care and skill ordinarily used in like cases by reputable members of his or her profession practicing in the same or similar locality under similar circumstances, and to use reasonable diligence and his or her best judgment in the exercise of professional skill and in the application of learning, in an effort to accomplish the purpose for which he or she was employed. A failure to fulfill any such duty is negligence.” (Underlines by author) It is important to note that the qualifications of learning and skill, and the expected use of those qualifications, are compared to those possessed and used by other reputable professionals who are practicing in the same or similar locality and under similar circumstances. The professional not only has to have the qualifications but also has to use them together with reasonable diligence and best judgment.

There are other definitions as well, created by attorneys and engineers. A simplified working definition of the standard of care, composed by the author, is: “That level and quality of service ordinarily provided by other normally competent practitioners of good standing in that field, when providing similar services with reasonable diligence and best judgment in the same locality at the same time and under similar circumstances.”

When examining an engineer’s performance on a project, it is important for the expert witness to recognize that the applicable standard of care is not what an engineer should have done in someone’s opinion, or what someone claims that he or she and/or others would have done, nor is it what in someone’s opinion an engineer would do now but,

rather, what normally competent engineers actually did do at the time under similar circumstances.

It is important to note that expert consultants/witnesses are also held to a standard of care—indeed to an elevated standard of care because they hold themselves out as experts.

Contractors are held to a “duty to perform” (the phrase coined by this author), that is, to follow strictly the plans and specifications and the provisions of their contracts, as well as the regulations of the authorities having jurisdiction. For these reasons, the evaluation of a contractor’s responsibility for a failure can be more objective than that of an engineer’s.

Forensic engineers must have a clear understanding of all these performance requirements because with their opinions and testimonies as expert consultants/witnesses, they can ruin long-built reputations and destroy precious livelihoods of people.

The Forensic Investigation Process The process of forensic structural engineering investigation is different from that of conventional structural engineering design.

Perhaps the most basic difference is that in design, we use predicted loads multiplied by load factors to increase their values on one side and material strengths multiplied by resistance factors to reduce their values on the other so as to result in desired margins of safety through the life of the structure; in failure analysis, we use the actual loads and actual strengths at the time of the failure.

The process of a typical forensic investigation of a failure, and hence the activities of the expert consultant investigating the case, may be coarsely outlined as follows: • First-response and preliminary assessment • Development of investigation strategy • Fact gathering and document review • Engineering analyses to determine cause(s) and responsibilities • Reporting If the failure results in legal action, further investigation and review of the work of other experts may be required as part of the technical support in the

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dispute resolution or litigation, possibly followed by giving expert testimony.

rules of evidence and, as such, is a legal term.

Except for litigation support, which is a function of the prevailing judicial system, the process is not country-dependent. Rather, it depends on the nature and magnitude of the failure, fatalities if any, the monetary value of damages and claims, the time and money available and, sometimes, on the particular experts and attorneys involved.

Engineers should be thorough in their investigations and careful with the statements of their opinions. They should be aware that forensic investigation, advice to a client and expert testimony can be subject to claims of negligence, and the liability may extend far beyond the fee for services.

The scope of the forensic engineer’s work also depends on the timing of his or her engagement and first involvement in the case, which can occur when a “condition” surfaced, when a failure or complete collapse occurred or when dispute resolution or litigation is already in progress. The conditions at these times are briefly noted in the Types of Cases section earlier in this paper; the scopes of work are listed in Refs. [1–3].

An adequate discussion of the investigation process is outside the scope of this paper, so for additional information with sufficient detail, see Refs. [4,12].

Let it suffice to say here that the scope of work can vary widely from a simple walk-through visual inspection—essentially a structural condition assessment—when a condition surfaced to a multi-phase complex investigation when a significant failure or total collapse occurred and on to technical support when dispute resolution or litigation is in progress.

Of particular importance after a partial or complete collapse is the “first response.” The engineer’s first steps after a failure can be critical for at least two reasons: preventing further damage and human casualty and documenting the collapse configuration and securing perishable evidence that may influence the validity of the entire investigation. For an informative discussion of good practices, the reader is referred to the 30-page chapter The First Steps after a Failure, listed here as Ref. [13].

Reasonable Degree of Engineering Certainty Professional opinion cannot always be given with 100% certainty. The courts in the United States do not require that the expert’s opinion be with absolute certainty; they only require a “reasonable degree of engineering certainty.”

The phrase “reasonable degree of scientific certainty” comes from courtroom

A credible, honest, ethical investigator satisfies himself or herself of the legitimacy of his or her conclusions and supports his or her opinion with sound technical foundation. Nevertheless, in his or her report, when expressing an opinion, he or she should use a phrase such as “it is my opinion with reasonable engineering certainty that . . .”

Resolution of Disputes The Code of Laws of Hammurabi (circa 1750 BC), King of Babylonia, settled damages arising from structural failures in a simple and straightforward manner, which, in essence, was: 1. If a builder builds a house for a man and does not make its construction meet the requirements and a wall falls in—that builder shall strengthen the wall at his own expense.

2. If it destroys property—he shall restore whatever it destroyed, and because he did not make the house which he built firm and it collapsed, he shall rebuild the house which collapsed at his own expense.

3. If a builder builds a house for a man and does not make its construction firm and the house which he has built collapses and causes the death of the owner of the house—that builder shall be put to death.

4. If it causes the death of the son of the owner of the house—they shall put to death a son of that builder.

5. If it causes the death of a slave of the owner of the house—he shall give to the owner of the house a slave of equal value.

To the good fortune of architects, engineers and contractors, the gruesome provisions of Hammurabi’s laws have not survived.

In the United States, as in most other countries, damage disputes are resolved in one of several ways, which include:

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• Settlement discussion – among the parties, usually involving compromise by all.

• Mediation (nonbinding) – with an impartial mediator, agreed upon by the feuding parties, seeking to bring about a resolution through diplomacy and conciliation.

• Arbitration (binding) – with one or more arbitrators hearing evidence and arguments of the parties, asking probing questions and deciding on a resolution through a quasi-judicial process.

• Trial in court – before a judge and sometimes a jury, resulting in a binding verdict.

Successful settlement discussions are the best, least bruising, most private and least expensive ways of resolving disputes. Mediations tend to bring out all, or nearly all, the facts and are successful and satisfying in their outcome if the parties act in good faith. Arbitration has the attraction of relatively speedy outcomes after the completion of testimonies, and the relative privacy of the parties in comparison to a trial; however, they may be time-consuming and, at the end, costly. The feuding parties usually try to avoid a trial in court because of its lengthy procedures, great expense and often unpredictable outcome; hence, only a handful of cases of structural failures end up in trial in court.

The Expert Consultant/Witness The role of the forensic structural engineer, as an expert consultant, is to determine the cause(s) of failure of a constructed facility and to provide the technical bases upon which the responsibilities for the failure can be identified. A forensic engineer investigator, or expert consultant, is not an expert witness until a court of law or some other dispute resolution forum formally recognizes him or her as such.

In order for a forensic engineer to intelligently investigate the cause of a failure and, subsequently, to point to the party(ies) responsible for it, he or she has to have an understanding of not only loads, strength and stability but also of the processes of design and construction, and even the prevailing business practices, in order to know where, when, how, why and by whom a cause of failure can originate.

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investigation of the cause(s) of failures, and a truthful expert witness is valuable and often necessary in the resolution of claims.

In the United States and some other countries, following a structural failure, each party involved usually retains a forensic expert consultant to investigate the causes of and responsibilities for the failure and, if needed, to provide technical support to that party’s legal team in the resolution or litigation of the resulting claims. This same consultant usually continues as the expert witness of the client to provide expert testimony and expert opinion at arbitration or trial if the case goes that far.

For the system to work—and it does work—it is of the utmost importance that, as stated by Canon 3c of the ASCE Code of Ethics and noted in Ref. [4], “engineers, when serving as expert witnesses, shall express an engineering opinion only when it is founded upon adequate knowledge of the facts, upon a background of technical competence, and upon honest conviction.”

In yet some other countries, the court appoints an expert to advise it on the technical aspects of the case. Each of the feuding parties also usually retains an expert who may or may not be allowed to challenge the courtappointed expert’s opinion.

The system of opposing experts retained by the feuding parties versus a court-appointed expert is often debated. What is best: (i) opposing experts expressing diversity of opinion—with possible partiality to their clients—and challenging each other or (ii) one court-appointed expert—with possible pre-existing opinions and prejudices—who may or may not be challenged by other experts or (iii) is there something else? The opinions are fodder for another paper.

Under the United States system, justice’s demand for truth and clarity may be in conflict with the client’s need for helpful testimony. This creates the ethical dilemma that the expert witness must resolve to his or her own and everyone else’s satisfaction. One’s ethics, biases and impartiality versus advocacy in a case are subjects of much scrutiny. Because of the adversary nature of the proceedings, ethical conduct is taken very seriously and watched with vigilance in the United States.

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Under the court-appointed expert witness system of other countries, it is the competence, thoroughness and credibility of the expert that are the most heavily scrutinized.

In the United States, the judge is “gatekeeper” on qualifying an expert witness, that is, allowing him or her to testify, as well as on admitting the expert’s testimony as evidence. The person is qualified only if he or she has the satisfactory education, background and experience applicable to the case, and his or her opinions in the testimony are admitted (accepted) only if those opinions are shown to be both reliable and helpful, that is, the opinion “rests on a reliable foundation and is relevant to the task at hand.” [This was held in Daubert v. Merrell Dow Pharmaceuticals, 509 U.S. 579 (1993), a United States Supreme Court case determining the standard for admitting expert testimony in federal courts. It assigns to the trial judge the task of ensuring that an expert’s testimony both rests on a reliable foundation and is relevant to the task at hand.] Commonly referred to as the “Daubert test,” it is intended to keep “junk science” out of the courtroom.

For more discussions of the role, qualifications and ethics of the expert engineering witness, see Refs. [14–17] that view the subject from the United States, Australian, Canadian and British perspectives, respectively. A rather thorough discussion is the 21page chapter The Expert Consultant and Witness, listed here as Ref. [18].

Is Forensic Engineering Universal?

The short answer is No. As noted in the Introduction, at the time of this writing, the IABSE Working Group 8 Forensic Structural Engineering is completing a survey of the state of forensic engineering in the 49 IABSE member countries. Early findings from the survey, reported in Ref. [7] included that “Forensic Engineering, as such, is practiced in some form in every country, it is not designated and treated as a field of professional practice in many of them, and only three or four of them have engineering organizations named and dedicated to forensic engineering … Organized and structured forensic investigation practices, maintenance of data bases, and dissemination of structural failure information exist in only a few countries.”

In the United States, the engineering and legal response to failures is almost institutionalized—which is not necessarily a good thing. Parties damaged by failures have an almost “knee-jerk” reaction to call their attorneys rather than attempting “peaceful” discussions; insurance companies and law firms have developed routine defensive responses to claims of damage arising from failures. Expert consultants are soon retained by each party in an adversarial environment to conduct the investigation of the failure and to provide technical assistance to the party in the litigation of the resulting claim. The American Society of Civil Engineers (ASCE) has a unit called Forensic Engineering Division, FED (formerly Technical Council on Forensic Engineering, TCFE), a booklet called Guidelines for Forensic Engineering Practice, 2nd ed.,4 a quarterly publication called Journal of Performance of Constructed Facilities,19 tri-annual national forensic engineering conferences, seminars, short courses and active forensic engineering groups in many of the ASCE local sections. In addition to the ASCE Forensic Engineering Division, there is the independent organization of National Academy of Forensic Engineers with its own activities.

The UK, mainly through its Institution of Civil Engineers (ICE), also has a vibrant forensic engineering community that, among other things, sponsors tri-annual international conferences and publishes the quarterly Forensic Engineering Journal.20 A rather unique organization is the office of the Forensic Science Regulator (FSR) in the UK. As explained in Ref. [21] the FSR had been created to ensure that forensic science services are subject to an appropriate regime of quality standards. To that end, the FSR publishes codes of practice and conduct and guidance on the issue of bias.

As noted in the section on the Expert Consultant/Witness earlier in this paper, in some countries, the feuding parties each retain an expert to be its consultant and witness; in some others, the court appoints an expert.

The resulting dynamics of the litigation are quite different.

In some countries, the engineering and legal professions follow established practices; in others, they are relying on ad-hoc procedures struggling to respond to the technical and legal challenges.

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Yet others are seeking to establish technical, business and legal procedures often modeled on US practices—with both their effectiveness and faults.

The knowledge and information available in Europe, the United States, Japan, Asia and in the Pacific area are noted very briefly in Ref. [22].

Literature There is a large amount of informative literature on forensic engineering and investigations in the United States, in the UK and a fair amount in other countries. A comprehensive guide to the practice of forensic engineering in the United States is the ASCE Guidelines for Forensic Engineering Practice, 2nd edition.4 ASCE’s Journal of Performance of Constructed Facilities19 and the ICE’s Forensic Engineering20 journal both deal specifically with articles on forensic engineering. A rather comprehensive treatment of forensic practices within the fields of structural engineering and construction is the 800-page Forensic Structural Engineering Handbook, 2nd edition.5 A condensed overview of the investigation of structural collapses is the Investigation and Analysis of Structural Collapses23 in the Elsevier Encyclopedia of Forensic Sciences, 2nd Ed. A book that addresses the practice of forensic engineering beyond the fields of structural engineering and construction is Forensic Engineering, 2nd edition.24 A way of gaining a good understanding of forensic investigations of construction failures is to read published reports of cases among the Construction Incidents Investigation Engineering Reports25 of the OSHA of the US Department of Labor. Condensed descriptions of some 50 cases of structural failures in the book Beyond Failure26 offer a good representation of the types of cases and investigations in forensic structural engineering. An informative series of both philosophical and case study articles originating from Australia on forensic engineering is published monthly online under the collective title of BHCitations.27 There are many more applicable and useful references dealing with forensic structural engineering in the worldwide literature. The reader is cautioned, however, to recognize what represents accepted and good practice and credible valid investigation and

what represents untested opinions of individuals.

[11] Book of Approved Jury Instructions. BAJI 6.37, Duty of a Professional. Jury Instructions, Civil, West Publishing Company: Los Angeles, 1986.

Closure

[12] Cuoco D, & Panariello G. Chapter 6: The engineering investigation process. In Forensic Structural Engineering Handbook, Ratay RT (ed), 2nd edn. McGraw-Hill, Inc.: New York, 2010: pp.

6.1–6.18.

Forensic structural engineering is gaining recognition as a field of professional engineering practice in many parts of the world. At one time or another, nearly all engineering firms do forensic engineering work as part of their practice or as occasional service to their favored clients. There are firms and individuals who specialize and are well known for their expertise in forensic work. It is an intellectually challenging, at times physically and emotionally demanding, but altogether satisfying professional practice. The often combative environment of adversity and the legal arena, however, are not for everyone’s temperament.

References [1] Ratay RT. Forensic Structural Engineering – Focus on the USA. IABSE: Zurich, 2007.

[2] Ratay RT. Professional practice of forensic structural engineering – what every engineer should know. STRUCTURE 2007, July: 50–53.

[3] Ratay RT. Forensic structural engineering practice in The USA. Civil Eng. J. 2009; 162(5): 52–56.

[4] Technical Council for Forensic Engineering of the American Society of Civil Engineers (TCFE of ASCE). Guidelines for Forensic Engineering Practice, Kardon JB (ed) ASCE: Reston, 2013.

[5] Ratay RT (Ed.). Forensic Structural Engineering Handbook, 2nd edn. McGraw-Hill, Inc: New York, 2010.

[6] Terwel K, et al. An initial survey of forensic engineering practices in some European countries and the US. Proceedings of the ASCE 6th Congress of Forensic Engineering, San Francisco, 1–3 November 2012

[13] Peraza D. Chapters 5: The first steps after a failure. In Forensic Structural Engineering Handbook, Ratay RT (ed), 2nd edn. McGrawHill, Inc.: New York, 2010.

[14] Ratay RT. The forensic structural expert consultant/witness – US practices. Proceedings IABSE 2006 Annual Meetings and Symposium, Budapest, 13–15 September 2006.

[15] Brady S. The structural engineer as expert witness – forensics & design. Proceedings of the 2nd National Conference of the Society of Construction Law, Brisbane, 5–6 August 2011. Portions of the paper titled ‘Causation, the structural engineer, and the expert witness’ published in Construction Law International, Volume 6, Issue 2, June 2011 [16] Association of Professional Engineers of Ontario. Guideline: Forensic Engineering Investigations Association of Professional Engineers of Ontario (PEO): Toronto, 2016.

[17] Clancy B. Briefing: reflections on the role of the expert witness. Forensic Eng. J. 2016; 169(2): 31–43.

[18] Thompson DE, & Aschraft HW. Chapter 19: The expert consultant and witness. In Forensic Structural Engineering Handbook, Ratay RT (ed), 2nd edn. McGraw-Hill, Inc: New York, 2010.

[19] Journal of Performance of Constructed Facilities. Forensic Engineering Division (formerly Technical Council of Forensic Engineering), ASCE, Reston.

[20] Forensic Engineering Journal. Institution of Civil Engineers (ICE) UK. [21] Mackay A. Impact of the forensic science regulator (FSR). Forensic Eng. J. 2016; 169(2): 44–45.

[22] Wenzel H. Forensic engineering in the process of structural innovation. Forensic Eng. J. 2016; 169(2): 50–52.

[7] Palmisano F, Ratay RT. The practice of forensic structural engineering in IABSE member countries: review of a survey. Proceedings of the IABSE Conference, Geneva, 23–25 September 2015.

[23] Ratay RT, & Peraza DB. Investigation and analysis of structural collapses. In Encyclopedia of Forensic Sciences, J. A. Siegel and P. J. Sankko, 2nd edn. Elsevier Ltd: New York, 2013; 461–465.

[8] Kardon JB. The standard of care and negligence of the forensic engineer. Proceedings of the 37th IABSE International Symposium, Madrid, 3–5 September 2014

[24] Carper KL (Ed.). Forensic Engineering, 2nd edn. CRC Press: Boca Raton, 2000.

[9] Kardon JB. Chapter 4: Standard of care. In Forensic Structural Engineering Handbook, Ratay RT (ed), 2nd edn. McGraw-Hill, Inc: New York, 2010: pp. 4.1–4.15.

[10] Boot W, Terwel K, & Strang H. Legal matters related to structural damage in the Netherlands. Forensic Eng. J. 2015; 168(3): pp. 117–126.

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[25] Office of Safety and Health Administration (OSHA), US Department of Labor. Construction incidents investigation engineering reports. https://www.osha.gov/doc/engineering/ EXengrptsr.html [26] Delatte NJ. Beyond Failure ASCE Press: Reston, 2009.

[27] BH Citations. http://www.bradyheywood. com.au/bhcitations/archive

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Forensic Investigation on Cracking in Hanger-to-Girder Connections of Long-Span Suspension Bridges Zhongxiang Liu, PhD Candidate; Tong Guo, Prof., Dr. Shengjun Pan, MSc.; Jie Liu, PhD Candidate, Key Laboratory of Concrete and Prestressed Concrete Structures of the Ministry of Education, Southeast University, Nanjing, P. R. China.

Contact: guotong@seu.edu.cn DOI: 10.2749/101686617X14881937384963

Abstract The hanger-to-girder connection is a vital load-bearing component of suspension bridges. However, two types of premature cracking (types A and B) in hanger-to-girder connections of the Runyang Suspension Bridge have been observed only several years after the bridge was open to traffic. This paper presents the field inspections on the pattern, location and propagation of cracks.

To further investigate the mechanism of cracking, a multi-scale finite element (FE) model of the bridge is developed. It is observed from FE analysis results that under vehicle loads, large stress range and significant stress concentration exist. The type A cracks are mainly due to the discontinuity of hanger-to-girder connections and in-plane shear/bending deformation, and out-of-plane bending deformation of the welded connections accelerated type B cracking. To this end, probabilistic FE analyses are performed, and the predicted fatigue lives of the hanger-to-girder connections are obtained and compared with those in the field, where good agreement is observed.

Keywords: forensic investigation; hanger-to-girder connection; fatigue crack; suspension bridge; out-of-plane deformation; probabilistic analysis.

Introduction Suspension bridges have been extensively used in modern long-span bridges due to their aesthetic shape, relatively easy construction and exceptional spanning capability as well as excellent antiwind and anti-seismic properties.1–3 Hung below the main cable through hangers, bridge girders are subjected directly to vehicle loads,4–7 and the hanger-to-girder connection is thus among the most vital load-transferring parts of suspension bridges.8,9 Figure 1 shows several typical hanger-to-girder connections. The bearing-type anchor connection, as illustrated in Fig. 1a, consists of the bearing slab, backing slab, anchor plate and stiffeners. To provide adequate capacity for transferring force, the bearing-type anchorages are always bi-directionally strengthened into an anchor box with stiffeners. The main disadvantages of such an anchor connection include complicated load path and incompatible deformation with transversally inclined hangers. The San Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: June 24, 2016 Paper accepted: August 30, 2016

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Francisco–Oakland Bay Bridge10 and Taohuayu Bridge in China11 adopted such an anchor connection. The pipetype anchor connection in Fig. 1b, as compared with the bearing-type anchor connection, has a simple configuration with a concise load path and is particularly suitable for hangers with transversal inclination. Such a connection has been applied in the Yongjong Bridge9 and the Jiangdong Bridge.12 However, due to the vibration and deformation of the main cables or hangers, the hanger may be subjected to both tension forces and bending moment at the contact region of the hanger and long pipe, resulting in the fracture of wires.13 Similarly to the bearing-type anchor connection, monitoring and maintenance of the pipe-type anchor connections below the girder are inconvenient.

The auricular plate anchor connection, as shown in Fig. 1c, consists of an auricular plate, web and transversal diaphragm of the main girder. The pin and the connection between auricular plates and welds are critical regions. The auricular plate anchor connection provides simplified fabrication with clear and smooth load transmission; moreover, almost all elements are in pure tension. In the Xihoumen Bridge in China14 and Akashi Kaikyo Bridge in

Japan,15 the auricular plate anchor connection has been adopted with double auricular plates and stiffeners.

Owing to the complex geometrical configuration and loading conditions, the hanger-to-girder connection is among the most fatigue-prone components of suspension bridges.5 However, so far, investigations regarding the performance of hanger-to-girder connections are very limited, which focus on laboratory tests on segmental specimens16,17 and finite element (FE) simulations.11,18 For example, an investigation was conducted on the local stress and fatigue behaviour of an anchor pinned to a deck-welded joint and the underneath reinforcement structure of a box-type suspension bridge.17 The weak point and fatigue strength of the anchor under ordinary load and wind load conditions were examined experimentally.

However, the field performance (i.e. cracking) of the hanger-to-girder connections has not be reported and investigated previously.

This paper reports the observed cracking in hanger-to-girder connections of a long-span suspension bridge in China. To provide a basis for rational design and repair of hangerto-girder connections, the pattern, location and propagation of cracks are evaluated based on non-destructive field tests, which were conducted every year from 2007 to 2015. To further look into the failure mechanism, a multi-scale FE model of the bridge is developed, and probabilistic FE analyses are performed to predict the fatigue lives of the specific details in conjunction with a vehicle load model developed using traffic information from the toll station; good agreement is observed in the model.

Field Inspection and Statistical Analyses Runyang Suspension Bridge The Runyang Bridge, open to traffic in 2005, consists of a cable-stayed

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(a)

(b)

(c) Hanger

Hanger

Top deck

Top deck

Hanger

Anchor plate Auricular plate Stiffener

Anchor pipe Web of main girder

Bearing slab Transversal diaphragm of main girder

Transversal diaphragm of main girder

Web of main girder Backing slab Web of main girder

Fi g. 1: Typical hanger-to-girder connections; (a) bearing-type anchor connection, (b) pipe-type anchor connection and (c) auricular plate anchor connection

bridge and a suspension bridge (Runyang Suspension Bridge, RSB) with a main span of 1490 m, as shown in Fig. 2a. The RSB consists of streamlined, closed, flat, all-welded box-girder segments, and the girder cross-sections with wind fairing are 36.3 m wide and 3.0 m high, as depicted in Fig. 2b. The box-girder is

cable through a cable band, while the lower end is linked to the stiffening girder through an auricular plate, as illustrated in Fig. 2d. As shown in Fig. 2a, the hanger-to-girder connections are labelled from north to south as locations 1–45, 46 (mid-span) and 47–91, respectively. Figure 2e further presents the three-dimensional (3D)

supported by two main cables through 88 pairs of hangers and one central clamp19 (in Fig. 2c) on the upper or lower stream side; the rigid central clamps are employed at the mid-span instead of short hangers to increase the fatigue performance of the hangers near mid-span. The upper end of the hanger is connected with the main

(a) Yangzhou North

1–45

46 Mid-span

Zhenjiang South

47–91

1490 m

(b)

(c) High strength

Main cable

2%

Bearing plate

Centerline

Auricular plate 4500

bolt (M30)

Cable band

1834

34300/2=17150 Hanger

1765 906

3184

Main cable

3000

West 1000

diagonal beam

Transverse diaphragm 2760

27300/2=13650

3220

3220

600 Vertical beam

Wind fairing leading line 3220

3220

2760

18400

(d)

Cable band

(e) Main cable

0,5m

Bracket Stiffening rib A

B

Inclined deck

hanger Bearing plate

Auricular plate

Auricular plate Inclined deck Bearing plates

Fi g. 2: The Runyang Suspension Bridge (dimension unit in mm); (a) profile, (b) cross-section of steel box-girder, (c) central clamp, (d) hanger anchor connection and (e) 3D configuration of the anchor connection

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(a)

(b)

Crack Crack

F ig . 3: Field-observed cracks (types A and B) at hanger-to-girder connections; (a) crack between bearing plate and stiffener (connection A) and (b) crack between auricular plate and inclined deck (connection B)

configuration of the hanger-to-girder connection, where the stiffening ribs and brackets are connected to the middle and the side bearing plates through welds. All the plates are of Q345D-grade steel with a nominal yield strength of 345 MPa. To ensure the weld performance and impact ductility, the carbon, phosphorous and sulphur contents should be no more than 0.18, 0.025 and 0.015%, respectively. In the impact ductility tests,20 the impact energy of the Charpy specimens at −20 C is required to be no less than 34 J, and no crack is allowed in the 180 cold bending tests, with the diameter of curve centre d = 15a, where a is the thickness of plates. The single full-penetration fillet weld has a height of 7 mm, and during the welding, CO2 is used for protection.

Field Inspection Periodical non-destructive field inspections were conducted every year since 2007, using ultrasonic detection,21 a

dye-penetrant22 and magnetic particle inspections23 so as to monitor the performance of steel box-girders. According to the field inspection in 2015, two typical types of cracking (types A and B) were detected at the hanger-togirder connections, and up to 50 cracks were detected in the anchor connections, of which 37 corresponded to type A and the rest were type B, as shown in Fig. 3. The type A cracks at the connection A started from the end of T-type fillet welds and propagated along the welds to the soldering holes, and some of them eventually propagated to 150 mm in length, as shown in Fig. 3a. The type B cracks at the connection B were observed in single fillet welds presented in Fig. 3b, which grew along the welds with crack lengths ranging from several to hundreds of millimetres. However, all type B cracks were arrested between two bearing plates due to the constraint of bearing plates. According to Fig. 3,

Figure 4a shows the total length of type A cracks in the upper and lower stream anchor connections detected in 2007, 2011 and 2015, being 1170, 3050 and 7170 mm, respectively. Although the propagation of an individual crack is highly uncertain from a probabilistic view, the crack propagation becomes faster with time. Figure 4b–d further illustrates the total length of each anchor connection along the bridge from north to south by the end of 2007, 2011 and 2015, respectively, where it is observed that premature type A cracks were found only several years after the bridge was open to traffic and gradually developed to an unexpected level in 10 years.

Although type A cracks were found dispersed over the complete bridge, more cracks were found at anchor connections near the one-eighth and

Mid-span

North Pylon 8000

7170

Lower stream Upper stream Sum

6000

4530

4000

3050

2420

2000

1640

1170

1120

630

50

0

2011

2007

2015

400 350 300 250 200 150 100 50 0

2007

10

20

400 350 300 250 200 150 100 50 0

North Pylon

South Pylon

Mid-span

2011

10

Lower stream Upper stream

20

30

40

50

30

40

50

60

70

80

90

Serial number of anchor locations

60

70

Serial number of anchor locations

80

(d) Crack length (mm)

(c)

South Pylon Lower stream Upper stream

Year of inspection

Crack length (mm)

Statistical Analyses of Type A Cracks

(b) Crack length (mm)

Total crack length (mm)

(a)

both types of cracks were initiating at the toe of the weld.

90

North Pylon 400 350 300 250 200 150 100 50 0

10

South Pylon

Mid-span

Lower stream Upper stream

2015

20

30

40

50

60

70

80

90

Serial number of anchor locations

F ig . 4: Total length of type A cracks in different years; (a) total crack lengths in different years, (b) in 2007, (c) in 2011 and (d) in 2015 346

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(b) 6000

4000 3000

4580 2950 2290

2000 1000

5245

Lower stream Upper stream Sum

5000

Crack length (mm)

Total crack length (mm)

(a)

690

660

660

665

2011

2015

30

2007

0

1400 1200 1000 800 600 400 200 0

North Pylon

10

Lower stream Upper stream

20

North Pylon

Mid-span

South Pylon

Lower stream Upper stream

2011

10

20

30

40

50

60

70

80

(d) Crack length (mm)

Crack length (mm)

1400 1200 1000 800 600 400 200 0

30

40

50

60

70

80

90

Serial number of anchor locations

Year of inspection

(c)

South Pylon

Mid-span

2007

1400 1200 1000 800 600 400 200 0

South Pylon

Mid-span

North Pylon

Lower stream Upper stream

2015

90

10

Serial number of anchor locations

20

30

40

50

60

70

80

90

Serial number of anchor locations

Fi g. 5: Total length of type B cracks in different years; (a) total crack lengths in different years, (b) in 2007, (c) in 2011 and (d) in 2015

one-fourth spans than the others.

Partly due to the existence of rubber bearings at girder ends and central clamps at the mid-span, no or few type A cracks were found in the anchor connections near the pylons and at mid-span. However, significant cracks were detected in the two anchor connections nearest to the mid-span (i.e. in anchor locations 45 and 47), as illustrated in Fig. 4b and c, which is probably due to the sudden change in stiffness because of the existence of central clamps. Besides, the type A cracks are asymmetrically distributed in both the upper and lower streams of the bridge. Traffic volume in the upper and lower streams of the bridge were obtained from the toll station, which respectively take up 52.39 and 47.61% of the total. Therefore, higher traffic volume might be one reason for more cracks being found in the lower stream than in the upper stream of the bridge.

in the lower stream and in the northern half-bridge than the other parts.

The asymmetrical distribution of type B cracks might be due to the different traffic volumes in the upper and lower streams. It is noted that no significant cracking can be found in the pylon up to one-fourth of the span in both the northern and southern half-bridges, and a similar phenomenon exists in most parts of the southern half-bridge.

FE Modelling and Analyses FE Modelling To investigate the mechanism of cracking in hanger-to-girder connections, a multiscale FE model of the RSB was developed using the FE programme ANSYS with the decks near the midspan refined twice, as shown in Fig. 6.

The steel box-girders were developed

Sub-model

Statistical Analyses of Type B Cracks For type B cracks, the total lengths recorded are 690, 2950 and 5245 mm in 2007, 2011 and 2015, respectively, as shown in Fig. 5. The average growth ratio of type B cracks was 565 mm/year for 2007–2011 and 573.75 mm/year for 2011–2015, which can be observed from Fig. 5a. According to the crack lengths of each connection presented in Fig. 5, the type B cracks appeared only a few years after the bridge was opened to traffic and are mainly near the one-fourth and the mid-spans, especially in anchor connections at the mid-span. Similar to the type A cracks, more cracks were found in the anchor connections

using the Shell63 element; the pylons and central clamps were built using the Beam4 element, having six degrees of freedom at each node, and the main cables and hangers were modelled using the tension-only Link10 element.

The initial strain/stress, material properties and real constants (i.e. areas of cross-section, moments of inertia) were strictly calculated and assigned to the corresponding elements. To reduce the number of elements, the orthotropic decks and bottom plates of the girder were modelled using a layer of plates without U-ribs, and these plates were assigned with orthotropic material properties.24 The contact between girders and the crossbeams of pylons were defined such that no penetration was allowed, while separation due to tension was allowed. Spring elements (i.e. Combin14 element) were used to

Anchor Structure

Box-girder

Connection A

Connection B

Global model

Fi g. 6: Finite element model of the RSB

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model the effect of the expansion joints installed at each end of the box-girder.

In total, there were 82 696 shell elements, 342 link elements, 444 beam elements and 24 contact elements in the global FE model.

1.35 m 120 kN 120 kN

1.9 m

60 kN

1.8 m

D=273.7 m

Deck centerline

Case 1

… …

4 m4 m 4 m 4 m

According to Saint Venant’s Principle,25 a sub-model of the boxgirder with a length of 19.32 m (longitudinal) and a width of 18.15 m (transversal) was developed, in which all the structural details were rigorously modelled, as shown in Fig. 6. To simulate the influence of vehicle loads on the connections A and B, fine mesh of size of 0.01 m was applied to the anchor connection and mesh size was increased gradually away from the studied connections in order to obtain a smooth transition between elements. The boundary conditions of the sub-model were obtained from the global model using the shell-solid sub-modelling technique, where the cut planes of the sub-model were perpendicular to the box-girder in the global model.

3.5 m

… …

6×16.1 m

Mid-span

… … … …

6×16.1m Pylon

1/4 span

F i g. 7 : Truck configuration and positions

Stress and Deformation Analyses The above FE model has been verified based on the measured modal and stress results obtained from the bridge completion test.26 For preliminary investigation of the stress levels of connections A and B, the load case 1 in the completion test, as shown in Fig. 7, was used for FE analyses. A total of 52 trucks (see Fig. 7) were symmetrically loaded about the deck centerline in four lines and 13 rows. Trucks were distributed near the one-fourth span of the bridge, where the distance between the middle axle of the truck in the seventh row and the mid-span was 273.7 m, as shown in Fig. 7.

According to the truck arrangement in load case 1, static loads are applied to the global model in Fig. 6. Nodal (a)

displacements in this load case are obtained as boundary conditions for the sub-model, based on which the FE analysis is carried out, and the stresses of the anchor connection shown in Fig. 8 are obtained. It is observed that significant stress concentration exists in the connections A and B. A closer view of the connection A (in Fig. 8b) indicates that the stress concentration around the middle bearing plate is more significant than that around the side bearing plate. Such phenomenon is consistent with the inspection results that for connection A more cracks were found at the middle bearing plate rather than at the side bearing plate. Moreover, stresses at the lower weld end of connection A are significantly larger than those at the upper

weld end. Accordingly, most cracks in connection A started at the lower end of welds. Considering that the stress is below the yield strength of the steel, and the failure was observed after several years of operation, the cracking is mainly due to the fatigue damage under repetitive traffic loads.

In order to investigate the stress concentration in connection A, Fig. 9a illustrates the x-components of top and bottom stresses at points M1–M9 and points R1–R9, where a local Cartesian coordinate is set. The x-direction corresponds to the longitudinal direction of the bridge. It is observed that the stresses at points M1–M9 vary from positive to negative, while those at points R1– R9 have an opposite changing pattern.

(b)

(c) B

B

A

A

0

von Mises (MPa)

von Mises (MPa)

10.9

25.5

40.1

47.4

65.5 0

29.6

69.0

10.8

148 177

F ig . 8: Stress isovalues in load case 1 (a) hanger-to-girder connection, (b) connection A and (c) connection B 348

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(b)

(a)

Stress σ x (MPa)

60

R9

M9

M1

40

y

M1

R1

x

20

Face Back

0 -20

Face Back

R1

Relative deformation (mm)

80 0.2

y

H15

H1

x H15

0.1

In-plane Out-of-plane H0

H15

0.0

H1

-40 0

20

40

60

80

100

120

140

160

0

100

200

300

400

500

600

700

Distance to y-axis (mm)

Distance to x-axis (mm)

Fi g. 9: Stresses σx and relative displacement in connection A (a) stresses σx and (b) relative deformation

stresses at a series of points in the inclined deck shown in Fig. 10a are obtained, with paths labelled as J60 –J–J6, J–P and Q40 –Q–Q4, respectively. As presented in Fig. 10b, the red line and blue line stand for the stresses at the top and bottom of the plate, respectively.

According to the variation of stresses along the paths J60 –J–J6 and Q40 –Q–Q4, a sharp change and an uneven stress distribution exist in the inclined deck at the ends of the gap (i.e. points J and Q). In addition, the top stresses of the plate vary from positive to negative along the path J to Q, while the variation trend is opposed in the bottom stresses of the plate. The maximum difference is up to

plates, resulting in significant displacement difference between the two bearing plates under loads. Because the two ends of the stiffening rib are welded to the middle and side bearing plates respectively, and the two bearing plates are of different stiffness, the in-plane shear/bending and out-of-plane bending deformations of the rib are somewhat inevitable. One improved measure to mitigate the undesirable deformation of the rib may be using a stiffening rib through the slots in three bearing plates.

As a result, connection A is subjected to bending moment and shear force. In addition, top stresses (i.e. stress on the upper surface of the plate) are all larger than bottom stresses (i.e. stress on the lower surface of the plate) by at least 8 MPa, indicating that out-ofplane bending exists in connection A. Figure 9b further describes the relative in-plane and out-of-plane displacements at points H1–H15, where in-plane shear/bending deformation and out-of-plane bending deformation are apparent. Being directly connected to the transverse diaphragm of the girder, the middle bearing plate has higher rigidity than the side bearing

To investigate the cracking at connection B, the x-components of

(a)

(b)

J

Inclined deck

K

4× 20 mm

×

5

Q4–Q1

y

Q

z x

Q1’–Q4’

J

J6–J1 6× 20 mm

M

J1’–J6’ 6× 20 mm

39 0 J6′

0

Auricular plate

O M

-39

K

Q P

N

J6

z y J3 J1 J1′ x J3′

J

L

Q2

P

N

Q1′

Q4′

Q1

Q2′ Q3′ Q

5

Q4

Q3

120 40 80 m) 0 s (m

i -40 Dista 10 15 20 -ax nce t -80 25 to x o y-a xis (m 30 35 -120 stance m) Di

Side bearing plate

(c)

Relative deformation (mm)

L

78

m m

Stress σx (MPa)

7

4× 20 mm

Top stress Bottom stress

(d) 0.035

J

K

0.030

Out-of-plane

L

Auricular plate

M

0.025 0.020

z x

0.015

J

0.010

N

Q

O

Inclined Gap deck

0.005

P Q

Inclined deck

0.000 0

5

10

15

20

25

30

35

Side bearing plate

Distance to z-axis (mm) Fi g. 10: Stresses σx, relative displacements and deformation pattern in connection B (a) serial number of points, (b) stresses σx, (c) serial number of points and (d) deformation pattern

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(a)

(b)

66

120

64

Lane 1 Lane 2 Lane 3

Lane 1 Lane 2 Lane 3

115

Stress (MPa)

Stress (MPa)

62

125

Connection B

Connection A

60 58

110 105 100

56

95

54 -80

-60

-40

-20

0

20

40

60

90 -80

80

-60

Distance to 1/4 span (m)

-40

-20

0

20

40

60

80

Distance to 1/4 span (m)

F ig . 11 : Stress–time histories under a truck running in various lanes for (a) connection A and (b) connection B

145.4 MPa in the top stresses and is up to 202.2 MPa in the bottom stresses. It is indicated that the inclined deck at the gap is subjected to significant back-bending deformation. Figure 10c further shows the relative displacements in the z-axis of the points J to Q, where the maximum relative out-of-plane deformation of the inclined deck is 0.033 mm in such a short gap (with a length of 35 mm). According to the deformation pattern of connection B shown in Fig. 10d, significant out-of-plane deformation exists in the inclined deck at the gap due to the displacement difference between the middle bearing plate and the side bearing plate. Although the relative deformations are <1 mm, such relative deformation in this gap can lead to excessive stress,27 which might be the main reason of the cracking in connection B. Such a phenomenon has also been observed previously at similar connections.28–30 Stress Range Assuming that a truck in Fig. 7 runs in different lanes at a constant speed across the bridge, static time history analyses are made to obtain the stress–time histories of different connections at the one-fourth span, as shown in Fig. 11, where the stresses are perpendicular to the crack. In general, the stress responses are similar, and significant stress ranges can be found under this passing truck, where Variable designation A

34,35

the maximum stress range is 9.63 MPa at connection A and 29.23 MPa at connection B. Although the shapes of these curves are similar at the same connection, the peak values are slightly larger when the truck is in the outer and middle lanes (i.e. Lanes 2 and 3) than in the inner lane (i.e. Lane 1). For example, when the truck is in different lanes, the peak stresses of connection B are 115.84, 118.60 and 119.81 MPa, respectively.

In addition, according to Fig. 11, the connections may be subject to alternating tensile stresses under the moving vehicle loads, resulting in significant stress ranges.

the bridge to obtain the stress influence lines of specific details, and the stress–time histories of a particular detail under various travelling vehicles are obtained by multiplying the influence line with axle weights and superposing them according to the axle spacing. Thereafter, stress ranges and the corresponding number of cycles Nc are achieved using rain-flow counting.33 The equivalent stress range Seq is calculated according to all stress ranges, as shown in Eq. (1): " #1=3 X ni ðΔSi Þ3 X ð1Þ Seq = ni

Probabilistic Fatigue Analysis

where ΔSi is the ith stress range, and ni is the corresponding number of cycle.

Analysis Approach As the fatigue lives of the specific details are by nature highly random, affected by a number of factors, a probabilistic analysis approach based on S–N curves is adopted in this study, which is in conjunction with a comprehensive vehicle load model developed using traffic information from a toll station.31 Random samples for the vehicle-load-related variables, such as transversal position of vehicles (NL), vehicle type (VT), axle load (AW) and axle spacing (AP) are generated using the Latin hypercube sampling.32 In this approach, a group of unit loads are applied to each lane respectively along

Meaning of variable Fatigue detail coefficient

Based on the probabilistic analysis approach, the fatigue lives of specific details can be expressed in terms of service year, as shown in Eq. (2): N 365 × ADT Nc A Δ =

m 365 × ADT Nc DLA Seq

T=

where N is the actual fatigue life in cycles calculated using the S–N curves of the detail, while T is the fatigue life (in years); ADT is the average daily traffic; A and Δ are the constants regarding certain detail category and

Distribution type Lognormal

ð2Þ

Mean value 3.61 × 10

12

3

MPa (Cat E)

COV 0.45

1.28 × 1012 MPa3 (Cat E’) Δ36

Miner’s critical damage accumulation index

Lognormal

1.0

0.3

DLA34

Dynamic load amplification

Normal

0.057

0.8

Table 1: Stochastic fatigue-related variables

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Normal PDF 0.08 μ = 10.99 Sre σ = 2.27 0.06 Sre γ = 0.413 0.04

Normal PDF μ = 12.97 Sre σ = 1.74 Sre γ = 0.587

0.6 0.4 0.2

0.02 9

10

11

12

13

14

15

0.0 16

0.08 0.07 0.06 0.05 0.04

1.0 FEA results Regression analysis Normal PDF μ = 24.97 Seq σ = 8.66 Seq γ = 0.535

0.6 0.4

0.02

0.2

0.01 0.00

10

15

20

25

30 35 Seq (MPa)

40

45

50

0.0

(d) 0.10

FEA results Regression analysis

FEA results

0.07

Regression analysis

0.06

Probability density

Probability density

0.8

0.03

Seq (MPa)

(c) 0.08

Normal PDF μ = 37.96 Seq σ = 4.77 Seq γ = 0.465

Cumulative distribution function

0.8

Cumulative distribution function

Probability density

0.10

0.00

(b) 0.09

1.0 FEA results Regression analysis

Probability density

(a) 0.12

Lognormal PDF µ T =10.01 σ T =6.6

0.05 0.04 0.03 0.02

0.08

Lognormal PDF 0.06

µ T =12.54 σ T =6.25

0.04 0.02

0.01 0.00

0

5

10

15

20

25

30

0.00

0

5

T (year)

10

15

20

25

30

T (year)

Fi g. 12: Probabilistic density functions of equivalent stress range and fatigue life (a) Seq of connection A, (b) Seq of connection B, (c) T of connection A and (d) T of connection B

the Miner’s critical damage accumulation index, respectively; and DLA is the dynamic load amplification. The slope of S–N curve on log–log axes, m, has a value of 3.0. Considering the high level of uncertainties in vehicle loads, material properties and prediction model, all variables are treated as random variables, with their probabilistic properties depicted in Table 1.

According to previous investigations, including field tests, laboratory experiments and simulation,37–39 the detail Categories E and E0 (with the constant amplitude fatigue thresholds of 30.8 and 17.8 MPa, respectively) are assumed for connections A and B, respectively, considering the influence of out-of-plane deformation.34

provide a desirable fit. Based on the distributions of Seq and Nc, the probabilistic fatigue lives of these connections are obtained using Eq. (2), and regression analysis is conducted to obtain the mean values and standard deviations through lognormal PDFs, as shown in Fig. 12c and d. It is observed that the mean lives of connections A and B are approximately only 10 and 12 years, respectively; however, the standard deviations are considerable, being 6.6 and 6.25 years, respectively, indicating that there are high possibilities for the connections to have a short life, which is consistent with the fielddetected cracking phenomenon.

Conclusions Results and Discussion According to the above analysis procedure, the distributions of Seq and Nc of the connections A and B are obtained, as shown in Fig. 12a and b, where the double-peak normal probability density functions (PDFs) are used to describe the distributions of Seq. According to the cumulative distribution function (CDF) of Seq, multi-peak normal distribution can

Hanger-to-girder connections are extremely critical force transmission components of cable-supported bridges.

This paper reports the field-observed cracking in hanger-to-girder connections, and both deterministic and probabilistic FE analyses were conducted to investigate the cracking mechanism.

According to the methodologies and assumptions adopted in this paper, the following conclusions can be drawn:

Structural Engineering International Nr. 3/2017

1. Unexpected cracking was observed at the connections between the bearing plate and stiffening rib and between the auricular plate and inclined deck, which occurred just a few years after the bridge was open to traffic. The anchor connections in the lower stream of the bridge in general are subjected to more severe damage than those in the upper stream of the bridge, which is mainly due to different traffic volumes on the upper and lower stream sides.

More type A cracks are found at anchor connections near one-eighth and one-fourth spans than the others, while type B cracks are mainly near the one-fourth and onehalf spans, especially in anchor connections near the mid-span.

2. According to the FE analysis, the two connections are subjected to significant stress concentration and large stress ranges. Connection A suffers from severe in-plane bending/shearing and out-of-plane deformation due to the displacement difference between two bearing plates with different stiffness, while connection B is subjected to significant out-of-plane deformation due to the existence of a gap.

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3. Probabilistic FE analyses further show that the mean lives of the two connections at one-fourth span are very short (approximately only 10– 12 years), as compared with the design life of the bridge (i.e. 100 years), and the lives might be even shorter considering the high uncertainties in the analyses and different locations along the bridge. The causes of such damage may include defective welding, irrational configuration and discontinuity of the details, etc.

4. In future studies, the steel plates can be instrumented with strain gages near the cracks to measure the real fatigue stresses onsite. The influence of such cracking on the performance of the anchors should be further discussed, and the connections can be optimised (i.e. using through stiffening rib, bearing plate without cope hole). The dynamic time–history analysis, rather than the static one, may be adopted, with the real section of the weld and its penetration degree in the local FE model taken into account so as to further increase the accuracy of the analysis.

Acknowledgements Support from the Natural Science Foundation of Jiangsu under Grant No. BK20130023, the Education Department of Jiangsu under Grant No. JHB2012-1, the Jiangsu Transportation Department under Grant No. 2014Y02, the Graduate Research Innovation Project of Jiangsu Province under Grant No. KYLX15-0087 and the Priority Academic Development Program of Jiangsu Higher Education Institutions under Grant No. CE02-1-17 is gratefully acknowledged.

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Structural Engineering International Nr. 3/2017

Investigating Structural Stability of Historic Buildings: A Case Study of a Rampart Wall of Hilandar Monastery c Tovarovi c, PhD, M.Sc., B.Sc. Arch, Asst. Prof.; Nenad Šekularac, PhD, M.Sc., B.Sc. Arch, Assoc. Prof.; Jasna Ciki Jelena Ivanovi c-Šekularac, PhD, M.Sc., B.Sc. Arch, Assoc. Prof., Department of Architectural Technologies, Faculty of Architecture, University of Belgrade, Belgrade, Serbia. Contact: jelenais@orion.rs DOI: 10.2749/101686617X14881937385250

Abstract

Literature Review

The objective of this research is to establish an appropriate methodology for the restoration of historic buildings and detecting the deformations of structural elements. The basic method adopted includes long-term in situ observation of historic buildings in use, analyses and condition assessment of the construction and performance monitoring for structural elements based on the case studies of Hilandar Monastery, Mount Athos, which is included in the UNESCO World Heritage List. The subject of this research is the rampart wall (10th century) and the buildings directly connected to the wall, i.e. the Refectory and the Abbot’s Lodge. This research deals with the causes of deformation and the subsequent consequences—the leaning of the rampart wall and deformation of the adjacent structural elements. The abovementioned motions and deformations led to the occurrence of cracking in the chimneys of the Abbot’s Lodge, where a disastrous fire (2004) destroyed nearly 55% of the monastery structures. The present research discovers the cause of the problem that has been present for more than two centuries and also prevents further deformations on the rampart wall.

The cultural heritage of a country is its legacy from the past and the natural and cultural values that are worthy of being preserved for future generations as authentically as possible.

Keywords: heritage conservation; UNESCO cultural heritage; structural stability of rampart wall; leaning of the rampart wall; cause of fire.

Introduction

The objectives of the research are to:

There is a need to implement a safeguard plan to manage rampart walls, historic structures and their remains against further deterioration. This paper investigates a method to increase the stability of the rampart wall of Hilandar Monastery on Mount Athos, Greece, and to prevent further leaning without negatively affecting the authentic visual appearance and materialisation in accordance with the conservation requirements.

• Undertake active investigation of the site to ensure that effective measures are taken for the protection and safeguarding of the rampart wall and historic structures as well as the unique spirit and historic quality of the entire complex of Hilandar Monastery • Increase awareness and knowledge of historic heritage and archaeological, architectural and historic values in order to fully develop the abilities necessary to carry out the restoration and preservation of the 1000 year old rampart wall and monastery complex and pass our legacy from the past on to future generations • Develop appropriate structural interventions to carry out rehabilitation and prevent further defects and leaning of the rampart wall and adjacent structures • Systemise the obtained facts about the causes of leaning and deformations of the adjacent structures and suggest necessary measures to restore and preserve these historic structures and prevent further deterioration • Make the results available to the academic community

The basic hypothesis of this paper is that appropriate rehabilitation and structural interventions, in compliance with conservation requirements, are capable of increasing the stability of the rampart wall, thus preventing further leaning, deformations and damage to adjacent structural elements and preserving the ancient appearance and harmony of the protected heritage site.

Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: July 24, 2016 Paper accepted: October 18, 2016

Structural Engineering International Nr. 3/2017

From the last century onwards, there has been a significant development in the preservation of architectural structures and their remains. The application of structural interventions on relics enhances the quality of their preservation for the future and renders them complete, thus safeguarding their historical importance, as presented in Ref. 1. All historic buildings have a set of typological, structural, constructional, functional, aesthetic, formal, historical and symbolic values, as given in Ref. 2. These elements reflect the way a conservation project is planned and a critical methodology to evaluate the project results as they are.

Mount Athos was included in UNESCO’s list of World Heritage Sites in1988.

As the objective of this research is a circuit wall, the following factors may be responsible for particular problems and deformations, as presented in Ref. 3: • Self-weight load • Earth pressure (depends on the quality and the compaction of artificial fillings, the excavations, the refilling compaction, the rainwater effects, the live load of the surface and the seismic activities) • Hydraulic action (weight increase caused by the degree of saturation, smoothened joints of the stone blocks, increased hydrostatic pressure due to the low permeability wall, loss of mass and natural degradation of the stone blocks) • Permanent load due to the construction of other structures on the wall itself • Military interventions (bombing, etc.)

• Seismic activity As a result of the mentioned factors, the rampart circuit wall, being the Scientific Paper

353

protected structure, shows a series of typical problems of a structural nature, which form its pathology.

These structural problems of the protected rampart circuit wall can be divided into three main categories, as discussed in Ref. 3: • Geometrical: related to the changes in the form of the structure, without structural defects such as displacement of structural members, their rotations, barrelling and vertical axis inclination (deviations in relation to vertical axis). These failures are caused by earth pressures and seismic activities.

• Mechanical: related to strains and failures of structural elements, without negatively affecting their form but still causing cracks, breaks, weathering and detachments of stones. All these structural problems are capable of creating local weakness of structural elements and general concern as to the capability of the structure (the wall) to respond to increased loading (such as hydrostatic pressure cause by heavy rains, earthquakes). These failures are the direct result of great compressive and/or tensile stresses on the wall mass due to various loadings.

• The consequences of unfavourable effects of natural physicochemical agents, such as cracks, loss of mass, soilification of the stone, mortar leach and others.

As for the restoration of cultural heritage, it is necessary to carry out geotechnical site investigations, especially in the area where the structural elements are exposed to soil-related effects (e.g. retaining the walls from ancient times), as presented in Ref. 4.

Failure of a retaining wall is a common structural occurrence that can cause severe damage—rotation, dislocation, disorganisation and fractures of the stone blocks. The main causes are excessive rainfall and earth pressures due to seismic activities.

•

•

•

•

•

issues and Hilandar Monastery and the analyses of previous research findings, data processing and systematisation in order to arrive at hypotheses.

In situ observation: protection and preservation of this historic structure is based on the conducted analysis and assessed state of the rampart wall and the adjacent Abbot’s Lodge.

Undertaking a case study research to investigate the rampart wall of the Hilandar Monastery, its leaning, the factors causing the problems and the impacts on the structural elements of the Abbot’s Lodge.

Study of the causal relationship between the current state of the site and the rampart wall and structural elements of the Abbot’s Lodge and the analysis and observation of the already-implemented construction interventions on the rampart wall in order to compare the obtained findings with facts from various literature.

Setting the criteria to guide rehabilitation and structural interventions considering the previously conducted in situ analysis and selected structural interventions that are to be taken in order to prevent further leaning of the rampart wall and structural failures of the adjacent buildings.

Implementation of structural protection measures and properly selected rehabilitation measures in accordance with the conservation requirements and preservation of the visual appearance of the

heritage site, its structural elements and materialisation.

• Design a rehabilitation project for the rampart wall in order to prevent further leaning and structural failures of the Abbot’s Lodge.

• Accessing the experience and knowledge gained from previous studies of restoration.

Case Study—Hilandar Monastery and Holy Mount Athos Site Specificities Mount Athos is located in the unapproachable peninsula of Athos in the Aegean Sea, Greece. The name Mount Athos comes from the Orthodox Christian world, when Athos became the spiritual capital of monastic life.

Hilandar Monastery is one of 20 monasteries on Mount Athos. It is ranked fourth in the hierarchical order and has the second largest territory. One of the most significant spiritual and cultural centres, it was founded in 1198 on the remains of the old and abandoned monastery Helandariy, built at the end of the 10th century, as presented in Ref. 5. The monastery was rebuilt in the 12th century by the Serbian noblemen, Saint Sava and his father Saint Simeon. A brotherhood of primarily Serbian monks has lived there ever since. The Hilandar Monastery is an isolated medieval town inside the walls (Fig. 1).

The reconstruction and rehabilitation of a structure in order to return it to its original form is inefficient for the longevity of the structure unless all the factors causing the problems are addressed.

Methodology The research methodology includes: • Analyses of the collected articles and literature related to these 354

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F i g. 1 : Hilandar Monastery, Mount Athos, Greece

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Mount Athos has a monastic community with a wide range of heritage elements that are imbued with intangible values (sacred, spiritual) and influenced by century-old traditions and special ways of life.

What makes the site of Hilandar Monastery and Mount Athos unique is the fact that the principles and standards of living have been unchanged for eight centuries. Therefore, the site itself is a living religious heritage.

The same, rigorous rules have been applied for centuries ever since the beginning—only men are permitted to enter the territory, and they first have to secure a special visa issued by the monks. Even today, Mount Athos is inaccessible by land, which means that one has to take a boat across the sea.

There is no road connection with the mainland. The secluded site and living in isolation have contributed to various challenges and difficulties in carrying out the restoration interventions.

The responsibility for cultural heritage management in the Athonite peninsula lies within each monastery (for its relevant territory) and with the collective administrative bodies situated in Karyes, the capital of Mount Athos.

Furthermore, a significant role in the implementation of various heritagerelated projects is held by the Hellenic Ministry of Culture through its peripheral agency, the 10th Ephorate of Byzantine Antiquities. The latter shares responsibilities, since the 1980s, with KEDAK (KEΔAK, the Centre for the Preservation of Athonite Heritage), a multidisciplinary body of academics, heritage professionals and Athonite representatives, as presented in Ref. 6.

The Research Results on the Rampart Wall of Hilandar Monastery Historical Review When Saint Sava and Saint Simeon came to Hilandar in the 12th century, the above rampart wall had already been there—it had been a barrier wall in the west built for defensive purposes by the Byzantine Empire. The rampart wall was made of roughly dressed stone (the bricks were laid in some parts), 1.60 m thick and 11.4 m high on the outside. Shallow pilasters line the exterior and interior walls.

Above the pilasters, there are shallow arches of brick. The width at the top

of the west rampart wall is 1.90 m, and there is a walking path in the central part at the top of the wall, where one can walk freely even today.

The external facade of the wall is arranged in a specific pattern as a defence mechanism. The rampart wall has no openings for safety reasons, meant to provide defence and protection to the monastery complex. The openings currently on the outside of the rampart wall were made later, which can be concluded from the observations of the stone processing around the openings.

The Refectory was built next to this wall in 1293, during the rule of King Milutin, as presented in Ref. 5. The roof rests on the rampart wall, where there is a walking path that is still usable today. There is a wine cellar underneath the Refectory, with the floor level in line with the ground outside the rampart wall (Fig. 2a and b).

There are two external three-sided apses on the south rampart wall of the Refectory and the north separating wall between the Refectory and the Abbot’s Lodge—the small apse is on the south side and the big apse is on the north side. The north wall of the Refectory touches the west rampart wall, without bonding of stone walls (Fig. 7). This confirms that the Refectory was built after the rampart wall.

On the north wall, the cornice above the top of the apse was built of bricks, and the remains of the upper side of the apse show that it used to be covered with stone.

The Abbot’s Lodge is located close to the north side of the Refectory, extending from the west rampart wall (Fig. 2a). The building was named after the Abbots—the heads of the monastery. The exact period of construction is unknown, but the inscriptions on the building, which were carved in stone or made of brick, reveal that the Abbot’s Lodge was rebuilt in 1651/1652, meaning that it had been built before 1650 and after the construction of the Refectory.

This is confirmed by the jagged cornice around the apse of the separating wall between the Refectory and the Abbot’s Lodge, as well as by the upper surface above the apse. It appears that this used to be the external wall during the construction of the Refectory. Another confirmation that it was an external wall is the walled window located on the north apse.

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Considering the east facade design, during the 1652 reconstruction, the building consisted of a basement, ground floor and two floors, as presented in Ref. 7. The building material was a combination of dressed stone and brick. Looking at the external west side of the monastery complex, confirms that the wall of the Abbot’s Lodge was erected on the rampart wall itself, and it was two or three levels high.

After the great fires that occurred in 1711 and 1775, as given in Ref. 8, when the entire structure of the Abbot’s Lodge was burned down, the new reconstruction began in 1777, and another floor was added—this is the final appearance of the building. The current level of the basement is 4.80 m above ground level on the outside of the rampart wall (Fig. 2c).

Considering the basement level of the two adjacent buildings, the Refectory and the Abbot’s Lodge, and the previous conclusion that the north wall of the Refectory used to be the external wall, it seems that, at the time of the construction of the rampart wall, the ground level was about the same on the inside and the outside, which means that the primary function of the wall was defensive. Given the different levels of the basement floors of the Refectory and the Abbot’s Lodge, it is evident that backfilling had been carried out, thus causing such differences (Fig. 2d). As stated in the literature, two fires that happened in the 18th century must have caused the collapse of the Abbot’s Lodge, the rubble fill and the formation of an embankment along the west side of the rampart wall, next to the north side of the wine cellar wall, inside the peripheral walls of the building.

The north direction of the west rampart wall ends with the perpendicular wall—the one that had been the rampart wall itself before the monastery was extended. In 1869, a counterfort was built on the north end of the rampart wall. It was 5.75 m wide on the outside. Based on the inscriptions carved in stone in 1896, three smaller counterforts were also built on the rampart wall in front of the Abbot’s Lodge—two of them were 1.8 m wide, and the third one was 2.4 m wide (Fig. 3). All three counterforts were made of stone, and all of them rested on the same footing. Nowhere in the literature one is able to find the Scientific Paper

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(a)

were made, primarily defining the foundation plans and visual appearance. The next decades were marked by the restoration interventions in several monastery buildings.

A

0 12 3 4 5

10 m

20 m

In March 2004, a devastating fire destroyed most of the monastery complex (Fig. 4). The fire broke out in the Abbot’s Lodge and spread over the entire monastery area. The fire caused massive damage and made impossible any in-depth investigation and analysis of the construction techniques and materials. The objective of the planned research was to develop appropriate rehabilitation and restoration measures. Urgent measures and structural interventions were required immediately after the inestimable damage to the buildings of the Hilandar Monastery, as presented in Ref. 9.

B

+47.90

+43.60

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+42.67 II

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The Refectory 220 mm 130 mm 20 mm 20 mm

I The Abbot’s Lodge

IV

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Deformations of the rempart wall

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50 mm

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h=11.50 m

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According to official records, the fire began to spread through the chimney crack of the Abbot’s Lodge, which initiated a comprehensive analysis of the current site condition in order to consider structural interventions as well as appropriate fire protection measures and eliminate the cause of such immense damage and cultural heritage loss.

h=4.5 m

h=9.90 m

245 mm 245 mm

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Cross section A-A

10 m

20 m

Cross section B-B

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+43.60

0 1 2 3 4 5

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Cross section C-C

F ig . 2: The Refectory and the Abbot’s Lodge built along the rampart wall: (a) basement floor plan, deformations of the rampart wall; (b) the Refectory—cross-section A–A; (c) the Abbot’s Lodge—cross-section B–B; (d) longitudinal section of the Abbot’s Lodge and the Refectory cross-section C–C

reason for these subsequently built counterforts in the 19th century.

Difference in ground level between interior rampart wall, in front of the Abbot’s Lodge, and exterior rampart wall is 5.25 m. (Fig. 2c).

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The Research and Its Results The second half of the 20th century saw the restoration of several buildings belonging to the Hilandar Monastery complex. It was then that the first technical drawings (drafting)

The analysis of technical documentation from the second part of the 20th century shows that the focus was primarily on foundation plans and visual appearances, while sections were inaccurate and rarely included in the drafting. Inaccurate cross-sections, in relation to the level next to the rampart wall, on the inside and outside, were caused by an inability to calculate height difference as the lower part of the rampart wall had no openings through which the calculation could have been performed. Also, the foundation plans showed different levels and semi-levels but without height levels.

The in situ observation and the rampart wall measurements that were carried out in 2004 brought to light a number of previously neglected facts.

Considering the observation of the Abbot’s Lodge, the Refectory and the level underneath these buildings—the basement level and the ground level on the other side of the rampart wall—it was clear that the obtained data did not match the previous technical documentation. The research was conducted on the principal cause of the cracks in the Refectory and the cracks on the arches perpendicular to

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Fi g. 3: The rampart wall with the counterforts and subsequently carried out interventions on the wall

it; archaeological conservation experts were ramped up and the obtained results showed that the foundation of the rampart wall was constructed on the ground. Based on the proximity to the rampart wall, the basement of the Refectory was likely constructed at the end of the 20th century. Due to the proximity of the stream, which is located at a distance of 10 m from the south end of the rampart wall, it was assumed that the cracks were caused by poor load bearing capacity of the soil, hence the subsidence of the rampart wall. The cause of the cracks appears to be poor bearing capacity of the soil, which led to the subsidence of the rampart wall.

After the 2004 fire, structural failures occurred in many places in Abbot’s Lodge, on the partition wall between the Refectory and the Abbot’s Lodge, next to the rampart wall, and at the points of separation between the arches that are perpendicular to the rampart wall. The following in situ investigations were started in 2006 to determine the causes of the structural failures, and continue to be conducted in 2016: • Geodetic survey: field survey, measurements of absolute horizontal movements of the benchmarks along the west rampart wall using the method of triangulation, observations of the subsidence and movements of the benchmarks.

• Geophysical survey.

• Geo-radar survey: using the GPR method and antennas of 25, 250, 800 MHz (Fig. 5a).

• Geological survey of five boreholes up to 15 m deep (Fig. 5b).

• Inclinometer pipes were inserted into three boreholes; the pipes are equipped with guides for

•

•

• •

• •

•

landslide monitoring in two directions (Fig. 5c).

• Piezometric pipes were inserted into the remaining two boreholes in order to monitor the level of ground water.

Measuring the verticality of the rampart wall on the outside of the pilasters using the surveying equipment.

Installation of triaxial dilatometers for monitoring the structural failures between the Refectory and the Abbot’s Lodge and the perpendicular wall between these two buildings (Fig. 5d).

Installation of fixed plumb bobs on the pilasters of the Refectory (Fig. 5e).

Measuring the changes in the distance between the benchmarks on the inner part of the walls (Fig. 5f ) using the convergence method.

Visual observation of the rampart wall and in situ comparison with the previous photo documentation.

Calculating height differences between the floor level next to the rampart wall, on the outside and on the inside of the Refectory and the Abbot’s Lodge using geodetic methods.

Opening the monitoring probes on the outside of the rampart wall and close to the counterforts, as well as in the basement of the Abbot’s Lodge (Fig. 6).

The results of the geological survey show that solid rock mass was found 5–7 m beneath the current ground level, along the rampart wall, on the outer side. The in situ investigations through the insertion of monitoring probes found that the foundation of the rampart wall of the Refectory and counterforts was established on the

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F i g. 4: The Abbot’s Lodge after the devastating fire (March 2004)

ground, while the structure of the Abbot’s Lodge was founded on the rock block.

The use of test pits to investigate subsurface conditions is not allowed on Mount Athos, not even in Hilandar Monastery, without active participation of archaeologists and without special permission issued by the KEDAK and their experts (Fig. 6).

Based on the results of the geological survey, it is evident that the embankment inside the Abbot’s Lodge is of a heterogeneous nature—in addition to the earth materials (sandy–clay dust particles) and fragments of broken stone (rubble), there are also ash particles, slag and the remains from the fire. Considering the level of bonding in relation to the embankment, there are areas that can be classified as having poor bonding qualities. These are the parts of the embankment that have an increased level of ash, slag and burnt remains. The secondary consolidation of the embankment, which is assumed to have occurred during the last 250 years, increased the values of bulk density, g = 19.5–20.5 kN m2.

The consolidation led to the total subsidence of the embankment. Height reduction was followed by lateral expansion and increased active pressure on the rampart wall.

Based on the research conducted on the rampart wall, neither subsidence nor deformations have occurred in the last 10 years.

Considering the in situ observation of the Refectory from the outside of the Scientific Paper

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(b)

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(e)

rampart wall, and the photo documentation completed in the 20th century, we can reach the conclusion that the joints had been filled with mortar several times—whenever the mortar fell out, the exposed joints had to be refilled with the new mortar (Fig. 3).

This is particularly visible in the case of the rampart wall, the part of the Refectory oriented towards the Abbot’s Lodge.

Checking the verticality of the rampart wall shows that the maximum deviation from the vertical direction is observed in the Abbot’s Lodge, between the three counterforts, at heights up to 245 mm and 9.9 m above the ground, which means that the inclined position of the upper point of the rampart wall is 2.47%; as for the part of the Refectory, at the first pilaster of up to 4.5 m height, the observed deformation is 2.66%, while at a height of 11.5 m, the total inclination is 220 mm, that is, 1.91% (Fig. 2a).

(f)

F ig . 5: Field research: (a) georadar investigation; (b) geological survey; (c) inclinometer measurements; (d) triaxial dilatometers; (e) fixed sinkers and (f) convergence investigation device

(a)

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F ig . 6: Monitoring probes inserted in the Abbot’s Lodge: (a) visual appearance and (b) archaeological layers 358

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Towards the end of the Refectory, the inclination disappears along the south wall. The performed calculations and the observation of the vertical plane revealed that the maximum concavity (curvature) was found at a height of 4.5 m above the ground level, which can be seen in the photographs and by conducting in situ observations.

Therefore, it is evident that the occurrence and development of this deformation caused the mortar to fall out of the joints, which led to periodical replastering. This explains the almost century-old photo documents in which the freshly filled joints are visibly different from the other parts of the wall.

The measurements found a vertical deformation (2 mm) of the wall between the Abbot’s Lodge and the Refectory in the direction of the Refectory. It was revealed a year after starting to take measurements, and so far, no other deformations have been detected. It appears that this deformation was caused by the fire, considering that the Abbot’s Lodge used to be uncovered at that time and filled with rubble to a height of up to 4.0 m (Fig. 4). Soon after the fire, a protective structure was built over the Abbot’s Lodge to protect the partition wall against dampness and to protect the extremely valuable frescos from the 14th century that cover the partition wall inside the Refectory.

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against the rock mass increase the deformation and rotation of the rampart wall due to the subsidence of the counterfort that is caused by the process of consolidation.

Fi g. 7: The consequence of the rampart wall leaning—the rampart wall separates itself from the boundary wall between the two structures, the Abbot’s Lodge and the Refectory

After the devastating fire and the collapse of all the floor structures and partition walls, a visible cracking developed between the rampart wall and the partition wall that separated the Refectory and the Abbot’s Lodge (Fig. 7).

Based on the analyses of the existing documents, various measurements and in situ observations and considering the undertaken restoration interventions on the rampart wall in the past, we can conclude that the subsequently built counterforts (the big counterfort was built on the former north end of the west wall from 1869, and three smaller counterforts were built in 1896) were caused by the rehabilitation interventions that were undertaken in order to prevent further deformations and rotation of the rampart wall, without identifying the cause of deformation.

The reason for building a big counterfort might be a large crack in the former north rampart wall of the monastery, caused by the embankment activity. These subsequent counterforts were founded shallow, compared with the solid rock mass, and their mutual footing is parallel to the rampart wall, which prevents measuring of the foundation depth for the rampart wall. These shallowfounded counterforts that do not lean

The cause of the deformation of the rampart wall is assumed to be the embankment within the Abbot’s Lodge. In accordance with the relevant historical facts, although the 1711 fire caused total destruction, not until 70 years later did the first rehabilitation and restoration interventions begin. This means that the rubble had been exposed to precipitation, deterioration and consolidation. The subsidence, rainfall and soil consolidation initiated lateral pressure on the rampart wall. Considering the construction time of the counterforts, it is evident that the deformation had existed even before the 19th century and that the deformation itself caused the construction of all four counterforts. The construction of three smaller counterforts in 1896 implies that the builders realised that the big counterfort had failed to prevent further deformations. Therefore, they decided to build three new counterforts.

The type of the deformation and the rotation of the wall indicate that the deformation was not caused by: • Translational or rotational soil movements as the wall is not displaced • Base sliding as the lower part of the rampart wall is not rotated • Soil subsidence as there are no cracks in the longitudinal wall This is about the deformation caused by the rotation of the wall around its horizontal axis, that is, the foundation of the rampart wall, due to the horizontal force of the embankment.

All the former sketches and crosssection drawings of Abbot’s Lodge and the Refectory overlooked the delevelling and provided no details in that field. The reason was the lack of thorough investigations and necessary drilling work for inserting monitoring probes, which had to be carried out under the supervision of the KEDAK. The opening of a monitoring probe, which was inserted beneath the basement of Abbot’s Lodge along the rampart wall, revealed that the rampart wall had been built in such a way that its facade had been directed towards the inside of the Abbot’s Lodge. This building technique

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implies that the wall used to be visible (Fig. 6a). In the test pit that was excavated along the rampart wall, there were two layers of ash at different depths, which is real material evidence of the fire (Fig. 6b). The holes in the wall were used to support the wooden beams—the floor structure joists that were subsequently built.

The test pit also contained the remains and fragments of the floor layers and floor structures, which confirms that certain parts of the burned structures had collapsed and remained buried at that place.

This evidence confirms that this part of the rampart wall used to be above the ground level. In other words—this wall was built for the purpose of defence; it was not meant to be a circular wall. It is assumed that the construction of the Refectory was soon followed by the construction of the Abbot’s Lodge along the rampart wall, and it was then that the rampart wall was turned into the perimeter wall of the Abbot’s Lodge.

After the fires and the collapse of the Abbot’s Lodge, the rubble and earth filling, one side of this wall remained behind the embankment, which turned the wall into the circuit wall and caused the occurrence of deformations that are typical of stone circuit walls.

Discussion All the structures presented in this case study have valuable architectural heritage. At the same time, these structures are functional buildings (or they used to be functional until the devastating fire) adapted to the needs of everyday monastic life, based on religion, spirituality and prayers.

When it comes to protected sites and the preservation or restoration of historic relics or “dead remains” or the preservation of the entire site, the usual procedures require expert advice and the suggested structural interventions must be approved by the wider community. Only then can the implementation of the measures be started. What makes this site different from others is the unique organisation of Mount Athos; under no circumstances could any potential structural intervention disrupt the harmony and peaceful life of this monastic community. All the measures that were to be taken to protect the site must comply with these rules.

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As the structures included in the case study are located in Greece, that is, the European Union territory, they are under the care and supervision of both Greece and the Republic of Serbia, considering the fact that the Serbian nobles founded the monastery and that most monks of the brotherhood are of Serbian descent.

All the decisions on the preservation and restoration of the protected structures are to be mutually agreed on between the Serbian and Greek expert teams working on the restoration of the monastery and approved by the Administration of the Hilandar Monastery.

The planning, design and management of the restoration work was carried out by both Serbian and Greek expert teams appointed by the Administration of the Hilandar Monastery. The monitoring of the conservation and restoration is performed by the experts from the KEDAK, from Thessaloniki.

The Administration of the monastery and KEDAK agree on the most important principle—it is necessary to preserve the authenticity and the cultural heritage of the site. Authenticity is a fundamental criterion defined by most conservation requirements, and preservation of authenticity is the goal of utmost importance in conservation practice. This is a true example of the Convention for the Protection of the Architectural Heritage—the Granada Convention.

The financing for the restoration of the Hilandar Monastery is provided

by Greece, the Republic of Serbia, numerous donors, Orthodox believers and admirers of Hilandar Monastery from all over the world.

The investigations conducted on the rampart wall were monitored and managed by the Expert Council for the Reconstruction of Holy Monastery Hilandar, with help from architects, surveyors, geophysicists and archaeologists. The case study of this 1000-year-old rampart wall required the application of protection measures to prevent further deterioration and leaning.

• identified the problem; • determined the measures to ensure efficient protection of certain structural elements or entire structures; • conducted numerous investigations and in situ observations together 360

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with his colleagues, architects, archaeologists and geologists from the Serbian team as well as with the colleagues from KEDAK; the in situ observation has lasted for more than a decade, and it is still going on for the purpose of monitoring and detecting potential further leaning and deformations of the rampart wall; • insisted on creating technical documentation (pre-engineering) as there was none; the technical documentation, accurate cross sections and longitudinal sections enable a precise and logical understanding of the current situation, thus reaching logical conclusions on the required structural interventions in order to prevent further deformation and deterioration and protect the valuable cultural heritage; • came to a conclusion that was totally different from the previous opinions given by his colleagues, geologists who claimed that the deformations of the wall were caused by land sliding, that is, foundation subsidence; • proved that the existing embankment and permanent dampness increase the pressure that the rabble exerts on the rampart wall, thus causing the deformations that has been there for centuries, weakening the stability of the wall.

The study confirmed the theory developed in Ref. 4 that based on the detected and analysed factors affecting the site and its development, this restoration is considered to be a complex and difficult technical and scientific problem. As for the analysis of the structural problems regarding certain sites, all the planned interventions must be thoroughly tested and comply with the principle of broad interdisciplinary collaboration.

The study of the rampart wall of Hilandar Monastery partially confirms the opinion presented in Ref. 4, but it has to be pointed out that this wall has changed its function over the centuries—the original rampart wall was partially turned into a circuit wall due to the backfilling caused by the collapse of the structures after the fire, rubble filling and earth filling.

The study confirmed the opinion that earth pressure caused by filling and water effects can cause problems and deformations in the circuit wall. This study also confirms that seismic activity can cause certain changes in the circuit

wall, and Mount Athos is a highly seismic region. The results of these investigations confirmed the findings presented in Ref. 4, that all of the above factors cause a number of typical structural problems in relation to the rampart wall of Hilandar Monastery, which is part of a monument under the protection of UNESCO. As stated in the literature, the earthquake caused the occurrence of cracking, but the opening of the crack was not entirely caused by horizontal seismic activity— it was also caused by embankment displacement and horizontal earthquake effects on the rampart wall.

The authors confirmed the opinion of Egglezos et al., presented in Ref. 4 regarding the importance of forming an interdisciplinary team to deal with the design of case studies (in-depth investigations) on restoration, rehabilitation and preservation of cultural heritage because it is the only way to obtain the best results. The team should consist of structural engineers, archaeologists, conservators, geodetic engineers and geologists, and the team would deal with detailed and accurate analyses of the effects of soil composition and soil moisture on deformations in circuit walls.

Based on the findings of in situ observations and the analysis of all the mentioned factors, it was concluded that the theory of deformation—rotation of the rampart wall did cause other structural cracks of the Abbot’s Lodge. Since the Abbot’s Lodge has had the rooms (kelions) for the monks’ dwelling for the last 230 years, its technical maintenance is still carried out on a regular basis—minor cracks in the partition walls inside the building are regularly repaired. The mentioned continuous maintenance keeps newly appeared cracks constantly covered; therefore, this problem could not have been clearly depicted and analysed.

The experts dealing with this type of structural failures found that on the night between 3 and 4 March 2014, the fire started in one of the chimney cracks of the Abbot’s Lodge and spread over the monastery complex, causing the collapse of most structures. The unfortunate event caused disastrous damage to the archaeological, cultural and architectural values and irreparable loss to the world heritage. It was the fundamental motive and reason for this research.

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Currently, it is not possible to say whether the 1711 fire had been caused by the deformations and cracks in the chimneys of this building that consists of numerous kelions for the monks who used the fire and chimneys during the winter months. However, the rubble layers must have caused the deformations of the rampart wall. The exact time of the embankment construction is still unknown, but its destructive effects are evident.

All the above evidence and the resulting conclusions reached through the research confirm the hypothesis that appropriate rehabilitation and structural interventions, in compliance with the conservation requirements, are able to eliminate the causes of deformations of the rampart wall and prevent its further leaning by increasing the stability, thus preventing the occurrence of new deformations of the adjacent structural elements. The implementation of the above rehabilitation measures enables the preservation of the authenticity of the rampart wall and surrounding structures without disrupting the harmony of this unique protected site in Mount Athos.

Further research in this field will include archaeological excavations and insertion of monitoring probes in order to check and confirm the conclusions reached on the cause of the deformations. If these investigations uncover the remains of the floor layers and the place where the wooden floor structure and girders used to rest, the conclusions will be confirmed; this could lead to the removal of the embankment beneath the Abbot’s Lodge basement, thus permanently eliminating the cause of deformations of the rampart wall.

Any structural intervention to be undertaken will have to be carried out in accordance with the conservation requirements. Also, all the surrounding structures (the Refectory and the Residence from 1640) will have to be protected during the restoration.

Conclusions All the conducted in situ observations, surveys and analyses of the structural elements of the Hilandar Monastery and their mutual effects led to the conclusion that the centuries-long effect of the embankment beneath the Abbot’s Lodge basement is the cause

of the deformations of the exposed parts of the rampart wall.

Based on the research and the above evidence presented in this paper, the authors came to the conclusion that the only way to prevent further deformations of the rampart wall and preserve its current condition (including all the existing deformations) is to take the following measures: • Thorough archaeological research involving large number of test pit excavations in order to obtain necessary evidence and confirm that there is a layer of rubble few metres thick beneath the basement of the Abbot’s Lodge.

• Removal of all rubble layers formed after numerous devastating fires caused the collapse of structural elements; also, the rubble layers might have been the result of the fillings carried out beneath the basement of the building, which has been exerting enormous pressure on the rampart wall for centuries, thus causing the occurrence of various deformations.

• All the existing foundations of the counterforts built along the rampart wall are to be supported by solid rock mass through underpinning or mega pipe installation, thus eliminating the pressure that is currently exerted on the rampart wall; due to the lower bearing capacity of the soil supporting them, the consolidation of the soil beneath them and the fact that their foundations are worse compared with the rampart wall, they are constantly exerting additional pressure on the rampart wall, thus weakening its stability.

• Protect the wall of the Abbot’s Lodge towards the porte; once the embankment is removed, the pressure on the rampart wall is reduced, which means that the parallel wall of the Abbot’s Lodge becomes the retaining wall that is exposed to the effects of soil structure and the embankment from the porte (the courtyard).

• Implementation of the above structural interventions, in accordance with the conservation requirements, can eliminate the causes of deformations of the rampart wall, prevent its further leaning and increase its stability, thus preventing the occurrence of new deformations in the adjacent structural elements.

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• Implementation of appropriate rehabilitation measures and compliance with the conservation requirements ensure the preservation of the authenticity of the rampart wall and the surrounding structures as well.

The main conclusion reached in the study is that the removal of the embankment layers beneath the basement of the Abbot’s Lodge will eliminate the cause of the deformations in the 1000 year old rampart wall in Hilandar Monastery. The greatest contribution of this study is the discovery of the cause of deformations of the west rampart wall in Hilandar Monastery and the recommendations for eliminating the cause of this problem in compliance with the conservation requirements and preservation of the authentic visual appearance of all the structures and the rampart wall itself.

References [1] Lambrinou L. Preserving a monument: the example of the Parthenon. Conserv. Manage.

Archaeol. Site 2010; 12(1): 60–74.

[2] Gómez Robles L. A methodological approach towards conservation. Conserv. Manage. Archaeol. Site 2010; 12(2): 146–169.

[3] Egglezos D, Ioannidou M, Moullou D, Kalogeras I. Geotechnical Issues of the Athenian Acropolis, Geotechnics and Heritage CRC Press, Taylor Francis Group: London, 2013; 13–48.

[4] Egglezos D, Moullou D, Ioannidou M. The role of the geotechnical engineer in archaeological work: the Greek experience. Proceedings of the Second International Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites, Napoli, Italy, 2013.

[5] Bogdanovi c D, Đuri c J, Medakovi c D. Hilandar, Beograd, Srbija: Republi cki zavod za zaštitu spomenika kulture Srbije, Jugoslovenska revija, Vuk Karadži c, 1978; 26.

[6] Alexopoulos G. Management of living religious heritage: who sets the agenda? The case of the monastic community of Mount Athos.

Conserv. Manage. Archaeol. Site 2013; 15(1): 59–75.

[7] Nenadovi c S. Osam Vekova Hilandara – Građenje i Građevine Republi cki Zavod za Zaštitu Spomenika Kulture: Beograd, 1997; 279–296.

[8] Raki c Z. Crkve Svetog Dimitrija i Svetog Save Srpskog u Hilandaru Matica Srpska: Novi Sad, 2008.

[9] Šekularac N, Ivanovi c-Šekularac J. Protection from the effect of horizontal forces of remains of high walls within Hilandar Monastery Block, on Holy Mount in Greece. Proceedings of the 5th International Conference on Structural Analysis of Historical Constructions, Possibilities of Numerical and Experimental Techniques, New Delhi, India, 2007.

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Pathology, Assessment and Treatment of Structures Affected by Delayed Ettringite Formation Bruno Godart, Civil Eng., Université Paris-Est, IFSTTAR, Marne la Vallée, France. Contact: bruno.godart@ifsttar.fr DOI: 10.2749/101686617X14881932436771

Abstract Expansive internal sulphate reaction of concrete due to delayed ettringite formation (DEF) can damage concrete structures severely. DEF is defined as the formation of ettringite in a concrete after setting, and without any external sulphate supply but with water supply. This phenomenon occurs in concretes exposed to frequent humidity or contact with water, and subjected to a relatively high thermal treatment (>65 C) or having reached high temperatures during casting. This reaction may cause large structural disorders due to unexpected deformations and additional stresses in concrete and reinforcement, and therefore knowledge about how to deal with such structures affected by DEF is essential. After a presentation of the disorders due to DEF and an analysis of their causes based on cases encountered in practice, the paper presents the three steps of the appraisal methodology by developing the assessment based on numerical modelling, and reviews the various treatment solutions that may be implemented to deal with existing affected structures whose serviceability and structural safety may constitute a serious concern.

Keywords: concrete structure; delayed ettringite formation; disorder; appraisal; treatment.

Introduction The expansive internal sulphate reaction due to delayed ettringite formation (DEF) can damage concrete structures severely. In healthy concretes, the primary ettringite (a hydrous calcium trisulphoaluminate) is a normal reaction product formed from the reaction of C3A and C4AF with gypsum during the plastic stage of the hydration of Portland cement. But, when temperatures in concrete are above ~65 C during construction, the sulphates may be incorporated in other cement phases. After concrete hardening, very slow formation of higher volume secondary ettringite may occur as the water is incorporated in the ettringite crystal structure which can lead to potentially disruptive expansions.1 DEF is therefore defined as the formation of ettringite in concrete after setting, and without any external sulphate supply, but with a water supply. DEF appears in concretes exposed to water or to frequent humidity (at least over 90 to Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: July 20, 2016 Paper accepted: November 28, 2016

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95 % relative humidity), and subjected to a relatively high thermal treatment (>65 C) or having reached equivalent temperatures for other reasons (massive cast-in-place concrete, concrete casting during summer, etc).

The expansive process leads to macroscopic local effects similar to those created by alkali–aggregate reaction (AAR) and mainly consisting of material swelling and micro-cracking and reduction of the mechanical properties which may cause large structural disorders due to unexpected deformations and additional stresses in concrete and reinforcement. The micro-cracking is principally present at the interfaces between the aggregates and the cement paste and is also disseminated in the cement paste. These microcracks whose size is around 10–20 microns are filled with a compact ettringite. The ettringite that is generally observed in the concrete has a wide range of textures, but it is mainly poorly crystallised and massive.1 The swelling of the material may cause large structural disorders due to unexpected deformations and additional stresses in concrete and reinforcement, and the problem is therefore to know how to deal with such structures affected by DEF. After a brief review of some cases

encountered abroad and in France, and a presentation of the disorders due to DEF with an analysis of their causes, the paper presents the three steps of the assessment methodology and develops the assessment based on numerical modelling; then, it reviews the various treatment solutions that may be implemented to deal with existing affected structures whose serviceability and structural safety may constitute a serious concern.

Brief Review of Some Cases Encountered The first reported cases of DEF occurred in some precast concrete elements subjected to heat treatment unsuited to the composition and environment of the concrete. The main examples of DEF world-wide include railway sleepers,2–9 and massive castin-place concrete components.10–13 In France, the first case of DEF was observed in 1997 in a cap beam resting on two circular columns and supporting the deck of the Ondes Bridge.14 Then other bridges with DEF were discovered; the bridge parts damaged by DEF were primarily massive structural elements (piers, cap beams on piers or abutments, etc.) in contact with water or subjected to high moisture. A brief description of these bridges may be found in Ref. [15] and a more detailed description is presented in Ref. [16].

More recently, DEF was discovered in some precast and prestressed concrete beams of about 25 bridges in France. These bridges were built around the year 1995, and DEF occurred only in the edge beams that are particularly exposed to high humidity (rain). The cracking on the external face of the edge beams is mainly longitudinal and relatively important, particularly at the ends.

Cracking on the internal face of the edge beams is much less important, and all internal beams are free of cracks. The investigations conducted on these bridges showed that the damaged precast beams were

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subjected to a severe heat treatment during their fabrication.

Presentation of Disorders The structures damaged by DEF in France are mostly civil engineering works, while buildings appear, at first glance, to be affected only slightly.

This reaction is mainly found in bridges and in some dams. So far, most of the bridges damaged by DEF were built after 1980 and before the adoption of preventive measures in 2007.17,18 The disorders presented are similar to those observed on structures damaged

by AAR, except for pop-outs and exudation of gel.19 It shows that map cracking is the most frequent disorder observed on the faces of structures affected by DEF. Cracking is generally anarchic and can take the form of a crazing with small mesh size (20– 50 mm) and a rather small crack depth (20 to 50 mm), or take the shape of a larger crack network (300 to 400 mm size) with greater crack depth (greater than 100 mm) (Fig. 1).

The crack opening is variable in each observed zone according to the evolution rate of the reaction. It can be a few tenths of millimetres for a crazing, and can reach several millimetres for a wide mesh cracking. The crack depth also varies with the degree of evolution of the disorders: cracking may be superficial (a few centimetres deep) or can propagate in depth, for a through cracking.

In some structures, map cracking leaves room for an oriented cracking.

This happens in the case of a structural element where there is compression stress in a predominant direction, and the cracks open in the direction perpendicular to the main compression axis. This is particularly the case when horizontal cracks develop in prestressed precast concrete beams (Fig. 2), or when vertical cracks occur in columns or piers.

The expansion of the structure is generally associated with a change in the mechanical properties of the concrete material resulting from the formation of micro-cracks inside the material. In DEF cases, at the beginning of expansion (say, 0.2–0.3%), micro-cracking first reduces the tension strength of concrete and its Young modulus.

Then, significant reductions in compressive strength can be found for

Fi g . 1: Cracking of the base of a pylon (Bourgogne bridge: crack width >0.3 mm)

Fi g . 2: Longitudinal cracks in the middle of a prestressed precast concrete beam

high and very high expansion levels (e.g. an expansion of 1%).20 The swelling of the material leads to significant internal stresses within the structure that are superimposed on existing constraints, and causes overstress in the reinforcing and prestressing steels. This leads to an expansion of the structure, as well as formation of cracks that occur when the stress state resulting from permanent loads, live loads and swelling of the material exceeds the tensile strength of the concrete. It is thus possible to observe cracks in theoretically compression areas, and there is usually a network of cracks that are more open on the bridge facings. This latter observation is probably due to differential swelling between the surface and the centre of the structure. The most credible assumption is that the centre expands more than the surface, because the temperature at construction is higher in the centre than at the surface and also because gradients of moisture exist within the concrete

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elements. Other factors such as a decrease in pH caused by carbonation of the surface concrete and leaching of surface alkalis also have to be considered.

Causes of Disorders In Ref. [16], a comparative study was done of the factors that could be encountered in several investigated bridges. Table 1 is a synthesis extracted from this report, and shows some common features of the investigated bridges that were damaged by DEF: a massive structure, exposure to water or rain, casting in summer, a maximum temperature inside concrete above 75 C (except for one) inside the concrete, a high cement content (above 380 kg/m3) and a rather high content in sulphates (above 2.5% of SO3) and aluminates (above 7% of C3A).

There are several reasons that explain the occurrence of these disorders.

Among them are the popular trend to increase the performance of concrete, Scientific Paper

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Bridge name

Ondes

Bourgogne

Lodève

Bellevue

Beynost

Cheviré

Date of construction

1955

1990

1980

1988

1982

1988/1989

Casting of concrete

August

August/ September

August/ September

August/ September

August

July/August

Cap beam

Base of pylon

Cap beam

Pier

Cap beam

Base of pier

27

6

9

10

10

8

Water-proofing problem

Immersed and variable immersion

Lack of drainage

Immersed and variable immersion

Exposure to rains

Rains and capillarity

Tmax ( C)

80

79

80

80

69

75

W/C ratio

0.50

0.45

0.47

0.54

0.49

0.48

CEM I

CEM I

CEM I

CEM II/A

CEM I

CEM II/A

Structural part Delay in occurrence of damage (years) Environment

Nature of cement 3

Cement content (kg/m )

430

400

400

380

350

385

SO3 content (%)

2.5

2.8

2.6

2.5

3.4

2.5

C3A content (%)

11.2

8.2

9.8

7.0

10.4

7.0

Table 1: Comparative study of factors encountered in some investigated bridges

with the development of (very) high performance concrete, and the tendency to increase the speed of erection of structures. Another additional reason is the increase in the size of the structural elements permitted by heavier manufacturing equipments and audacious designs. This has led to the development of high-performance cement resulting in an increase in the hydration heat. Consequently when some basic rules to control the heat release in structures are neglected, the temperature inside the structural elements may exceed the thresholds and leads to DEF. An analysis of the bridges affected by DEF shows that the structures or parts of structures in constant contact with water or wet soils, and parts constantly sprayed with water (e.g. due to a leakage of expansion joints) are the most damaged. This analysis confirms the key role played by water in the development of this pathology.

Until now, it was admitted that structural parts exposed to humidity should be subjected to an internal relative humidity of 90–95% to show distress.

A recent experimental study conducted at IFSTTAR21 on cylinders with 110 mm diameter and 220 mm height concluded that specimens immersed in water as well as specimens stored at 100% relative humidity (RH) develop an expansion very quickly (maximal expansion 1.4 %), specimens stored at 98% RH show an expansion after a long period (600 days), and specimens stored at 96, 94 and 91% RH do not expand.

Based on this study, it appears that 364

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the threshold of internal humidity to develop expansion is around 98% RH, and that it is influenced by the leaching of alkalis on the surface.

Some Other Considerations Related to DEF A research conducted in France21 showed that a concrete specimen exposed initially to a relative humidity of 90% without expansion, and then to an immersion, expands at the same rate had it been immersed since the beginning; this means that the expansion potential is latent in a structure and that any unfavourable change in the exposure condition can trigger the reaction.

Another critical problem may arise due to this potential latent DEF expansion in a structure: a study conducted at IFSTTAR22 confirmed that concrete specimen subjected to different thermal scenario after 100 days of maturation, showed the risk of a significant expansion associated with DEF for a late heat treatment on concrete. Moreover, a pessimum effect with the heating duration has also been highlighted. Therefore, structures heat-cured after certain storage duration, or exposed to a second heat treatment, can develop a significant expansion, and this should be carefully considered, e.g. in the case of radioactive waste concrete containers.

However, it is reassuring for those involved in construction that the survey of the 25 bridges damaged by

DEF in France shows a great many of them presenting a slowdown in the evolution of cracks 10 years after their occurrence, while in only four bridges cracking is shown to continue to be growing.

Assessment of Structures Three Steps of the Assessment Methodology Once the presence of significant DEF in the concrete has been established on samples taken from the structure with the help of laboratory investigations—among them the most valuable is the scanning electronic microscopy examination—the next action is to analyse the implications of the problem as regards the overall safety and the management of the structure. Following the methodology developed in Ref. [23] for AAR, the assessment of structures affected by DEF may be broken down into three steps: the initial appraisal, the consideration of severity of reaction and the structural assessment.

The first step of appraisal begins with a review of the available documentation on the design and construction of the structure (inspection records, any modifications and/or repair measures).

The extent to which configuration of reinforcement and prestress contains expansion in all three directions determines the vulnerability of the structure to DEF, so detailed reinforcement drawings and prestressing tendon schematics are paramount.

The appraisal continues with site investigations such as the

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measurement of the cracking index to estimate expansion, the measurement of global deformation of structures with time and the determination of the moisture levels in the concrete.

Then this appraisal proceeds with laboratory investigations: various mechanical testing and residual expansion testing on cores (like the one developed by LCPC24) for estimating the potential magnitude of future concrete expansion.

In the second step, the qualitative assessment of the reaction severity is done by crossing site investigations (e.g. number of cracks, expansion, water availability) with laboratory investigations (e.g. microscopy results, mechanical reduction, core expansion). According to Ref. [23], this second step may be implemented by distributing the structures among the four classes of severity defined by LL, LH, HL and HH (see Table 2), assuming that the structural elements are well reinforced and there are no other structural or deterioration concerns. In order to move towards a quantitative assessment, some precise advices may be given where the results of the laboratory and site investigations have to be considered as low or high. In terms of site investigations, it may be considered that a part of a reinforced concrete structure having a cracking index above 1 or 2 mm/m or a single crack greater than 0.5 mm is classified as high; concerning the water availability, an immersion in water of a part of a structure has to be considered as a more aggravating factor than an exposure of the same part to alternative rain and drying conditions which is in itself more detrimental than an exposure to a humid environment. In terms of laboratory investigations, a severe microcracking of the observed samples under a microscope, or a reduction of 5–10% of the compressive strength of the concrete, or a residual expansion of concrete (measured according to Ref. [24]) greater than 0.05% has to be considered as high.

In the LL class are found structures where the diagnosed DEF is too little to cause significant damage in the structure. In this case, it is advised to practise the regular inspection regime of the structure in order to detect an eventual development of cracking that could merit a reassessment.

In the LH class, two cases are possible: either the sampling programme may have missed the locations of the severe disorders found in the in situ investigation, or the disorders observed during on-site investigation could have other causes. In both cases, a new investigation program is required. As the in situ inspection did find major deterioration features, monitoring on the evolution of the structural disorders is strongly advised.

In the HL class, either DEF is only a localised problem in the sampled part of the structure, or the laboratory results show the early stages of DEF which have yet to develop significant disorders in the structure. In this case, additional awareness during the regular inspection regime of the structure has to be applied in order to detect the development of cracking that could trigger a reassessment.

In the HH class, structures are clearly affected by DEF. Monitoring of the structures alone is not enough anymore and a structural assessment to identify vulnerable details is strongly advised, as well as an investigation program to predict the future development of the DEF. The third and last step of the methodology is devoted to structural assessment and is applied to critical elements of the structures classified in the highest reaction severity class HH. Its objective is to identify significant current and future serviceability and safety issues that have an impact on the function of a structure. This structural assessment comprises a “structural severity rating” of each structural element and detailed assessment of those critical elements identified and the overall behaviour of the Site investigation (# cracks, expansion, water availability)

Laboratory investigation (microscopy results, mechanical reduction, core expansion)

Low

High

Low

LL

LH

High

HL

HH

Table 2: Summary of reaction severity assessment based on site and laboratory investigations (from Ref. [23]).

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structure. The following section emphasises on the full structural analysis based on numerical modelling.

Structural Assessment Based on Numerical Modelling The structural analysis of critical elements is mainly based on a calculation involving inevitably a numerical modelling that can also be used to analyse the global behaviour of a structure.

In this analysis, the consequences of a local failure of part of the structure need to be considered. Numerical modelling can also provide a long-term management tool and allows evaluation of possible strengthening and moisture mitigation techniques.

Modelling DEF involves real structures with complex geometries and details, mechanical aspects such as the initial stress state due to loading conditions, shrinkage and creep, physical processes including temperature and moisture effects, chemical reaction rates causing swelling, and other degradation mechanisms that may also be present in the structure. As in the case of AAR, the best model is based on a finite element method (FEM) and considers thermo-hydro-chemo-mechanical couplings. To complicate things further, the determination of the model parameters requires extensive and longterm experimental tests. Due to this complexity, any simulation model should be calibrated based on the monitoring data and pertinent information obtained from in situ and laboratory investigation of the structure investigated.

One of the first comprehensive models used to assess DEF-affected concrete structures has been developed at IFSTTAR. Following the theoretical developments proposed in Ref. [25], the total deformation of concrete is the summation of the elastic deformation, the plastic deformation, the deformation due to shrinkage and creep and the chemical-induced deformation. The chemically induced part of the deformation is derived from Larive’s law26 successively improved in Ref. [27] and is based on Eq. (1).

In this equation, ε∞ corresponds to the final expansion, τc and τl are the characteristic and the latency times, respectively (corresponding to the swelling rate and the duration before the onset of expansion), ϕ and δ are two complementary parameters introduced by Ref. [27] into Larive’s law to Scientific Paper

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model an asymptotically linear phase at the end of the reaction which seems to be more relevant for DEF (Fig. 3).

(

1− e

1 ϕ

1− δ+t

−

1,0

− τt

C

ðt − τ L Þ τC

for AAR for DEF

with 0 ≤ ϕ ≤ δ

0,8

)

ð1Þ

Expansion (%)

εðtÞ = ε ∞

1 −e

Model of expansion 1,2

0,6

0,4

The chemo-mechanical model shall also take into consideration the consequences of the stiffness reduction of the concrete by introducing a damage coefficient, and the effect of anisotropy; this latter effect has two origins, one linked to the casting direction of concrete in the structure, and the other resulting from the restrained expansion by the stress state.

After calculation, the model is able to give displacements, strains and stresses at every point of the structure. It is therefore possible to check if the ultimate and service limit states are attained or not, keeping in mind that uncertainties in both loads and resistance parameters remain a difficulty for developing a suitable safety format in the context of applying a non-linear FEM analysis for the assessment of existing structures. The model has beneficial aspects that also check whether some bars have yielded, 366

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0,2

0 0

100

200

300

400

500

600

700

800

900

1000

Time (days)

F i g. 3 : Fitting of the Brunetaud’s law (pink) with experimental points (blue diamonds)

ɛ(T°,h)

Expansion

P1 n°1 ɛc(t0) ?

Residual expansion test (T = 20°C ; RH = 100%) t0

t

Time

F i g. 4 : Fitting of the Brunetaud’s law with the result of an expansion test 1,8 1,6 1,4 Déformation (mm/m)

This model can then be implemented in FEM software (RGIB module).

The chemo-mechanical computation is based on the results of two complementary modelling: one aimed at assessing the temperature field in the structure, and the other one consisting in solving the transient non-linear moisture-diffusion equation governing the evolution of humidity in the porous network of the structure. Fitting of Brunetaud’s law is based on experimental data such as the results of the residual expansion test on concrete cores extracted from the structure and the results of the cracking measurement and the global deformation monitoring of the structure.

Figure 4 shows an example of a fitting with the result of a residual expansion test applied on a core extracted from the bridge pier presented thereafter.

Once the expansion law of the concrete has been calibrated, the calculation is conducted and its result validated with the value of a cracking index measured on the pier (Fig. 5).

Measure of the cracking index IF

1,2 z

1

x

y

0,8 IF

0,6

Cas1

0,4

Cas2

0,2

Cas4

Cas3 Cas5

0

0

50

100

150

200

250

300

Time (months)

F i g. 5 : Fitting of the model with the result of the measurement of a cracking index IF on the pier (blue diamonds) – the curve “Cas5” appears to be the best fitted curve

whether cracking is developing, whether concrete is able to sustain compression, and so on as shown in the Ref. [28] which presents the calculation results of a massive bridge pier and a precast prestressed concrete

beam. Once the FEM calculation has been validated by comparison with the actual behaviour of the structure, it is then possible to predict the future evolution of the structure and to decide on treatment solutions.

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u (mm)

w (mm)

7.82 - 8.71

8.94 - 9.98

6.93 - 7.82

7.90 - 8.94

6.05 - 6.93

6.87 - 7.90

5.16 - 6.05

5.84 - 6.87

4.28 - 5.16

4.81 - 5.84 3.78 - 4.81

3.39 - 4.28

2.74 - 3.78

2.51 - 3.39

1.71 - 2.74

1.62 - 2.51

0.68 - 1.71 0.74 - 1.62

–0.36 - 0.68

–0.16 - 0.74

Fi g. 6: ISO values of radial (u) and vertical (w) displacements of a pier column 1000 AR_1 AR_2 AR_3

800

Stresses (MPa)

Present state

AR_4 AR_5

700 600

AR_1 AR_2

AR_6

AR_6

500

AR_3 AR_4 AR_5

400 300 200 100 0 –100

Protection of Users, Third Parties or Structural Elements

Z

Y

900

be designed for protecting users, third parties or structural elements, delaying the beginning of a reaction or slowing down its growth, reducing the mechanical consequences of the pathology, and rehabilitating the structure when the reaction is completed.

The main solutions to protect third parties and users from falling concrete blocks are the prohibition of access to dangerous zones, the removal of loose blocks and scales, and the installation of protection nets.

For structural elements exposed to shock in tidal zones, a mechanical protection against erosion may be done by surrounding the element with concrete shells. These shells may be either disconnected from the element to allow its free expansion, or bound to the element in order to bring a confinement effect to oppose the residual expansion.

Mitigation of the Reaction

0

50

100

150

200

250

300

Time (months)

Fi g. 7: Evolution of the calculated stresses in the reinforced steel bars of the pier column

To illustrate the results of such a calculation, Figs. 6 and 7 show, respectively, the expansion of a pier column and the computed stresses in the steel reinforcement of this pier column. In order to understand Fig. 6, it should be considered that the column is restrained at the bottom by the footing and at the top by the crossbeam that are both non-affected by DEF, and that the downward deflection of the crossbeam is due to the dead and live loads of the deck resting at the end of the crossbeam. The maximum vertical displacement at the top of the column reaches a maximum value of about 10 mm, while the transversal displacements reach 9 mm. The transversal swelling is more significant in the lower part of the pier where the saturation degree is higher (immersion of the footing and the bottom of the column in the river).

These results highlight the relationship between expansion and water content in the concrete structure.

Figure 7 shows that, after a latency period, the swelling of the structure causes tensions in the steel bars which become quite important in the case of

vertical (AR_6) and circular (AR_3, AR_4 and AR_5) bars, where the swelling is greater. The values of the tensile tresses in these reinforcing steel bars exceed 800 MPa for the final expansion, whereas they do not exceed 330 MPa in AR_1 and 500 MPa in AR_2 (Fig. 7). In these calculations, a purely elastic behaviour of steel bars has been assumed; however, in reality, stresses over 500 MPa cannot develop in the reinforcement and the results have to be interpreted as an incursion of steel in the field of plasticity. The most important point is that the tensile stresses in the reinforcing steel bars will probably exceed the yielding threshold after about 3 years.

Treatment Solutions Despite the fact that experiments of treatment are still extremely rare in the case of DEF-affected structures, some conclusions may be drawn from the few bridges damaged by DEF in France and from some laboratory tests.29 These treatment solutions may

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The main actions to mitigate the DEF are to limit water penetration into the concrete and to avoid contact with water. Different solutions with various efficiencies exist.

Injection and bridging of cracks are generally ineffective remedies which do not act on the reaction and do not produce noticeable improvement in the concrete strength to resist the development of cracking. This technique simply prevents water from penetrating inside the cracks, provided that the product injected or used for bridging is sufficiently flexible to follow the likely future opening of the cracks.

The application of paint on structures constitutes the simplest method to implement coatings. The painting layer has a small thickness (maximum 200–300 microns), and its efficiency is rather low and even null in the case of totally immersed structures, as proved in laboratory experiments.29 The application of a watertight coating having generally a significant thickness (a few millimetres) may act against the reaction while trying to reduce the penetration of water in the structure. It is a provisional solution that allows a possible extension of the lifespan of the structures.

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Finally, the drainage of water, the water collection, the renewal of a deteriorated waterproofing membrane, and the sealing of joints between elements are always beneficial to mitigate the reaction.

to be leading to much more expansion than AAR, the stress release could have a short-term effectiveness, and will presumably require repeating the operation insofar as it does not oppose the progress of the reaction.

Mechanical Limitation of Expansion

Other Solutions

Different methods of mechanical treatment exist: addition of reinforcing steels or composites when the objective is a simple strengthening, confinement with a high compression when the aim is a remarkable reduction of the expansion, the release of stresses, and the separation of swelling elements from non-swelling elements in the structure.

The demolition and then the replacement of a structure is sometimes an inevitable but expensive solution. It has been proposed in the case of the structure presented in Fig. 1 (reconstruction of the base of the pylon of the Bourgogne bridge), but facing the technical difficulty, the risk arising during the load transfer and the high costs of this replacement, this solution was abandoned in favour of a solution for extending the life of the affected structure. The expensive and traumatic aspects of a complete demolition can be attenuated by the realisation of a partial demolition followed by a rebuilding.

Fixing the amount of compression to be added to structures to fight against expansion is a real challenge. Based on previous cases of structures affected by AAR, the amount of additional compression stress to be introduced by prestressing is in the range of 5–10 MPa, and sometimes more.

Considering that in expansion tests conducted in the laboratory, the free expansion of specimens affected by DEF is five to ten times higher than the free expansion of specimens affected by AAR, it can be deduced that the level of compression stresses should be rather high. But sufficient experience in this domain is not available, and it is paramount to take into account the exposure condition of the structure and the residual potential swelling of the concrete.

Among the existing techniques, it is thus possible to insert, to enclose, or to hoop elements of structures with reinforcement or prestressing tendons and bars. These strengthening units are primarily made of steel, but the use of composite materials is increasing. Mechanical “restraining” has no effect in the untreated directions; it has a short-term effectiveness which is not always systematic, and a longterm effectiveness which is not proved. If a “restraining” is considered for treatment, then a threedimensional active strengthening (with prestress) is recommended.

Another type of active treatment consists in releasing the constraints by sawing whole or part of the structure.

This operation is often applied with success on gravity dams affected by AAR, and there is no reason why this type of action would not be effective in the case of DEF. But if DEF seems 368

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Finally, it should also be mentioned that monitoring may be a solution to manage the risks due to DEF or to give more time to take a decision concerning the future of the structure while ensuring its safety of use.

Conclusions DEF may be considered as a relatively new reaction to degradation of concrete. Presently, only some countries appear to be affected by DEF cases; it is probable that new DEF cases will be discovered in already affected countries and in other countries not yet affected.

Designers and owners of structures must be aware of this possible pathology. Massive parts and precast products are most vulnerable to this reaction, depending on the temperature reached in the concrete during the setting and on the conditions of exposure. The main challenge for the owners is therefore to know how to deal with these structures.

After a presentation of the disorders and their causes, an assessment procedure inspired by the appraisal methods developed for structures damaged by AAR has been proposed in this paper.

It comprises three steps. The initial step reviews the available documentation on the design and construction of the structure. The second step is a combination of material and structural approaches with the objective to

distribute the stock of structures into four severity classes. The third step is for structures in the highest severity class and consists of a structural assessment based on a non-linear FE analysis and a fitting of the expansion model with measures implemented on the structure. It helps to check the conditions of a structure affected by DEF in terms of safety, serviceability and durability limit states. An example of application on a pier of a bridge shows that the determination of the mechanical condition of the structure is predominantly governed by the variation of stresses in the different reinforcing steels.

Then, various treatment solutions are presented and discussed. They are classified according to their purpose: to protect users, third parties or structural elements, to mitigate the reaction, to limit the expansion by mechanical means, and to monitor the structure to limit the risks. It appears that there is no definitive and reliable solution to repair, on the long term, structures affected by DEF.

References [1] Divet L, & Randriambololana R. Delayed ettringite formation: the effect of temperature and basicity on the interaction of sulphate and C-S-H phase. Cement Concrete Res 1998; 28(3): 357–363.

[2] Tepponen P, & Eriksson B-E. Damages in concrete railway sleepers in Finland. Nordic Concrete Res. 1987; 6: 199–209.

[3] Heinz D, Ludwig U, & Rudiger I. Delayed ettringite formation in heat treated mortars and concretes. Betonwerk und Fertigteil-Tech. 1989; 55(11): 55–61.

[4] Vitouva L. Concrete sleepers in CSD tracks.

International Symposium on Precast Concrete Railway Sleepers, Madrid, 1991, 253–264.

[5] Shayan A, & Quick G-W. Microscopic feature of cracked and uncracked concrete railway sleepers. ACI Mater. 1992; 89(4): 348–361.

[6] Oberholster R-E, Maree H, Brand J-H-B. Cracked prestressed concrete railway sleepers: alkali-silica reaction or delayed ettringite formation. Proceedings of the 9th International Conference on Alkali-Silica Reaction or Delayed Formation in Concrete, CS104, vol. 2. London, 1992; 739–749.

[7] Mielenz R-C, Marusin S-L, Hime W-G, & Jugovic Z-T. Investigation of prestressed concrete railway tie distress. Concrete Int. 1995; 17(12): 62–68.

[8] Sahu S, & Thaulow N. Delayed ettringite formation in Swedish concrete railroad ties.

Cement Concrete Res. 2004; 34(9): 1675–1681.

[9] Santos Silva A, Gonçalves A-F, Pipa M. Diagnosis and prognosis of Portuguese concrete

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railway sleepers degradation – a combination of ASR and DEF. Proceedings of the 13th International Conference on AAR in Concrete, vol. 89(4). Trondheim, Norway, 2008; 1240–1249.

[10] Collepardi M. Damage by delayed ettringite formation. Concrete Int. 1999; 21(1): 69–74.

[11] Hobbs D-W. Cracking of concrete attributed to delayed ettringite formation. Proceedings of the Eleventh Annual BCA/Concrete Society Conference on Higher Education and the Concrete Industry, paper 6, UMIST, Manchester, 2001; 51–60.

[12] Thomas M, Folliard K, Drimalas T, & Ramlochan T. Diagnosing delayed ettringite formation in concrete structures. Cement Concrete Res. 2008; 38: 841–847.

[13] Ingham J. Delayed ettringite formation in concrete structures. Proceedings of the ICE – Forensic Engineering 2012; 165(2): 59–62.

[14] Divet L, Guerrier F, & Le Mestre G. Existe-t-il un risque de développement d’une réaction sulfatique d’origine endogène dans les pièces en béton de grande masse ? Le cas du pont d’Ondes. Bull. Laboratoires Ponts et Chaus. 1998; 213: 59–72.

[15] Godart B, Divet L. Lessons learned from structures damaged by delayed ettringite formation and the French prevention strategy. Proceedings of the Fifth international conference on Forensic Engineering, 15–17 April 2013, Institution of Civil Engineers, London.

[16] Divet L. Les réactions sulfatiques internes au béton: contribution à l’étude des mécanismes de la formation différée de l’ettringite. Etudes et recherches des laboratoires des Ponts et Chaussées, no. 40. LCPC, OA, 2001; 227 pp.

[17] Godart B, Divet L. Recommandations pour la prévention des désordres dus à la réaction sulfatique interne. Techniques et Méthodes des laboratoires des ponts et chaussées, LCPC, 2007, (English version published by LCPC in 2009).

[18] Godart B, & Divet L. DEF prevention in France and temperature control at early age. In CONCRACK 3 – RILEM-JCI International Workshop on Crack Control of Mass Concrete and Related Issues Concerning Early-Age of Concrete Structures, Toutlemonde F, Torrenti JM (eds) RILEM Publications: Paris, 2012.

[19] Martin R.-P., Bazin C., Toutlemonde F. Alkali aggregate reaction and delayed ettringite formation: common features and differences.

Proceedings of the 14th International Conference on Alkali Aggregate Reaction, Austin, 21–25 May, 2012.

[20] Martin RP. Analyse sur Structures Modèles des Effets Mécaniques de la Réaction Sulfatique Interne du Béton. Thèse de doctorat (PhD Thesis, in French), Université Paris-Est, LCPC, Paris, France, 2010; 577 pp.

[21] Al Shamaa SM, Lavaud S, Divet L, Nahas G, & Torrenti JM. Influence of relative humidity on delayed ettringite formation.

Cement Concrete Compos. 2015; 58: 14–22.

[22] Kchakech B, Martin R-P, OmikrineMetalssi O, Renaud J-C, Baron L, Toutlemonde F. Experimental study of the influence of late heat treatment on the risk of expansion associated with delayed ettringite formation. Proceedings of the 15th International Conference on Alkali Aggregate Reaction, Paper 087, Sao Paulo, 4–8 July 2016.

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[23] Godart B, & Wood JGM. Appraising structures affected by the alkali–aggregate reaction. Construct. Mater. 2016; 169(3): 162–171.

[24] LCPC. Réaction Sulfatique Interne au béton - Essai d’expansion résiduelle sur carotte de béton extraite de l’ouvrage. Techniques et Méthodes des Laboratoires des Ponts et Chaussées, Méthode d’essai des LCPC, 67, 2009; 28 p.

[25] Li K, & Coussy O. Concrete ASR degradation: from material modeling to structure assessment. Concrete Sci. Eng. 2002; 4: 34–46.

[26] Larive C. Apports combinés de l’expérimentation et de la modélisation à la compréhension de l’alcali-réaction et de ses effets mécaniques. Etudes et Recherches des Laboratoires des Ponts et Chaussées, LCPC, OA, 28, 1999.

[27] Brunetaud X. Etude de L’influence de Différents Paramètres et de Leurs Interactions Sur la Cinétique et L’amplitude de la Réaction Sulfatique Interne au Béton. Thèse de doctorat (PhD Thesis, in French), Ecole Centrale, Paris, 1996.

[28] Godart B, Omikrine-Métalssi O, Kchakech B, Lavaud S. The effects of delayed ettringite formation on the expansion of a massive bridge pier and a precast prestressed concrete beam of a bridge. Proceedings of the 15th International Conference on Alkali Aggregate Reaction, Sao Paulo, 4–8 July 2016.

[29] LCPC. Protection et réparation des ouvrages atteints de réactions de gonflement interne - recommandations provisoires. Techniques et Méthodes des Laboratoires des Ponts et Chaussées, Août, 2010, 141 pp. (In French)

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Human Errors in the Design and Execution of Steel Structures—A Case Study Izabela Tylek, Dr. Eng.; Krzysztof Kuchta, Dr. Eng., Cracow University of Technology, Faculty of Civil Engineering, Institute of Building Materials and Structures, Department of Metal Structures, Cracow, Poland; Anna Rawska-Skotniczny, Dr. Eng., Opole University of Technology, Faculty of Civil Engineering and Architecture, Department of Mechanics and Structural Engineering, Opole, Poland. Contact: anea.rawska@gmail.com DOI: 10.2749/101686617X14881937385287

Abstract Construction engineering activity requires decision making that involves a relatively large number of risk factors. These factors are related not only to the statistical variability of design data and model uncertainties but also to human proneness to commit errors. Human errors are inevitable and have a significant influence on the safety of building structures, which has been confirmed by the results of numerous structural failure analyses. This influence is also mentioned in the European Standard EN 1990, but human errors are not considered in the partial factor design. In this respect, the reliability required for the structure is achieved by applying quality management measures in design, execution, use and maintenance. The correct operation of the quality control system is then essential for the operational safety of the structure and the economic efficiency of the entire business project. This issue is illustrated with an analysis of the steel structure of an agricultural building in which human errors occurred at the planning, design and execution stages. Owing to the incorrect application of the quality management system, these errors had not been revised in due time, which led to the initiation of legal proceedings and finally to the decision to dismantle the building before putting it into use.

Keywords: human errors; reliability; robustness; quality control; steel structures.

Introduction An error is a part of human life; it may accompany any human activity. The nature of errors is complex, and their causes may be analysed from many points of view, not only technical but also psychological, sociological and organisational. This is probably the reason why there is no one, universal definition of error in literature.1 Hazards from human activity can be subdivided into two categories:2 • From outside the building process, which includes fire, explosion, collision and similar.

• From within the building process, which includes uncertainties due to acceptable practice and those caused by departures from acceptable practice.

Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: July 23, 2016 Paper accepted: November 21, 2016

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The practice was described as acceptable2 if no significant number of the most knowledgeable engineers finds it unacceptable. Common practice is not necessarily acceptable. Acceptable practice is not necessarily common.

Uncertainties due to acceptable practice are caused by the use of a new material or a new or unique type of structure, etc. Human errors as departures from acceptable practice may result from impudence, negligence or lack of knowledge. Acceptable practice is described, for example, in structural codes.

Human errors are inevitable, and they are the main source of uncertainties in the building process. Most structural failures are directly or indirectly connected with human actions or omissions. Contemporary studies2–5 estimate that about 70 to 90% of failures were caused by human errors.

Despite such a large influence on structural reliability, human errors are not included in standard design procedures of contemporary design codes, for

example, in the Eurocodes.6 In this respect, structural safety is provided by using quality management and control measures in the design, detailing, execution and maintenance of building structures. Therefore, an appropriate and efficient quality control system is crucial to ensure the required reliability level. The objective of a quality control system is to keep deficiencies and damages within acceptable limits. Quality may be quantified4—the less that has to be spent on repairs, deficiencies and damage, the better is the quality. Simultaneously, it should be remembered that the quality control system is created and implemented by people, so it may not be free from human errors itself.

The occurrence of human error is always undesirable, but it may be treated as a contribution to the development of engineering. This idea was very accurately expressed by Dostoyevsky in his famous novel Crime and Punishment: “through error you come to the truth!” The effect of an error is a guideline leading to the correct solution. The trial-and-error approach is widely used in the fields of computer science, biology and algebra.

Conclusions drawn from analyses of the structural failures3,7 still lead to improving design procedures; creating new solutions; and upgrading standards, guidelines, legal regulations, etc.

Error Classifications Human errors may be classified according to different criteria,1,2,4 for example, on the basis of causes, consequences and stages of building process. Generally, they may be divided into three basic categories: slips, mistakes and lapses (Fig. 1).

According to the definitions,8 a slip results from the incorrect execution of correct sequences of action, while a mistake is a correct execution of incorrect sequences of actions. The error is defined as a lapse when the action is omitted or not carried out at all.

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Unintended actions

Attentional failures

Lapses

Memory failures Knowledge-based

Mistakes errors

Intended actions

UNSAFE ACTS

Slips

Rule-based

Violation: routine violations, exceptional violations, acts of sabotage, terrorism

Fi g. 1: Taxonomy of human errors.

Adapted from Reason8

Depending on when the consequences occur, human errors may be categorised as manifest and latent. Manifest errors are committed by people who have direct contact with the structure, for example, crane operators and the consequences of their actions are perceptible almost immediately.

Latent errors are made by people who cannot check their own actions immediately, for example, designers, contractors and procedure writers.

Defects created by latent errors could exist in a structure for a long time, waiting for a combination of circumstances to occur, for example, abnormal or even maximum design load.8 Human errors can also be classified on the basis of information processing within the human mind. According to this system, there are three types of errors associated with three modes of performing the task by humans.9 The first mode is based on knowledge. In this mode, a person performs his or her work fully consciously, with a relatively large mental effort, trying to control the results as often as possible, which may be time-consuming. Errors associated with the knowledge mode are usually caused by incomplete/inaccurate understanding of the designed system or the behaviour of applied building materials, overconfidence, people’s inclination to prefer information that confirms their preconceptions and a psychological tendency for people to minimise cognitive strain. On the other hand, the rule-based mode takes place when the individual does not create his or her own solutions but uses known problem-solving rules.

In the rule-based mode, errors may occur if the incorrect rule is used, the correct rule is applied inappropriately or the rule is applied in a case of unknown rule exception. In this mode, problem solution is reached faster than in the knowledge-based mode because some part of mental activities is more “automated” with lower conscious control.

The last is the skill-based mode, which is

the least time-consuming type of problem solving. Human activities in this mode are well-practiced routine and have been done so many times that they require little conscious effort. Skill-based errors are slips, usually related to manual manipulation, for example, incorrect typing of numbers.

One of the most common classifications of human errors in civil engineering is based on the stages of building process; hence, there are planning, design, fabrication, transportation, erection, use and dismantling errors. This clear classification should be complemented with a type of error associated with another important activity accompanying all the aforementioned stages—quality control.

The observations and investigations of the Dutch building industry, reported in Refs. [10,11], also confirm that quality control is one of critical factors of building safety improvement.

Error Causes Most failures in the construction industry are the effect of multiple causes, which is confirmed by many researchers.2–4 Moreover, one error may result in the occurrence of others, or it may increase their probability. As it was accurately pointed out in Ref. [12], “Accident investigation is like peeling an onion.

Beneath each layer of causes and recommendations lie other layers. The outer layers deal with the immediate technical causes, the middle layers with ways of avoiding the hazards and the inner layers with the underlying weaknesses in the management system.”

The authors made an attempt to briefly characterise human error causes according to their own professional experience and literature study.

Organisational Causes The realisation of even the smallest building investment requires the cooperation of at least several people, while in the case of large projects, there may be thousands of them. Thus, the organisation of the building process should be clear and fully understandable for all the participants. Correct planning and management of the project, as well as proper organisation and coordination of the tasks, are necessary for the appropriate and timely completion of the work. Coordination of tasks becomes important already at the design stage because the structural

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bearing system and services are usually designed simultaneously. The design and execution of a structure should be sequential activities, but in the intensified time regime, project managers sometimes allow them to overlap. Such work organisation is possible, but if there is no appropriate supervision of the large number of tasks performed at the same time, and if the design and execution teams are poorly integrated, errors become more likely to occur. In this case, special attention should be paid to the procedure of changes introduction by the participants of the building process. The lack of control over changes may lead to the changes in the design concept and decreased structural safety.

The following two causes of human errors can be analysed from both organisational and psychological points of view. The first one is high personnel rotation, which may increase the probability of error occurrence, especially under the pressure of time, when a new employee is insufficiently acquainted with the documentation, applied technologies or the course of completed parts of work.1 However, if a person with more professional experience joins the team, personnel change may also have a beneficial effect. The second aforementioned cause of human errors may be the remoteness of top management from the place of real actions, where the problems arise. Such a situation is favourable for the superficial or incorrect assessment of situation, resulting in a wrong decision being taken.

Psychological and Sociological Causes The construction industry is inextricably linked with stress experience, which strongly influences the human psyche and may impede or even prevent the taking of rational decisions.

This stress may be caused by work overload and fatigue due to unrealistic schedules or interpersonal conflicts. In the building process, stress may also result from different kinds of pressure: business, economic and political. A person who works under stress is more prone to take risky decisions and less critically assesses his/her actions and applied solutions.

Incorrect decisions may also unsafe actions caused by the viour of an individual, such as sive (in relation to own

follow behaexcesskills)

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personal ambitions and boredom or lack of motivation generated, for example, by too low salary with respect to the difficulty of the task.

The origins of erroneous decisions may also be found in the manner in which the person functions in an organised community, such as a design office, execution company or inspection authority. The larger the scale of such an organisation, the more undesirable behaviours can be indicated. One of them may be the irrational succumbing to the influence of people who dominate in the given organisation.

Communication Causes Communication between the participants of the building process—understood as a flow of information presented in spoken, written or graphic form—is an essential element of every project realisation. Probably the first recorded case of a building project debacle caused by a lack of proper communication was the construction of the biblical Tower of Babel.

The lack of a common language is particularly visible in the communication process between the client and the designer. The designer may have difficulties understanding the investor’s or user’s needs, while the client may not understand the advantages and disadvantages of the presented solutions.

Additionally, the information may be incomprehensible because it is illegible, unclear, irrelevant, ambiguous or misinterpreted. It may also be distorted, delivered only partially or missing.

Today, due to the common availability of many communication media that the problem often does not consist of a lack of information but its excess. An overabundance of information may cause the overlooking of important details or the unintentional neglecting of essential facts. In extreme cases, information overload may cause “bureaucratic inefficiency”, leading to misunderstandings and delays in project realisation.

Errors may also be generated by intentionally withholding information, for example, alarming reports. The person responsible for this may fear a negative evaluation by supervisors and try to deal with the problem by himself or assume that the reported 372

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problem will “solve itself” while it requires actions taken on a higher organisational level.

Another case that can be placed at the intersection of organisational and communication causes of human errors is solving problems that an individual cannot handle. To avoid incorrect decisions, the organisation of the information flow should ensure the possibility of collective consultation of proposed problem solutions.

Causes Associated with Technological Background Human errors that are generally linked with technological background are associated mainly with ignorance.

The lack of knowledge is particularly perceptible in cases of using newly introduced structural codes. The complexity of contemporary design codes (e.g. the Eurocodes) and the increasing number of experimental instead of theoretical formulae may lead to misinterpretation and misapplication.1 The lack of professional experience in the design and execution of engineering structures triggers human errors, especially during the realisation of novel structures. It should be noted that for a novice engineer, almost every structure is a novelty, while for the experienced engineer, it is only unusual or experimental. A person with none or little professional experience often tends to apply non-technological, error-prone solutions, which may be the reason for the lower safety or serviceability of the structure and the increase in costs of a structure.

It cannot be denied that common usage of computers has contributed to the creation of new possibilities in the planning, design, execution and use of building structures. In particular, the positive influence of computer technology has taken place at the design stage, where the finite element method (FEM) is widely used, allowing the resolution of many complex issues that could not be solved until now. New numerical calculation models made it possible to reduce model uncertainties and, together with user-friendly graphical pre- and postprocessors, minimised the risk of committing gross errors. A negative consequence of computer use in the design process is the fact that engineers, especially novices, “let their guard down” and unconditionally

believe in numerical calculation results.1 Although computers practically do not make mistakes, with the exception of very rare cases, like errors in the early models of Intel Pentium CPU,13 human errors may occur in a computer software. Owing to the massive scale of design software distribution, an error made by one programmer may influence the safety of many building structures. It should still be remembered that computers and software are only tools and that the results of even the most accurate calculations will be erroneous if the model or input data (loads, material properties, joint characteristics, etc.) are incorrect.

Causes Associated with Quality Control Checking is one of the main tools for quality control assurance,1,2,4 but it may not be free from errors. The main cause of errors is the lack or insufficient design supervision or execution inspection, usually resulting from time pressure or concurrent or overlapping activities.

The rational assessment of skills and the mental and physical predispositions of an individual employee allow for a well-balanced assignment of tasks and responsibilities. Poor selection of technical personnel can be the beginning of a chain of errors caused by wrong people working in inappropriate positions. Inadequate training, resulting in insufficient knowledge and skills, may be the cause of an increasing frequency of errors made by employees of quality control division.

The source of errors may also be the quality control procedure itself. Inappropriate or unclearly formulated procedures distort the sense of quality control, both in design and execution.

Additionally, in case of design supervision, the verification of computeraided calculations may be impeded so that committed errors remain undetected. When the user has none or limited access to part of data used in the design model because authors of the software wrongly believed that they are irrelevant, the only effective form of design supervision is to repeat calculations using different software.

Other Causes The causes of human errors described above should be complemented by omissions, resulting from ignorance or

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indolence, and by greed, which is often manifested in a form of fraud or embezzlement. A frequent practice in the steel structures industry is that contractors change the original design to save weight and, consequently, the cost of structures.14 This action may be highly risky, and it usually results in the decrease of structural reliability.

When it is combined with several other unforeseen circumstances, such as human errors and abnormal loading, it may lead to a catastrophic failure of a structure with unimaginable consequences. An example of such a case was the tragic collapse of the steel structure building of the International Katowice Fair during the VIIth International Trade Fair of Carrier Pigeons on the 28 January 2006. It was one of the largest building catastrophes in Polish history: 65 people killed and over 140 injured. After experts’ investigations and court trials, the court ruled that the main cause of the collapse was human errors at the stage of redesign and maintenance of the building.

Other categories of human actions that may lead to the same consequences as unintentional human errors are sabotage and terrorist acts.

Strategies Against Human Errors There may be two approaches to taking account of human errors in the design of building structures. The first one is to incorporate the effect of human errors into the design procedures, for example, by assuming the appropriate value of probability of failure. Some attempts to include the effect of human errors in the design procedures by updating the probability of failure were made, for example, in Refs. [15,16]. The second strategy against human errors is to design the structure on the basis of notional probability of failure and, simultaneously, to apply relevant quality control measures.

The analyses of potential human errors and the strategies against these errors should be a part of individual risk analyses related to the particular building project. Both human errors and their mitigation measures generate costs for the investment and because of that cannot be neglected in cost optimisation studies. Usually, sensitivity to human errors is one of

the selection criteria of particular solutions during the cost optimisation analysis.

Contemporary design codes require that a structure shall be designed, executed and maintained in such a way that, during its design working life, it will sustain all actions and influences likely to occur during execution and anticipated use and that it will meet the specified serviceability requirements.

Additionally, the structure shall not be damaged by accidental events (e.g. impact) or consequences of human errors to an extent that is disproportionate to the original cause (robustness requirement). These specifications shall be fulfilled with appropriate degrees of reliability and in an economic way. The expression “with appropriate degree of reliability” means that the adopted level of reliability should take into consideration the possible causes, types and consequences of failure; public reaction to failure; and the expense, level of effort and procedures necessary to reduce the risk of failure.6 For the purposes of reliability differentiation, Annex B of EN 19906 introduces three consequences classes (CC1, CC2, CC3) and associates them with reliability classes (RC1, RC2, RC3). The higher the consequences of failure or malfunction of the structure, the higher the consequences class. The higher the risk to life, economic, social and environmental losses due to structure failure, the higher the reliability class. For each reliability class in the ultimate limit state, EN 19906 recommends a maximum value of probability of failure pf. As the reliability class is associated with the consequence class, the higher the consequences of failure, the higher the degree of reliability and then the lower the probability of failure pf. A design compliant with the Eurocodes leads to a structure with a value of probability of failure pf lower than pf0 = 7.2 × 10−5 for a 50 year reference period. It should be remembered that these values do not necessarily represent actual failure rates because they do not take into consideration the effects of human errors.1,4,6,11,17 Structures are exposed to two major types of hazards: natural and manmade. Natural hazards result from the unpredictability of actions such as wind, snow, earthquakes, etc. They also include natural variations of physical and mechanical material properties and dimensions. Manmade

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hazards are associated with any human activity. Recognition of possible hazards is a very important part of structural safety analysis because if the potential hazard is recognised, suitable measures can be applied to reduce its unfavourable effect on the structure.4,6,18,19 These measures may be divided into structural and non-structural ones. In case of structural measures, the structure is designed to have a sufficient strength reserve or alternative load path in case of local failure, while non-structural measures may include elimination of the hazard source, reduction of its probability of occurrence and reduction of action intensity or consequences of failure.

Hazard may also be by-passed by changing concepts or controlled by checking, monitoring or warning systems. Analyses of structural failure indicated that only about 10 to 30% of failures were caused by natural hazards, while the remaining 70 to 90% was derived from human activity. Most of the human errors were committed in the design and execution of a structure.2–,5,10,11 The application of structural reliability theory to structural design codes helped to better understand and introduce the risk of structural failure and, consequently, to design more reliable and economic structures.

As mentioned above, numerical values of probability of failure obtained from theoretical models normally do not take into account the influence of human errors and thus may lead to quite large discrepancies between the theoretical and observed failure probabilities. An example might be the bridge failure probability rate in the United States, which is 10−3 to 10−5 annually,2 while typical theoretical values are between 10−6 and 10−8. In this case, the failure20 is associated with Ultimate Limit States ULSs rather than with the serviceability criteria. Although the actual probabilities of failure are higher than theoretical values, they are still very low. Consequently, the database of failures is limited, and the development of structural reliability models that take into account human errors is quite difficult. Owing to this fact, contemporary European design codes do not include human errors in partial factor design.6 For that reason, the influence of human errors on structural reliability should be taken into account separately. The most common practice to reduce the frequency of human errors is to introduce quality Scientific Paper

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management and control measures at all stages of the building process. Consequences of human errors can be controlled by the application of sensitivity analysis to determine the impact of individual errors on the safety and serviceability of the structure.2 Detailed analyses of 800 cases of structural damages, conducted by Matousek in the late 1970s, indicate that 94% of failures were caused by human errors. Furthermore, 87% of all errors could have been detected in time: 32% by a careful review of the documents by the subsequent person in the process and 55% by additional checks resulting from well-planned quality management procedures.4 This confirms that quality assurance and supervision are essential in the prevention of human errors.1,2,4,10,11 In Europe, quality management and control measures aimed at the elimination of failure due to gross errors in design and execution of a structure and to ensure the resistance assumed in the design are provided in EN 1990.6 At the design stage, the concept of design supervision levels (DSL) is introduced.6 Design supervision differentiation is made on the basis of minimum recommended requirements for checking of calculations, drawings and specifications. At the stage of execution of a structure, EN 19906 proposes three inspection levels (IL) that are defined on the basis of the requirements for inspection during execution.

The consequences of human errors may be minimised not only by introducing quality assurance and control measures but also by the special shaping of a structure to ensure an appropriate structural robustness and to avoid progressive collapse.6,18,19,21 With respect to human errors, robustness is defined18 as the ability of a structure to withstand the consequences of human error without being damaged to an extent disproportionate to the original cause. Some rules on how to provide adequate robustness for a structure may be found in European codes.6,18 One of them is to introduce three-dimensional tying for the additional integrity of a structure.

Another important issue is limiting the extent of failure. Human errors may result in localised failure, or local failure may initiate the failure of other members and lead to progressive 374

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collapse. Some provisions aimed at reducing the extent of local failure are provided in EN 1991-1-7.18 Selection of an appropriate strategy for adequate structural robustness and minimising the extent of local failure may be based on three consequence classes defined in EN 1990.6 Adoption of the strategies recommended in EN 1991-1-718 should provide a structure with an acceptable level of robustness to sustain localised failure without a disproportionate level of collapse.

Case Study Consequences of human errors may be a key issue not only from the point of view of structural safety but also of the economic efficiency of the entire business project. The case described below concerns the steel structure of an agricultural building, which at planning, design and execution stages was not revised in due time and was dismantled before being putting it into use, due to human errors.

Selected Issues of Quality Assessment During the Building Process in Poland One of the measures to assure the quality of a structure at the design stage is the requirement, enshrined in Polish law, that any structure must be designed and checked only by certified building designers. At the stage of execution, construction work should be conducted by an engineer with a license in execution, and if an investor is obligated by the law (as in analysed case) to employ an execution supervisor, he or she must also hold a certificate in structure execution. The author of a building project is allowed to visit the building site any time and to stop the execution if it is not conducted according to the design documentation and if it poses a threat to the safety of people or property. Site managers and execution supervisors have the same right, for example, if they find an error in design documentation.

Workshop works, due to their large receptiveness for almost every building project, are easier to be controlled systematically than site work; an example is shown on Fig. 2.

It should be noted that the objective of the quality control system is to detect errors and to minimise costs of their repair. Additionally, it should help to improve structure execution

technology and the system itself. One of seven basic quality control tools is the checklist. The European Organisation for Quality recommends using checklists in quality control systems to achieve continuous quality improvement in production processes. The checklist is particularly effective in instances where the human activities have a high level of repeatability. In Poland, checklists are not obligatory during the building process, but in spite of that, they are used in some companies, which may confirm their efficiency. They are also commonly used during the technical inspections of bridges. The use of checklists is a wellaccepted quality control measure in the Polish civil engineering environment.

An exemplary checklist for the preerection stage of the steel structure is included as Fig. 3.

Although the questions included in the presented checklist may seem trivial, it may reduce the likelihood of errors for people with no experience working without supervision or for people who are experienced but work under time pressure. The checklist is also a form of recording the actions taken on the construction site and the assigning of responsibility for the supervision of these activities. The personal liability usually improves the diligence of performed actions.

In the described case study, the checklists were not used. The authors believe that the application of relevant checklists would help to prevent many mistakes.

The examples of issues specified in the checklist in Fig. 3, which might have prevented the errors shown in Table 1 if it had been verified thoroughly, are as follows: • “Does the foundation survey report for the location of anchor bolts/levelling pads axis marks on foundation comply with documentation?”

The verifications of this issue would prevent defective mounting of anchor bolts in foundation formwork (Table 1a), b)).

• “supports tolerance: holding-down bolts, cast-in concrete footing within tolerance”? The verifications of this issue would prevent the start of the assembly of steel structure and problems arising from the lack of fit between steel elements (Table 1c), d)).

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Third party suppliers

Project documentation: contract, specifications, drawings accepted Delivery: materials, fasteners, welding materials, anti-corrosion coatings others

accepted

recycling or disposal

no

Quality control: technical acceptance, commissioning inspection complaint

rejected usable?

Mechanical processing: cutting, shearing, sawing, punching, drilling, beveling storage recycling

no

repair usable?

yes

no

recycling

accepted

yes no

Quality control: technical acceptance

repair usable?

yes

no

repair rejected possible? accepted

recycling

yes

Assembling: welding, very rarely screwing or riveting; trial erection (when is needed) storage

no

repair usable?

yes

no

disposal

yes no

repair usable?

yes

no

Quality control: technical acceptance

repair rejected possible? accepted

Anti-corrosive protection layers execution: (painting, hot dip galvanising or both) storage

Quality control: technical acceptance

repair rejected possible?

Changing the geometry: straightening, bending, pre-cambering storage

rejected accepted

yes

storage

yes

Quality control: documentation review, check material quantities

QUALITY CONTROL DIVISION (error analysis, preventive action co-planning)

Documentation supplier rejected

Quality control: technical acceptance

repair rejected possible? accepted

On-site shipment

LEGEND: info feedback

Fi g. 2: Exemplary scheme work organisation in the steel structure, including quality control system

â&#x20AC;˘ â&#x20AC;&#x153;connection joints/splice joints constructed as per drawingsâ&#x20AC;?? The veriďŹ cations of this issue would force the contractor to contact the designer and would lead to the recalculation of the structure or replacement of improperly executed items (Table 1c), d)).

In case of errors, the signature under the checklist clearly indicates the person responsible for the lack of proper supervision, which can be used, for example, in the investigation or court proceedings.

Description of the Building Structure The analysed structure is a detached steel skeleton structure with a small mezzanine inside, located in Poland.

The main bearing structure was designed as a single-bay, two-hinged

portal frame with a truss roof girder and a relatively long cantilever truss (Fig. 4).

The structure was planned to serve as a small, root vegetables processing plant. The interior of the building was divided into two parts with different functionalities. In the ďŹ rst part of the building, vegetables are delivered, washed, processed and packed. The second part is a chilling room for vegetable storage in cold (2â&#x20AC;&#x201C;5 C) and moist (~97% humidity) conditions.

Owing to these service requirements, the roof and walls were covered with sandwich panels ďŹ xed to the internal surfaces of the framesâ&#x20AC;&#x2122; steel proďŹ les.

Error Analysis Planning Stage The ďŹ rst error was made at the early planning stage; despite the high

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humidity and low service temperature inside the building, the designer initially assumed that the roof and wall panels would be ďŹ xed to the external surfaces of steel proďŹ les, like in a typical building steel structure. In spite of that, the structure was designed with no regards for the loads generated by the thermal gradient. Probably because of the lack of knowledge, appropriate calculations were not conducted even after the decision about changing the place of sandwich panel ďŹ xation had been made.

Design Conceptual Stage Numerous errors made at this stage led to the fact that subsequent repairs became very expensive. Usually, there is no one and only optimal solution for the given real-world engineering problem because there are many variable criteria and parameters of solution evaluation. Moreover, the analyses conducted according to different criteria frequently lead to opposite conclusions.

In the described case, the main frame structure concept was correct from the point of view of building static analysis, but it was not economical. The frame bay length was too small to take advantage of using the truss roof girder because savings of material costs were completely lost as a result of increasing labour cost in comparison with a girder made of hot-rolled or cellular sections.

The selected solution also had a lower corrosion durability. Owing to the application of hinged columnâ&#x20AC;&#x201C;truss connections, the column-bending moments had to be taken by the columns themselves. Applying moment connections would reduce columnbending moments and cantilever deďŹ&#x201A;ections due to the incorporation of bending stiffness of all frame members. The designerâ&#x20AC;&#x2122;s lack of professional experience is also manifested in the manner he shaped the cantilever. It would have been a better idea to link the cantilever endpoint with the frame eave corner by a tie or strut, depending on the cantileverâ&#x20AC;&#x2122;s self-weight and its loadings.

Calculation Stage Structural analysis was conducted by means of decomposition of the threedimensional structure into plane static systems. The problem, and one of the main errors, is that the designer overlooked the superposition of action effects in members belonging to more than one plane system. Static analysis did not take into account the imperfections and second-order ScientiďŹ c Paper

375

and slips, for example, lack of bolts descriptions and discrepancies, between calculated (and built in) and drawn sizes and types of steel profiles (Table 1d). One of the main reasons of execution stage problems was the bolted end plate connection between column and wall rafter. This type of connection is very sensitive to the inaccuracy of rafter length, but the designer did not take it into account.

The rafters were designed without negative tolerance for their length, and consequently, it was impossible to apply corrective packs.

Size and length of weld according to approved plan and specifications?

Welding electrode strength comply with specifications?

Weld surface clean and free from dust, rust and scales?

Qualified welder and certificate provided?

Any welding defects? (undercuts, pits, surface cracks, lack of fusion) The remedy action? (attached report) Ultrasonic test?

MPI test?

Rradiographic test?

Penetration flaw detection test?

Others?

Location plan provided?

Any remedial action taken for welding test failures?

Anti-corrosive protection layers execution comply with the specifications?

Workshop Works and Erection Stages

Anti-corrosive protection layers execution comply with the specifications?

F ig . 3: Exemplary checklist for the pre-erection stage of the steel structure

effects. In the calculations report, no calculations of bracing systems were found. In consequence, purlins, columns and other structural elements were designed without any contribution of additional forces derived from bracing system action. Several mistakes were made during the modelling of loads. Some of the factors not included in the design model are the variability of environment temperature, self-weight of cooling units installed on the roof and accidental loads due to cargo vehicles and forklifts colliding with the structure. For unknown reasons, the designer assumed that the total load (selfweight and service load) acting on the mezzanine floor was only 0.2 kN/m2.

Additionally, the load combinations were also created incorrectly.

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Specifications Execution specifications were incomplete. There were no bolts, nuts and washer lists. The designer did not provide any specific information about the execution of column base grouts.

There were no operating instructions and no tips for geometrical imperfection compensation.

Bolt grade and type conform to specification?

Bolts dimension (diameter x length)?

Washer grade and type ?

Connection joints/splice joints constructed as per drawings?

Bolts tightened to correct torque?

Min 1,5 thread beyond nut?

Any defects? (eg tilted bolts / holes enlarged by torch cutting) The remedy action? (attached report) Embedment length and arrangement of holding down bolts location Plan provided? (attached hardcopy)

Welding connection Welding test?

Pre-erection of steel structure activity Bolted connection

Steel

Pre-planning

Quality checklist before erection work Yes No N/A Comment Documentation complete? (specifications, drawings) sequential erection procedure provide?

Erection crew inducted into sequential erection procedure?

Erection crew inducted into the safe work method statement? ground or supporting surface suitable for plant to operate safely Manufacturer has the necessary welding facilities, qualified welders and a shelter fabrication yard?

Weather conditions monitored for potentially hazardous conditions?

Foundation survey report for location of anchor bolts/levelling pads axis marks on foundation comply with documentation?

Supports tolerance: holding-down bolts, cast-in concrete footing within tolerance?

Materials (fasteners, welding materials, anti-corrosion coatings, others) comply with the specifications?

All elements appropriately marked?

Legibly and durably marked?

Free from defects? (warping, twisting, distortion, damaged section, pitting) Cutting and steel section dimension (d x b x I)?

Thickness (flange and web)?

Stiffener plates as per drawing?

Detailing Stage The first version of the structural design project assumed that steel elements were corrosion-protected with paint coatings, and most site connections were welded joints. In the initial phase of assembly, corrosion protection was replaced with hot-dip zinc galvanisation, and all site connections were redesigned as bolted splices. It should be emphasised that the design documentation did not contain connection calculations. Consequently, it cannot be said, for example, whether the bearing capacity of moment-resistant column base, executed without a shear key, is sufficient to take horizontal reactions only by friction (see Table 1).

Analysis of the workshop drawings showed that there are many lapses

There was a strong interaction between these two stages. In situ casted concrete foundations were executed with large geometrical imperfections, exceeding the code tolerances several times. The careless execution of foundations caused column axes to deviate from their designed positions (Table 1); code tolerances were fulfilled only in two columns. The contractor changed the type of wall rafter to column connection into shear connection, which is less susceptible to rafter length deviations, without notifying the designer about it. The structure was probably assembled without using any land surveyor instruments—this is the only possibility to explain why column out-ofplumbs also exceeded code tolerances.

Magnitudes of two aforementioned geometric imperfections exceed the allowable tolerances several times, so even shear type connections were unable to compensate it (see Table 1).

Generally, the assembly works were carried out very incorrectly. The contractor’s employers demonstrated a lack of elementary knowledge about steel structure erection techniques.

Table 1 presents a few examples of errors due to an evident lack of a proper sense of structure’s work, which results in change of intended structural behaviour.

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As designed

As built

• column axis deviations more than four times greater than tolerance; consequently, additional bending moments were introduced into the column and foundation • initial corrosion marks, probably due to unremoved anti-spatter agent layer before galvanising • small thickness of anchor washers, failure hazard due to punching shear on oversized anchor holes • lack of grout causing a hazard of local foundation overloading and concrete cracking

(a)

z

y

max max

y z

= 31 mm, = 22 mm,

max overall inclination 33 mm

Point of reference y, z

Remarks

5 mm

• anchor bolts too close to the foundation edge, resulting in deceased bearing capacity and corrosion durability of the tensioned anchor bolts

(b) 25

32 (c)

• lack of joint tension resistance that may cause loss of stability of the supported element • unacceptable change of the design solution—vertical forces should not be transferred between elements

120

41

(d)

Top view 160 171

Dimensions in mm 250

160

polyurethane core sandwich panels

SHS 100x100x4

Wrong profile

100

• discrepancies between calculated and drawn sizes and types of steel profiles. During the execution, the constructor used the profiles indicated in the calculations, not these marked on the drawings • large, more than 1/3 of section depth, eccentricity between centroids of wall rafter cross-section and group of bolts • lack of cropped gusset plate corners for proper drainage during galvanisation • lack of joint tension resistance that may cause loss of stability of the frame column

Table 1: Selected execution errors of the steel structure

Summary The detection of corrosion marks induced the investor to stop payment for the construction works and initiate legal proceedings. The court expert investigations showed that the total

effect of human errors on the reliability of analysed structure was as follows: the resistance utilization factor of the most loaded element was exceeded nearly six times, and the serviceability condition was exceeded almost seven times. Although it was

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possible to repair the building structure, it was not accepted by the investor because of its large costs and influence on the functionality of the plant. In this situation, a decision ordering the dismantling of the steel structure was made.

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377

8.70

(a)

(b)

ed steel she

ugat

eting

2.9

Corr

4.20

2.9

Polyurethane core sandwich panels

0.5

0.00

7

1

4

4.3

Chilled room

6.0

12.0

1 B 3.0

1.8 2.4

Processing & packing room

A 0.3

(d)

6.0

Projection of mezzanine floor

7

(c)

2.6

C 5.6

D 2.6

3.0

E 5.6

F 5.6

0.3

3.9

steel sections are intentionally overscaled, red colour indicates columns executed out of tolerances bold black dashed lines indicate planes of truss bracing systems blue lines indicate sandwich panels

F ig . 4: The root vegetable processing plant: (a) longitudinal wall; (b) gable wall; (c) horizontal projection; (d) static scheme of main frame; units [m]

It was an unfortunate coincidence that during the realisation of the structure described above, an inexperienced designer met a shoddy and ignorant contractor. Fortunately, there was no casualty or injuries. From the financial point of view, however, the case was a complete business failure. The investor’s losses due to unrealised contracts were several times greater than the building cost.

Conclusions Human errors are an inevitable part of building process. Analyses indicate that human errors are the major cause of structural failures, but they are not included in design methods of contemporary structural codes. Structural reliability is instead provided by the application of quality management and control systems. For that reason, it is very important to have this system run properly. In the authors’ opinion, particular attention should be paid to avoiding human errors during checking and revision stages because these actions may be the last barrier against human errors. In the analysed case study situation, it appears that the system had failed.

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The main question is how could it be possible? Why did such a simple structure, with so many checking and supervision procedures, become dangerous for human life? In the authors’ opinion, all the aforementioned human errors could have been avoided if people who are obliged to control the process had done their work properly. The following may be suggested in order to improve checking and supervision:

particular design and execution errors are. This could be the basis of the discussion on how the effects of human errors and efficiency of possible mitigation measures should be quantified. This problem is very important, because human errors may change the probability of failure by several orders of magnitude.2,15 This makes human errors one of the most significant uncertainties in the design of building structures.

• Frequent professional training.

• Using a checklist.

• Better selection of checkers and field inspectors, for example, selection according to psychological predispositions, increasing of knowledge requirements associated with appropriate salary.

• Gathering, analysis and dissemination of conclusions and recommendations formulated on the basis of structural failure investigations among civil engineers.

The observations and conclusions presented above concerned only the Polish engineering practice and legal system. However, generalisation of conclusions and formulating the code recommendations for other countries could not be justified; the paper may be treated as a voice in a discussion that may be taken into account in further revisions of European codes.

One of the issues that is planned for analysis in a further work is a more detailed numerical quantification of human errors’ effect on the reliability of the structure described above. The results of the analyses would indicate what the safety consequences of

References [1] Fröderberg M. The Human Factor in Structural Engineering: A Source of Uncertainty and Reduced Structural Safety Lund University: Lund, 2014.

[2] Nowak AS, Collins KR. Reliability of Structures CRC Press: Boca Raton, 2013.

[3] Palmisano F, Ratay R. The practice of forensic structural engineering in IABSE

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member countries: review of a survey. Paper presented at: IABSE Conference Geneva 2015.

IABSE Symposium Report of the IABSE Conference Geneva 2015: Structural Engineering: Providing Solutions to Global Challenges, Geneva, Switzerland, 22–25 September 2015; 1–4.

[4] Schneider J. Introduction to Safety and Reliability of Structures Zurich: IABSE-AIPCIVBH, 2006.

[5] Vrouwenvelder T, Holicky M, Sykora M. Modelling of human error. Paper presented at Joint Workshop of COST Actions TU0601 and E55 2009. Proceedings of the Joint Workshop of COST Actions TU0601 and E55, Ljubljana, Slovenia, 21–22 September 2009; 55–64.

[6] EN 1990. Eurocode. Basis of Structural Design. CEN: Brussels, 2002.

[7] Petroski H. Patterns of Failure. Modern Steel Construction [serial on the Internet]. Available from: http://msc.aisc.org/globalassets/modern-st eel/archives/2006/07/2006v07_patterns.pdf. Cited 16 July 2016.

[8] Reason J. Human Error Cambridge University Press: New York, 1990.

[9] Rassmusen J. Skills, rules, knowledge: signals, signs and symbols and other distinctions in human performance models.

IEEE Trans. Syst. Man Cybern. 1983; SMC13(3): 257–266.

[10] Terwel K, Boot W, Nelisse M. Structural unsafety revealed by failure databases. Proc.

Inst. Civil Eng. Forensic Eng. 2014; 167(1): 16–26.

assessment method for structural engineering.

MSc Thesis. T.U. Delft, Netherlands, 2012.

[16] Stewart MG. Modelling human error rates for human reliability analysis of a structural design task. Reliab. Eng. Syst. Saf. 1992; 36: 171–180.

[11] Terwel K. Structural safety: study into critical factors in the design and construction process. PhD thesis, T.U. Delft, Netherlands, 2014.

[17] Diamantidis D. Current safety acceptance criteria in codes and standards—a critical review. Paper presented at: ASCE Structures Congress 2008. Proceedings of Structure Congress 2008: Crossing Borders, Vancouver, Canada, 24–26 April 2008; 1–8.

[12] Kletz TA. Process industry safety. In Engineering Safety, Blockley DI (ed) McGraw-Hill Book Company: London, 1992; 347–368.

[18] EN 1991-1-7. Eurocode 1. Actions on Structures. Part 1-7: General Actions - Accidental Actions. CEN: Brussels, 2006.

[13] Chapman MR. In search of Stupidity: Over 20 Years of High-tech Marketing Disasters Apress: Berkeley, 2007.

[19] Vrouwenvelder T, Leira BJ, Sykora M. Modelling of hazards. Struct. Eng. Int. 2012; 1: 73–78.

[14] Ameer ALM. Irregularities, Frauds and the Necessity of Technical Auditing in Construction Industry AuthorHouse: Bloomington, 2013.

[20] ISO 2394. General Principles on Reliability for Structures. ISO: Geneva, 2015.

[15] De Haan, J Human error in structural engineering. The design of a human reliability

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[21] Sørensen JD, Rizzuto E, Narasimhan H, Faber MH. Robustness: theoretical framework.

Struct. Eng. Int. 2012; 1: 66–72.

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Rehabilitation and Lessons Learned from the Collapse of Viaduct 1 Located on the Caracas–La-Guaira Highway in Venezuela Luis B. Fargier Gabaldón, Civil Eng., Department of Civil Engineering, Núcleo la Hechicera Facultad de Ingenieria, Universidad de Los Andes, Merida, Bolivarian Republic of Venezuela. Contact: lfargier@dcintegrados.com DOI: 10.2749/101686617X14881937385007

Abstract The Viaduct 1 located on the Caracas–La-Guaira Highway in Venezuela collapsed in March 2006. The structure built in 1953, consisted of three double-hinged arch ribs made of plain concrete, spanning 152 m over the Tacagua creek, and two approach bridges for a total length of approximately 300 m. The reactivation of an ancient landslide affected the Caracas side of the Viaduct, shortening the arch span by more than 1 m, which eventually led to its collapse. This paper summarizes the results from the geotechnical investigation, the impact of the landslide on the structure from 1987 to 2006, the soil—structure interaction and the rehabilitation measures to extend the service life of the bridge, which included external post-tensioning of the ribs, design of a plastic hinge at the crown and lateral translation of the structure.

The causes of failure, mechanism of collapse and lessons learned will be presented.

Keywords: Concrete arch; bridge rehabilitation; external post-tensioning; plastic-hinge; landslide reactivation; structural collapse.

Description of the Structure The structure is shown in Fig. 1. The main span consisted of three doublehinged arch ribs spanning 152 m and two approach bridges with identical spans of 14.60 m: a three-span approach bridge located on the Caracas side (43.8 m) and a longer, sixspan approach bridge (87.60 m) on the La-Guaira side. The approach structures were connected to the main span through two hollow fourcelled elements (6 m × 24 m in plan and 42 m tall) referred to as pilasters1, which were “…placed at each side of the arch to create a major architectural feature avoiding the unpleasant appearance of a change without transition from the main structure to the approach viaducts”1.

The aesthetic of the bridge was of main importance during the design phase.

The piers consisted of three I-shaped columns with the cross-section shown

Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: August 5, 2016 Paper accepted: November 10, 2016

380

Scientific Paper

in Section B–B0 (Fig. 1), built-up with precast elements that were prestressed in the vertical direction.

The three I-shaped columns at each pier were linked at the top by a transverse beam made up of five precast elements, which were also prestressed. Boundary conditions for the piers were as follows: • I-shaped columns at the approach bridge located on the La-Guaira side had a Freyssinet hinge at the bottom and at the top and were rigidly connected to the deck.

• I-shaped columns supporting the approach bridge on the Caracas side were hinged only at the bottom, except for the shortest pier, closest to the abutment, which had hinges at both ends.

• I-shape columns located within the main arch span were rigidly connected to the ribs and the deck, except for the shortest piers, closest to the crown, which were hinged at both ends.

Freyssinet hinges behaved extremely well during the service of the bridge and demonstrated an exceptional rotation capacity before and after the reactivation of the landslide. Freyssinet hinges also

exhibit excellent behaviour in Viaducts 2 and 3 with arch spans of 145 and 138 m, respectively, which are in service and located on the same highway. Except for the approach bridges, these two structures are very similar to the Viaduct 1. Freyssinet2 concluded that the hinges he invented “…behaved like a liquid mass in an inextensible tube, without showing disruption or disorder”. This statement almost perfectly reflects what was observed in the field after the landslide affected the structure. A complete description of the Freyssinet hinge, and other concrete hinges used in bridge structures can be found in Ref. [3].

At the pilaster, girders rested on a roller-type support, which consisted of a rocker wall, 1.00 m deep, also hinged at both ends (Fig. 1). At the abutments, similar rockers supported the girders. The structure was proportioned to resist a wind pressure of 2.40 kPa and designed in accordance with the AASHTO HS-20-44 loading specification.

Each span consisted of eight girders.

Within the main span, six girders were attached to the arch ribs near the crown, and the two intermediate girders were longer and attached to a transverse diaphragm that connected the three ribs at the crown. It is also interesting to note that the deck was continuous from abutment to abutment. Prestressing cables placed in grooves over the top flange of the girders were anchored into the arch near the crown, providing continuity of the deck to resist wind forces. Thus, the deck had only two expansion joints, one at each abutment. With this configuration, it follows that the ribs and pilasters provided most of the deck stability in the longitudinal and transverse directions, respectively.

Ribs consisted of three parallel box sections, placed 8.4 m apart at the centre, with a width of 3.2 m and an

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La-Guaira

Caracas 4% 152.00

A

6.00

87.60

Pilaster

Abutment

Crown Pilaster

43.80

6.00

Arch rib

Abutment

Grade beams

A' Hill affected by the landslide

70.00

Section A–A' 0.70

20.80 1.20

9.10

9.10

1.00

2.80

1.80

0.40 Precast girders

0.70 Deck

Transverse beam

Section B–B' 0.40 0.80

B 0.49

3.20

6.00

0.40

0.14

B'

La-Guaira

Caracas

Chamber at the top of the pilasters1 I-shaped column

Arch rib 3.10

0.18 0.24

3.20 2.40

1.00

Rocker well

0.18 5.20

3.20

5.20

3.20

Fi g. 1: The Viaduct 1 Caracas–La-Guaira Highway.1,2(unit: m)

almost constant depth ranging from 2.85 to 3.10 m along a distance of approximately 20 m away from the crown, where the reduction of depth initiated. At the crown, the depth of the ribs was 2.0 m. Thickness of the bottom slab was kept to a minimum (240 mm) to alleviate the load imposed on the falsework. The thickness of the web and top flange were 180 and 490 mm, respectively.

The structure rested on shallow raft foundations except for three inclined cast-in-place shafts, 29 m long and 2.20 m in diameter, which transferred the arch thrust to the soil on the Caracas side and seven vertical shafts, 29 m long, 1.90 m in diameter, which supported the pilaster (Fig. 1). Deep foundations extended 4 m into the rock.

The pilaster and the springing of the arch on the La-Guaira side rested on a pear-type hollow structure that transferred the loads to the soil through a 800 mm deep concrete slab, which was connected to the ribs through three inclined walls (as shown Fig. 1) and five walls placed in the longitudinal directions (not shown in the figure): two exterior and three

interior walls. The entire assembly was post-tensioned in the three directions.

Most of the bridge components were precast elements, including the Freyssinet hinges, I-shaped columns, prestressed girders and transverse beams connecting the top of the Ishaped columns. During construction, a cable way facilitated the translation of the precast elements and minimized the use of equipment.

After the construction of the foundations and pilasters was completed, the erection of the ribs was initiated using timber falsework and the cantilever method. Three cast-in-place concrete rib segments were concrete-poured using the cantilever method to complete the erection of the ribs up to the quarter points (the entire crosssection of ribs was constructed this way, including web and flanges). The central span of the ribs was cast-inplace with the use of a central falsework attached to the previously poured cantilever segments (Fig. 2).

The construction sequence is summarized in Refs. [4,5] and deserves special attention:

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• After the precast Freyssinet hinges were placed at the base of the ribs, the falsework for the first cantilever segment was suspended by four inclined cables, which were anchored into the front face of the pilaster at mid-height.

• For the other two ribs segments constructed by the cantilever method, one end of the falsework was connected through a hinge to the falsework of the previous segment, and a single inclined cable, anchored to the pilaster, held the free end of the falsework. By adjusting the length of the cables, it was possible to adjust the geometry of the ribs.

• The back face of the pilaster was tied back to the ground to avoid overturning.

• After the ribs were constructed up to the quarter points, the remaining half rib or central span was cast-inplace. A central falsework (81.4 m, Fig. 2) was assembled on the ground and lifted by electrical winches located at the end of the cantilevers.

• The bottom flanges of the ribs were then concreted, and the hinges between falsework segments were Scientific Paper

381

F ig . 2: Construction of the Viaduct 1

•

•

•

•

•

•

blocked, except for the hinge between the central falsework and the third segment (the last segment constructed using the cantilever method).

Then, three of the four inclined cables holding the first segment constructed by the cantilever method were released before concreting the web on the central falsework section. Only the cables holding the intersections of segments 1 and 2, 2 and 3 and segment 3 with the central falsework were active.

After concreting of the web of the central span was completed, all cables were uninstalled. At this point, the concrete arch ribs, except for the top flange of the central span, carried their own weight and that of the falsework.

Subsequently, the central falsework was lowered as well as the falsework of the cantilever segments (Fig. 2). I-shaped columns and transverse beams were constructed on the main span before concreting the top flange in the central span section of the ribs.

Next, prestressed girders were placed in a symmetrical and simultaneous manner on both sides of the crown.

Before completing the construction of the ribs, a narrow open joint at the crown allowed the insertion of Freyssinet flat jacks, in combination with concrete wedges, to adjust the pressure line and possibly minimize the long-term effects of creep.

Finally, the deck concrete was poured and prestressed longitudinally and transversely.

Stresses in the range of 0.6 to 8.50 and 0.7 to 7 MPa (all compression) were expected at the bottom and top 382

Scientific Paper

fibbers of the ribs,6 respectively, and thus, no reinforcement was provided.

More details about the structure can be found in Refs. [1,2,4,5,7].

Geology of the Area Aerial photographs taken in 1936 suggest the existence of an ancient landslide, initially estimated as 500 m × 225 m, on the Caracas side of the viaduct.8–13 The rock mass below is a sequence of Jurassic–Cretaceous metamorphic limestone, interfoliated with calcareous mica schist, quartzite and some amphibolites.9 A fault runs east– west, parallel to the first 4 km of the highway. It belongs to the TacaguaAvila fault system, which is very active in producing small earthquakes within the central regions of Venezuela.10

Activation of the Ancient Landslide In March 1987, both expansion joints at the abutments were found completely closed, restraining the expansion of the deck. Additionally, an asphalt bump was noticed, and extensive cracking of the grade beams on the Caracas side was observed, with crack widths exceeding 2 mm. A complete investigation was conducted.

Based on the interpretation of the physiographic evidence observed in the aerial photographs, it was evident that the cause of the problems was the reactivation of the ancient landslide. It was also found that the soil mass movement affected the structure at a 45 angle (Fig. 3a).

An extensive surveying program was implemented with three control points located on the La-Guaira side (stable) and 15 points on the unstable area, including points on the deck. The slide caused displacement at an

approximate rate of 10 to 30 mm/ month in 1987, decreased to 10 to 15 mm/year from 1988 to 1993 and increased exponentially to 10 mm/day in 2005. The complete history of time versus longitudinal displacement measured at the base of the ribs on the Caracas side is shown in Fig. 3b. It is worth mentioning that a 1 mm longitudinal displacement, monitored at the base of the ribs on the Caracas side, translated into 1 mm upward displacement at the crown and a 0.5 mm longitudinal displacement of the entire deck toward the La-Guaira side, which lead to the upward displacement profile of the ribs shown in Fig. 3c. Note that an extrapolation of the data shown in Fig. 3b suggests that the landslide was reactivated in 1967, the year of the Caracas earthquake (Mw = 6.5, Richter scale). The active sliding landmass in 2005 was estimated at 4.3 million m3.12 The picture shown in Fig. 3d was taken weeks before the collapse and shows more than 1500 mm of upward displacement at the crown.

Effects of the Landslide Thrust on the Structure and Rehabilitation Measures The landslide progressively shortened the span length of the unreinforced arch, inducing tensile stress at the top fibbers. A study summarized in Ref. [6] concluded that cracking and collapse of the ribs would occur at approximately 600 mm of longitudinal displacement, measured at the base of the ribs. The study also recommended the closing of the structure to traffic at a longitudinal displacement of 300 mm.

Changes in the authorities of the Ministry of Transportation over two decades made it difficult to ensure continuous monitoring of the structure. It is probable that many of the authorities underestimated the magnitude of the problem. Repairs were conducted on the structure on needbasis, only when problems were encountered, as for example the emergency work undertaken during the almost 2 years before the collapse.

All works were conducted without virtually any traffic interruption. The main features of these rehabilitation works included: • Installation of passive and active soil anchors Approximately 245 grouted passive anchors with lengths ranging from 30

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(a) Plan view

Pilaster

Longitudinal displacement

Crown

Pilaster

Transverse displacement ≈ 45º

Landslide thrust

La-Guaira

(c)

(b)

Upward displacemnt of the ribs (mm)

1750

1250

H G F

1000

E

1500

750 C

500

D

B

2005 2007

2000

1995

1990

1985

A 1980

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0

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Landslide displacement at the base of the ribs (mm)

2000

Caracas

(d) 2000 1750 1500 1250 1000 750 500 250 0 100 125 La-Guaira

150

175

200

225

250 Caracas

275

Progressive (m)

Year A. July 291, 1967, 0 mm Possible landslide reactivation (1967 earthquake) B. July 07, 1987, 160 mm Viaduct damages are observed. Monitoring begins C. June 08,1999, 350 mm Velocity aproximately constant D. December 13, 2004, 470 mm Velocity increases E. November 02, 2005, 820 mm New increase in velocity F. January 04, 2006, 1160 mm Viaduct closed to traffic G. March 06, 2006, 1220 mm H. March 17, 2006 1390 mm Last measurement

March 17, 2006 January 5, 2006: Ribs fractured January 4, 2006 November 4, 2005 May 4, 2005 July 7, 1987

Fi g. 3: Deformations of the structure. (a) Landslide vector and plain view of the structure; (b) time versus longitudinal displacement at the base of the ribs; (c) profile of upward displacement of the ribs and (d) upward displacement of the deck (picture taken weeks before the collapse)

to 26 m were installed in 1989 over an area of approximately 2000 m2. It is worth mentioning that many of the anchors required an unusual amount of cement; a few of them took up to 2000 bags (425 N per bag).13 A total of 20 active anchors at an inclination of 30 were also installed at the base of the Caracas abutment and at the base of the two intermediate piers on the Caracas side. The force in the anchors varied between 450 and 600 kN. Of the 20 anchors, 11 were placed (anchor horizontal projection) parallel to the structure, and the remaining 9 were installed parallel to the horizontal component of the landslide movement. Anchors were designed to try to deviate a portion of the thrust away from the bridge and not to stop the landslide because their strength was only a small fraction of that required to stabilize such a large unstable soil mass. Anchors probably helped to maintain the integrity of the soil mass surrounding the foundations of the structure on the Caracas side but showed little effectiveness in deviating the landslide as initially intended.

• Controlled demolition of the deck to open the gap at the expansion joints The gap at the expansion joints at both abutments repeatedly closed due to the displacement of the deck toward the La-Guaira side, compromising the integrity of the bridge. The rib–deck intersection at the crown generates an important modification in the way the ribs deform under the longitudinal displacement demand imposed by the landslide. This is because the deck that is connected to the ribs at the crown (Fig. 1) acts as a horizontal strut if the gap at the expansion joint is closed on the La-Guaira side. Thus, the longitudinal displacement of the half-rib portion located toward the La-Guaira side is restrained by the deck. The other halfrib, located on the Caracas side, absorbs the totality of the imposed longitudinal displacement at the base because the crown cannot move toward La-Guaira.

Consequently, the ribs bend in double curvature under longitudinal displacements and experience a significant increase in the bending stiffness.

Results from computer simulations

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suggested that axial stresses up to 14 times higher were likely to occur in the ribs if the deck cannot displace toward La-Guaira (the gap at the expansion joint is closed) when compared with the expected stresses under identical circumstances, except for an open gap.

To solve this problem, a short segment of the deck near the expansion joints was demolished repeatedly, especially at the La-Guaira side, to release the restrain in the longitudinal direction imposed on the ribs by the deck in the vicinity of the crown. This proposal was accomplished with the use of hydraulic jacks reacting against a corbel, which was attached to the deck at one end of the jack and against the abutment at the other end before the demolition of both expansion joints.

After the demolition of a small segment of the deck was completed to create a 200 to 300-mm gap, the force in the jacks was simultaneously and slowly released to permit a progressive unloading of the axial forced sustained by the deck, thus allowing the relaxation of the ribs and the bending in Scientific Paper

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single curvature. This methodology was first introduced in 1987 with the use of eight jacks stressed to 1250 kN for a total axial force of 10 000 kN applied to the deck at each abutment.

• Installation of steel trusses on the Caracas side to allow the recovery of transverse deck displacement The transverse landslide displacement of the arches and the approach bridge on the Caracas side induced shear in the transverse direction, out-of-plane bending and torsion on the ribs, which added to the effects of the longitudinal displacement. In order to reduce part of the effects caused by the transverse displacement, a mechanism was designed to uncouple the approach bridge on the Caracas side from the transverse movement of the soil mass.

The design included steel trusses resting on rollers, which allowed displacements in the transverse direction of the structure. The truss–roller assembly was placed at the top of the reinforced concrete redundant piers built in 1993 on either side of the original piers, as shown in Fig. 4a and b. An additional truss on rollers was placed very close to the Caracas abutment (Fig. 4c). Inside the top chamber of (a)

the pilaster located on the Caracas side, a series of steel paired struts were installed to support the precast girders, as shown in Fig. 4d. The ends of the prestressed girders were connected through a top transverse reinforced concrete beam, which transferred the load to the paired struts. As shown in Fig. 4d, these struts on rollers also allowed transverse displacements and rested on top of the bottom slab of the chamber, which was heavily reinforced to resist the loads from the prestressed girders.

A series of European Standard Beams (IPN 80), spaced at 300 mm in the transverse direction of the bridge, were placed on top of the existing slab, which supported a row of IPN 120 spaced at 400 mm in the longitudinal direction, for a total depth of 200 mm. After the mesh of steel sections was placed, and the top flange of the IPN 80 was fillet-welded to the bottom flange of the IPN 120, concrete was poured to a depth of 300 mm, forming a stiff and strong composite slab. The set of trusses and struts previously described remained as a passive redundant system for gravity loads until 2005.

A wall-strut-cable device was designed to release the transverse (b) Truss

Truss Top Chamber

Original Piers Truss

Redundant Pier

Redundant Piers

(c)

(d) Transverse Beam Prestressed Grider

Paired Struts Roller

Abutment Truss on rollers

F ig . 4: Redundant piers, steel trusses and struts. (a) Redundant piers built in 1993; (b) redundant piers and steel trusses; (c) truss on rollers close to the Caracas abutment and (d) paired steel-struts inside the top chamber 384

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deformation of the deck. It was clear that only the elastic portion of the deformation was recoverable. Permanent deformations, such as those induced by creep and nonlinear material behaviour, were unrecoverable.

This device was built adjacent to the two piers located on the Caracas side and included three elements (Fig. 5a): • Two reinforced concrete squat walls, built on either side of the original piers.

• A horizontal steel strut, connecting the three original I-shaped columns at the bottom base just above the Freyssinet hinge.

• A cable system consisting of two sets of 21 low-relaxation strands (16 mm in diameter), each placed in three 7-strand tendons wrapping around the old piers in a double Ushape, which were anchored at the ends of the squat walls. Tendons overlapped inside the squat wall.

Strands were initially stressed to 30% of their ultimate capacity as a passive system. This device was activated in 2005 to recover the elastic portion of the transverse deformation of the deck. For this purpose, it was necessary to transfer gravity loads from the original I-shaped columns, abutments and girders inside the top chamber of the pilaster to the trusses and struts.

The solution was to cut the base of the I-shaped columns below the Freyssinet hinge with a diamond saw (Fig. 5b) and to cut the bottom hinge of the rocker wall located inside the pilaster (Fig. 5c). After the demolition was completed, the portion of the deck located between the pilaster and the abutment on the Caracas side effectively rested on rollers. Immediately after the cut was completed, the wall-strut-cable device restrained the deck movement in the transverse direction. Then, by slowly releasing the tension in the cables, it was possible to activate the roller displacement and to recover approximately 250 mm of the transverse deck deformation, as shown in Fig. 5d (measured at the pier closest to the pilaster). The entire operation was completed with almost continuous traffic on the structure, except for brief interruptions at night, ranging from 30 min to 2 h.

• Prestressing of the ribs and the design of a plastic hinge at the crown Concrete tensile strength in the range of 2.4 to 2.8 MPa was obtained from

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cylinder split tests of the structure.

According to the structural analysis, such stresses could have been attained near the crown at a longitudinal displacement of approximately 500 mm, measured at the base of the ribs (close to the 600 mm reported in Ref. [6]).

The eminent cracking of the ribs forced emergency works to strengthen the structure in early 2005.

between the prestressing cables and the ribs. Old and new concrete were connected at these locations using 12 mm reinforcing bars as shear keys, which were embedded into the web of the box section with an epoxy adhesive in previously drilled holes.

The web thickness of the ribs was increased from 180 to 430 mm in the sections below the piers at which tendons experienced an angular change in order to facilitate the force transfer

A prestressed concrete section can exhibit significant rotation ductility and a behaviour similar to that of a plastic hinge, as observed in reinforced concrete structures.14 Note that a plastic hinge at the crown and the two Freyssinet hinges at the springing of the ribs, form a statically determinate structure, insusceptible to stress increments due to longitudinal landslide displacements15,16 (Fig. 7a and b). Following this reasoning, a plastic hinge was designed. The prestressing and mild reinforcement were bonded to the arch ribs. Yielding of the hinge reinforcement was self-induced by the moment increment at the crown produced by the rib shortening and was expected at nearly 800 mm of longitudinal displacement. At the crown, the thickness of the bottom slab and webs of the ribs were increased by 260 and 370 mm, respectively, and were heavily reinforced, using shear keys to connect the old and new concrete in order to prevent concrete crushing and to increase the plastic hinge’s rotation capacity (Section B–B0 in Fig. 6a and c). An insight of the plastic

(a)

(b)

The arch ribs were strengthened with mild and prestressing reinforcement.

Two small concrete beams were added to each rib near the top corners of the box section (Fig. 6a and b). These beams encased the mild and prestressing reinforcement and were connected to the ribs using 16 mm reinforcing bars embedded in the concrete with the aid of a rotary hammer to drill the holes required to anchor the rebar (shear keys), with the use of an epoxy adhesive. Each small beam encased five 25 mm, Grade 420 MPa reinforcing bars and two set of tendons, each of which had 30 and 22 strands, stressed to a total force of 5880 and 4100 kN, respectively, as shown in Fig. 6a. Strands were low relaxation, 6 mm in diameter and made of Grade A 1860 MPa steel.

Hinge Horizontal Struts

Squat Wall

Original Pier

U-Shaped Cables

Diamond Saw

Redundant Pier

(c)

(d) Paired Struts

250 mm

Downstream Hinge Demolished

Fi g. 5: Recovery of transverse deformations. (a) Wall-strut-cable device; (b) cut at the base of the I-shaped column with a diamond saw; (c) cut at the bottom hinge of the rocker wall inside the chamber of the pilaster and (d) recovery of 250 mm of transverse displacement of the deck (picture at the pier closest to the pilaster)

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hinge rotation capacity was obtained through moment-curvature relationships. However, it is worth mentioning that uncertainties related to the material properties, plastic hinge length and the stresses in the prestressing steel were very high, and it was found that the analysis is quite sensitive to the assumptions made. For example, a plastic hinge length of 1.5 m (approximately 3/4 of the rib depth at the crown) yielded plastic rotations of approximately 1.8% (lower bound) to 4.5% (upper bound), which translated into 0.54 and 1.35 m of longitudinal displacement measured at the base of the ribs on the Caracas side, respectively. More information related to the behaviour of plastic hinges in prestressed concrete members can be found in Refs. [14, 17, 18].

These proposals aimed at extending the life of the bridge until the completion of a new viaduct founded on competent soil. Unfortunately, the landslide velocity increased exponentially in 2005, which led to its collapse before the completion of the new structure.

Behaviour of the Structure Landslide velocity increased exponentially in 2005 up to 10 mm/day. In November 2005, the structure was still in service (Fig. 7d), without alternate roads to accommodate 50 000 vehicles per day, in each direction, on average.

Despite all the efforts made to extend the service life of the structure, it was essential to estimate a potential day when collapse was likely to occur.

The methodology summarized and explained in Refs. [19–21] in which the time is plotted versus the inverse of the velocity was used. Collapse was found to occur when the inverse of the velocity intersects the time axis. An extrapolation of the data led to the date of 16 December 2005. However, from 11 December to 6 January, the rate of movement remained nearly constant at 10 mm/day while the structure was still in service.

Longitudinal displacement of approximately 1100 mm at the base of the ribs was measured in January 2006, probably the most critical month before the collapse. During the first week of the month, rainfall of more than 107 mm accumulated in 3 days, resulting in a sudden landslide displacement (approximately 250 mm of displacement was measured at specific points on the hill) that closed the gap at the expansion joint on the La-Guaira side, Scientific Paper

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(a)

La-Guaira

structure was closed. Traffic was diverted through the old road, and a 40-min drive from Caracas to LaGuaira and vice versa became an 8-h journey due to the relatively high traffic volume for such a small road.

Caracas

Plastic hinges B B′ A A′

Dead end anchorage

P-4100 KN

Dead end anchorage

P=5880 KN Section A–A′

Hinge

Hinge

Section B–B′ Small beam

2.40

4.00

2 Tendons

2 Tendons

.18

2.00

.37

.49

.26

2.58

1 with 22 6 mm strands + 1 with 30 6 mm strands

3.20

.24 3.20

(b)

(c)

The thrust continued at a rate of approximately 8 mm/day. On 19 March 2006, the bridge collapsed before the completion of the adjacent structure (Fig. 7f). The maximum net or absolute upward displacement measured at the crown of the ribs before the collapse was approximately 1800 mm. No casualties or injures were associated with the failure of the structure.

An interesting quote from Freyssinet in Ref. [7], captures the very important aspects of this case history that every engineer shall always remember: “no one knows better than I that risks are inherent in every enterprise no matter what precautions are taken, since human beings are always fallible and nature is full of the unexpected.”

Conclusions and Lessons Learned F ig . 6: Reinforcement of the ribs. (a) Prestressing of the ribs; (b) installation of the concrete beam inside the ribs and (c) reinforcement at the crown of the ribs and concrete casting (unit: m)

thus restraining the longitudinal displacement of the crown. In consequence, the statically determinate structure (Fig. 7b) became a half-rib (a)

statically indeterminate one (Fig. 7c).

The concrete ribs fractured in a section located 35 m from the crown, towards the Caracas side (Fig. 7e), and the (b)

(c)

Plastic Hinge

• The importance of conducting extensive geological and geotechnical investigations should be emphasized. Sufficient resources and time should be allocated during the design phase to obtain high-quality aerial photographs and to conduct a complete study of the surface Expansion joint closed

x

Deck

x La-Guaira

(d)

35.00 Fracture of the ribs

x

Caracas

La-Guaira

Caracas

(e)

La-Guaira

Caracas

(f ) La-Guaira

Caracas

Caracas

La-Guaira

Fracture of the ribs

F ig . 7: The structure before and after the rehabilitation and during the collapse. (a) Original structure (statically intermediate); (b) plastic hinge at crown (statically determinate); (c) deck restraining rib deformation; (d) the viaduct in service; (e) fracture of the ribs and (f ) collapse of the bridge (Photo courtesy: A. Fonseca) (unit: m) 386

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•

•

•

•

•

•

geology, in addition to the traditional standard penetration and/or cone penetration test.

Major infrastructure should be continuously monitored to identify problems as soon as they appear. It is likely that the landslide was activated in 1967, the year of the Caracas earthquake, but it was detected around 20 years after.

Landslides of large magnitude, which cannot be stabilized by traditional methods, such as tie-back walls and/or pile curtains, are valid reasons to modify the route of a future highway.

Arch structures similar to the Viaduct 1 can withstand significant longitudinal and transverse displacement demands at their bases given that sufficient reinforcement is provided to resist tensile stresses.

Freyssinet hinges exhibited excellent behaviour during the service life and rehabilitation of the structure. Designers are invited to investigate the mechanics behind these devices and to use them when it is considered appropriate.

Engineers are encouraged to revise the construction process of the viaduct to suit applications in projects with limited budget and equipment located in regions with relatively economical labour. Similar construction procedures can be implemented today by which precast elements allow the optimum use of falsework, and only minor construction equipment is required.

Moment–curvature analyses provide a powerful tool for investigating the plastic rotation capacity of prestressed concrete sections.

Because of the high degree of uncertainties in existing structures related to material properties, plastic hinge length, loading magnitudes and stresses in the prestressing steel, the upper and lower bound estimates of plastic rotation are recommended as a potential tool for decision making

and not as an instrument for the prediction of behaviour.

• Landslides behave erratically. This experience suggests that it might not be safe to rely on long-term behaviour estimates based on past monitoring.

• Forecast theories can be useful tools to help make better short-term decisions in cases where limited information is available but should be used with extreme caution.

Acknowledgements Rosendo Camargo Mora made the main contributions to extend the life of the structure and designed most of the rehabilitation measures, with the author working as his assistant engineer. Daniel Salcedo played a key role conducting the geological investigation and articulating the team of specialists. They deserve special credit.

Andres Pesti, Celso Fortoul, Claudio Ramírez Corredor, Claudio Ramírez Soto, Eli Abadí, Isaac Salcedo, José Capobianco, Josefina Ortas, Mathías Brewer, Nelson Rodríguez and Salvador Pulido provided valuable support. The workers from PILPERCA carried out the repairs in very tough conditions. The author thanks Laura Dávila Carrasco, Teresita León and Peter Weatherer for help in the preparation of this manuscript.

References [1] Muller J. Largest concrete spans of the Americas. How the three bridges were designed. Civ. Eng. ASCE 1953: 45–49.

[2] Freyssinet E. Le pont de Candelier. Annales des Ponts et Chaussées 1923; 1 (in French): 165.

[3] Schacht G, Marx S. Concrete hinges in bridge engineering. Eng. Hist. Herit. 2015; 168(EH2): 64–74.

[4] Podolny W, Muller J. Construction and Design of Prestressed Concrete Segmental Bridges John Wiley Sons: New York, 1982.

[5] Shama R. How they were built. Civ. Eng. ASCE 1953: 50–55.

[6] Jean Muller International.

Viaduct No. 1 sobre la Autopista Caracas La-Guaira Highway. Reporte técnico para el Ministerio de Transporte y Comunicaciones de Venezuela, 1988.

[7] Freyssinet E. Largest concrete spans of the Americas. Three monumental bridges built in Venezuela. Civ. Eng. ASCE 1953: 41–44.

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[8] Salcedo D. ¿Es predecible el comportamiento del macrodeslizamiento que afecta al Viaducto No. 1 de la Autopista Caracas-LaGuaira? Boletín Sociedad Venezolana de Mecánica del Suelo e Ingeniería de Fundaciones. N 58, Caracas, 1989.

[9] Salcedo D, Ortas J. Investigation of the slide at the southern abutment hill of Viaduct No. 1. Caracas–La-Guaira Highway. Venezuela.

Proceedings of the 6th International Symposium on Landslides, Christchurch, 1991.

[10] Salcedo D. Observaciones geológicas en galerías exploratorias excavadas en la zona del Viaducto No. 1. Autopista Caracas–La-Guaira.

IV Congreso Suramericano de Mecánica de Rocas, Santiago de Chile, 1994.

[11] Salcedo D. El deslizamiento de la ladera sur de la autopista Caracas–La-Guaira. Ingeniería forense y estudios de sitio. Guía para la prevención y gestión de riesgos. Ediciones Cetici, Caracas, 2006.

[12] Salcedo D, Ortas J. Comportamiento del deslizamiento de la ladera Sur del Viaducto No. 1. Autopista Caracas–La-Guaira. Venezuela. Memorias I Simposio Panamericano de Deslizamientos, Guayaquil, 1994.

[13] Salcedo D. Behavior of a landslide prior to inducing a landslide failure. Eng. Geol. 2009; 9(1): 16–30.

[14] Naaman A, Harajli M, Wight JK. Analysis of ductility in partially prestressed concrete flexural members. PCI J. 1986; 31(3): 64–87.

[15] Camargo R. Reforzamiento del Viaducto 1 de la autopista Caracas–La-Guaira. Ingeniería forense y estudios de sitio. Guía para la prevención y gestión de riesgos. Ediciones Cetici, Caracas, 2006.

[16] Fargier L, Camargo R. Rehabilitation of a viaduct subjected to a landslide thrust. IABSE, 17th Congress, Chicago, 2008.

[17] Burns N. Moment curvature relationships for partially prestressed concrete beams. PCI J. 1964; 9(1): 62–63.

[18] Suri KM, Dilger WH. Steel stresses in partially prestressed concrete members. PCI J. 1986; 31(3): 88–112.

[19] Fukuzono T. A new method for predicting the failure time of a slope. Proceedings of the IV International Conference and Field Workshop on Landslides, Tokyo, 1985; 145–150.

[20] Voight B. Materials science law applies to time forecasts of slope failure. Landslide News 1989; 3(14): 8–11.

[21] Voight B. A relation to describe ratedependent material failure. Science 1989; 243(4888): 200–203.

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Determination of Defect Factors for a Road Bridge Made of Prestressed Concrete Arkadiusz Madaj, Prof.; Wojciech Siekierski, PhD, Faculty of Civil and Environmental Engineering, Pozna n University of Technology, Pozna n, Poland. Contact: wojciech.siekierski@put.poznan.pl DOI: 10.2749/101686617X14881932436735

Abstract The paper concerns determination of defect factors for a road bridge made of prestressed concrete. Cracking and permanent deformations of the continuous triple-span road bridge made of prestressed concrete were reported by road administration. Inspection of the bridge was carried out. Bridge design and erection documentation were researched. Hypotheses concerning defect factors were put forward. Parametric analysis of the bridge was carried out. The analysis showed that the cause of observed defects was faulty design of prestressing, namely: underestimation of prestressing force losses, underestimation of the influence of concrete shrinkage, and inappropriate tendon layout which introduced rapid changes to prestressing force. The general conclusions are: (a) symmetric prestressing of continuous main girders should be applied in zones where sagging and hogging bending moments occur in turns, (b) long prestressing tendons should be applied with care due to substantial prestressing force losses that may occur, (c) anchoring numerous tendons at the same crosssection should be avoided since it results in rapid variation of prestressing force, (d) variations of structure deformations due to prestressing should be recorded and checked against the predicted ones.

Keywords: prestressed concrete bridge; defect factor; static-strength analysis.

(cross-section B shown in Fig. 4) over the abutments, through twin T-beam girders of slender webs (cross-section A shown in Fig. 3) in the spans to singlecell box-girder (cross-section C shown in Fig. 4) over piers. Symbol D in Fig. 2 denotes zones where the span crosssection changes between cross-sections A (Fig. 3) and C (Fig. 4). Neutral axis of cross-section A is about 500 mm higher than neutral axis of cross-section C. The change of the neutral axis level in the vicinity of the interface of C and D cross-sections is rapid. It is followed by a rapid change of the bending moment due to prestressing.

Cross-beams are made of reinforced concrete. So is the deck slab, which has variable thickness (see Figs. 3–5).

Bridge Inspection Results Introduction The process of determining factors of structural defects recorded usually starts with the inspection of available erection documentation.

Then hypotheses concerning the factors are put forward.

Verification of the hypotheses may require various activities: further detailed inspection, auxiliary testing and/or computational analysis. At present, finite element method (FEM) analysis software is able to deal with the complex phenomena of prestressed concrete.1,2 However, sometimes simplified modelling3 is capable of reflecting factors of structural behaviour.

Together with the choice of the hypotheses verification method, the criteria that allow one to make a final judgment have to be established. In the case of the prestressed concrete bridge, such criteria may be taken as: Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: July 20, 2016 Paper accepted: January 30, 2017

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crack distribution and permanent vertical displacement distribution.

Both phenomena are being widely researched4–7, and usually result from insufficient prestressing.8,9 A hypothesis is considered proven if the computed distribution of cracking and deflection, based on assumed lack of prestressing force, complies with the observed structural defects.

The approach for determination of defect factors is presented in this paper. Development of serious cracks in the main girders and permanent span deflections were observed in a triple-span road bridge made of prestressed concrete. This was followed by an attempt to determine the defect factors.

Bridge Description The analysed bridge is a twin-girder triple-span structure made of prestressed posttensioned concrete. Span theoretical lengths are: 57.3 + 57.3 + 57.3 m. General view of the viaduct is shown in Fig. 1 and longitudinal layout in Fig. 2.

The bridge cross-section changes from twin T-beam girders of stocky webs

Vertical cracks in both girders were recorded during inspection. Location and morphology of the cracks are shown in Figs. 6 and 7, respectively.

The cracks were found in the vicinity of the piers. They extended up to half the depth of the girder. In the absence of service loads, the cracks were about 0.2–0.4 mm wide at the bottom edge of the girder. An example is shown in Fig. 8.

Permanent deflections of all spans were observed during bridge inspection. Their magnitude was also recorded (see Fig. 9).

Inspection of Available Documentation The available documentation, including design documentation and reports created during bridge erection, were inspected.

The bridge was designed according to Polish standards.10,11 The load and resistance factor design was applied.

It was built in 2002 while the bridge inspection described above was done in 2015.

The concrete superstructure of the bridge was completed in a formwork

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Fi g. 1: General view of the viaduct

on scaffolds. The structure is made of C40/50 concrete. The tendon layout applied is shown in Fig. 10. Tendons 1–3 (continuous lines in Fig. 10) run along the entire bridge length. Tendons 4–6 (dashed lines in Fig. 10) are situated in the terminal spans while tendons 7–13 (dash-dotted lines in Fig. 10) are situated over the piers, in two groups: longer (7–10) and shorter (10–13). As is shown in Fig. 10, in general, tendons are straight. Curves are used only to create a turn of tendon axis.

Reinforcement of main girders at the cracked cross-sections was also studied. According to the final prestressing design documentation, the degree of longitudinal reinforcement in tension at the cracked cross-sections was about 0.04%, i.e. 10 times smaller than the minimum required by the Polish standard10 for prestressed concrete bridges (0.4%). At the same time, the degree of transverse reinforcement (stirrups) at the cracked cross-sections was just above the required minimum.10

Total length of tendons 1–3 exceeds 170 m. For active anchorage at both tendon ends (as it was applied) the active tendon length, in terms of prestressing force losses, is over 85 m. In the literature on prestressed concrete one may find general suggestions to avoid such tendon length.8

It was also found that the span deflections observed during inspection had been reported earlier, just after formwork dismantling. The intended bridge profile line was not achieved.

Observation during the service period proved that the deflections were stable.

The available data on prestressing, found in the erection documentation, included tendon type (the cross-section) and initial prestressing force for each tendon. The prestressing report suggested that each tendon was prestressed up to its nominal capacity.

The particular tendons and their nominal capacity are: tendons 1–3: 19T15S, 0.75 fpk = 3976 kN and tendons 4–13: 25T15S, 0.75 fpk = 5231 kN. In general, the tendon layout was characterised by rapid changes of prestressing force, due to deep anchorage of group of tendons of large bearing capacity, and lack of uniform bearing of tendons against duct walls, due to limited length of tendon curvatures.

B

A

3,0

46,3

D

Method of Determination of Defect Factors The occurrence of cracking was assumed to coincide with the tensile stress at the girder bottom edge being greater than the tensile strength of concrete (ultimate strain exceeded). Since the reinforcement coefficient at critical cross-sections of main girders does not exceed 0.04%, tension stiffening does not occur and it was justified to assume that the member behaves similar to that made of unreinforced concrete8,12 (strain energy concentrates in a single cross-section, resulting in a single, relatively wide crack).

Distribution of prestressing force along the bridge also influences bridge deformations. Determination of prestressing force distribution can be done by comparing recorded vertical displacements of a structure after prestressing with computed vertical displacement, based on certain assumptions concerning prestressing force distribution.

Hypothesis on Defect Factors Based on in situ bridge inspection and analysis of available design and erection documentation, the following hypothesis was put forward: the observed defects (cracking and permanent bridge span deformation) are caused by faulty design of prestressing, underestimation of prestressing force losses, especially for tendons running along the entire bridge length (1–3, Fig. 10). This issue, combined with insufficient reinforcement near the bottom edge of main girders in the vicinity of intermediate supports, resulted in a few cracks of considerable width.

C

D

4,0 4,0 4,0 4,0

A

41,3

D

Computational Model FEM was used in the analysis of the bridge. The computational model is shown in Fig. 11. The model was created in the Structural Robot environment. Main girders, cross-beams and the deck slab were modelled with two-node linear beam elements of 6 degrees of freedom per node. Longitudinal beam elements represented main girders while transverse elements represented cross-beams and deck slab. The structure and member dimensions were assumed based on C

D

4,0 4,0 4,0 4,0

A

B

46,3

3,0

Fi g. 2: Variations in bridge cross-sections (Units: [m])

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220

0,278

0,216 1600

375 3250

264

500 19

940

700

500

254

2680

1600

2680

3450 6650

1921

1600

2380

700

1921

3250

300

232

0,000

168

0,031

3180

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F ig . 3: Cross-section A (see Fig. 2) (Units: [mm])

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F ig . 4: Cross-section B (see Fig. 2) (Units: [mm])

940

design documentation. Variations in bridge cross-section and neutral axis level near the piers were considered.

Slab thickness variations were also taken into account.

Flexural and torsional stiffness of beam elements were computed on the basis of structural dimensions. Influence of change in cross-section span (open twin girder to box-girder) on torsional stiffness of beam elements modelling the main girders was taken into account.

The following factors were taken into account: weight of structural members, weight of fittings, prestressing loads (including losses due to friction and concrete shrinkage), and live loads. Self weights and live loads were taken as per Polish standard (as in the design documentation). Prestressing forces were computed on the basis of erection documentation. Because of lack of data concerning the prestressing sequence, simulations of various cases were carried out.

Since the computational model involved beam elements, the prestressing effects were modelled as additional forces and bending moments.13 At tendon anchorage (red dot in Fig. 12a) the following effects were considered3: Horizontal component of prestressing force S: ð1Þ

Px = S cos⁡φ

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Vertical component of prestressing force S:

F ig . 5: Cross-section C (see Fig. 2) (Units: [mm])

1 GIRDERS 2

CRACKS

SUPPORT

SUPPORT

SUPPORT

D C B

E

SUPPORT

13,39

CRACK

A 9,97

14,30

F ig . 6: Crack location seen in bottom view of spans (walls of webs are marked with dashed lines) (Units: [m])

B

E C

166

146

142

175

D

50

138

268

A

F ig . 7: Crack morphology over the side walls of main girders (for reference see Fig. 6) (Units: [mm]) 390

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Py = S sin⁡φr −S sin⁡φl = 2 S sin

φO Δφ sin 2 2

ð5Þ

where S is prestressing force, Px horizontal component of prestressing force S, Py vertical component of prestressing force S, M bending moment caused by prestressing force S, and φ, φl, φr, Δφ are the angles as in Fig. 12.

Analysis of Normal Stress Distribution Since open cracks and span deflection were observed in the absence of service loads, static strength analysis was carried out for combination of selfweight of bridge spans and fittings as well as prestressing. To verify the hypothesis that the main cause of the observed defects was underestimation of prestressing force losses, parametric analysis was carried out.

Fi g. 8: View of the crack E (see Fig. 6)

ð2Þ

Py = S sin⁡φ

Bending moment caused by prestressing force S: M = Px e = S cos⁡φ el

ð3Þ

First, only time-dependent losses of the prestressing force were considered in the FEM analysis. The analysis was carried out for the average coefficient of long-term prestressing force losses, η = 0.85 (the relationship of final to initial value of prestressing force). Under such assumption, the dead loads and prestressing generated very small compressive normal stresses at the bottom edge of the main girders at the sections where the cracks were observed. Such circumstances resulted partially from the tendon layout and partially from variations of main girder cross-section

The layout of each tendon was given as a set of co-ordinates of turning points of the tendon axis. At these points (red dot in Fig. 12b) the following were considered3: Horizontal component of the prestressing force: Px = S sin⁡φl + S sin⁡φr φ Δφ = 2 S sin O cos 2 2

ð4Þ

Vertical component of the prestressing force:

–20

0

–40

Distance from support axis at abutment (m) 0

0

0

171,90

0

45

114,60

20

Vertical distance from Support line (mm) 30 57,30

40

0,00

60

–60 80

–80

Fig. 9: Deflections recorded during bridge inspection (piers at 57.30 and 114.60) 3200 2800 2400 2000 1600 1200 800 400 0

Distance from tendon axis to girder bottom edge (mm)

that introduced rapid change in neutral axis level where a twin T-girder crosssection transforms into a single-cell box-girder cross-section. It caused rapid increase of the prestressing force at the lever arm which, at bottom edge, magnified tensile stress due to prestressing.

Second, immediate loss of prestressing forces was assessed; data concerning the prestressing process suggested that such losses had not been taken into account while setting up prestressing forces to be applied during the bridge erection. Elastic shortening of concrete and friction of tendons against the inner surface of the duct were considered. The dates of tendon prestressing were taken from the prestressing reports and considered as basic data to establish probable prestressing sequence. Several possible cases were analysed. For tendons 1–3, which were assumed to be prestressed prior to all others, the computed prestressing force loss due to elastic shortening of concrete was 3–8%. Prestressing force loss due to friction of tendons against ducts walls was computed for all tendons, taking into account their length and curvature. The losses fell in the 10– 25% range.

Finally, it was established that total losses of prestressing force in tendons 1–3 may have reached over 45% in the middle span. Bridge static analysis was carried out taking into account immediate losses as well as timedependent losses of prestressing forces. It was assumed that total losses (a) vary along long tendons (1–3), (b) are constant along short tendons (4– 13) and (c) depend on tendon length and curvature. The following total coefficients of prestressing force losses were assumed: η = 0.85/0.75/0.60/0.50 (three changes between anchorage over abutments and beam symmetry plane) for continuous tendons (1–3), η = 0.75 for longer tendons over piers (7–10), η = 0.80 for short tendons in terminal spans (4–6) and for shorter tendons over piers (11–13), see Fig. 10 for tendon numbers.

no.11-13

Computational results show (Fig. 13) that the location of minimum compressive stresses coincides with the location of cracks observed in the middle span (Figs. 7 and 10). These cracks were the first to occur and now, after 14 years of service, they have the biggest width and length.

Fi g. 10: Prestressing tendon layout (half of the bridge is shown; pier at 57.30) and the recorded crack location

The stress diagram in Fig. 13 does not include the stress component due to

Cracks

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no.2

no.3

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no.5

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F ig . 11 : FEM of the bridge (a)

Δφ

(b) φ

. eɭ

φɭ

e

Neutral axis φr

Neutral axis O

L M

φo

P

Influence of standard live loads11 on bending moment in main girders was analysed. Standard live loads on road bridges, i.e. a single vehicle of 800 kN factored by a dynamic coefficient (1.064 for the analysed structure) and a uniformly distributed load of 4 kN/m2, were considered. The envelopes of total bending moments in the main girders showed that cracking occurred in the zones where sagging and hogging of bending moments may act in turns, due to live loads action.

According to the literature on prestressed concrete structures, in such zones symmetric prestressing should be applied.9

Py Px Px Py

F ig . 12 : Application of prestressing force effects in FEM a) at the point of tendon anchorage, b) at the point of tendon flexure 14,0

Normal stress (MPa)

Top fibre

12,0

Bottom fibre

10,0 8,0 6,0 4,0 2,0 171,90

57,30 (PIER)

0,00

Approximate location of observed cracks

114,60 (PIER)

Distance from support axis at abutment (m)

0,0

F ig . 13 : Diagram of computed normal stresses in extreme fibres of main girder under dead loads (shrinkage excluded); positive sign denotes compression 0,08

Displacement (m)

0,06

Formwork

Dead loads

Prestressing

Total

0,04 Recorded 0,02 171,90

114,60

57,30

0,00

0,00 -0,02

bottom edge of the main girders at the cracked cross-sections was about ten times smaller than the required minimum. No prestressing tendons were present near the bottom edge of the girders. Therefore the cracked cross-sections had to be analysed as if it was made from unreinforced concrete.

Moreover, the analysis of the envelopes showed that certain arrangements of live loads produced additional tension near the bottom edge of the girders at the cross-sections. In view of the above-mentioned lack of appropriate reinforcement, it posed a threat to overall safety of the structure.

It was concluded that the observed cracks were initiated by service loads; the latter, however, were not included in the analysis due to unknown intensity and spectrum.

Analysis of Span Deflection Distribution

-0,04 -0,05 Distance from support axis at abutment (m) -0,08

F ig . 14 : Diagram of computed and recorded in situ vertical displacement

concrete shrinkage. This stress component was estimated at cracked cross-sections separately.14,15 Concrete properties and erection technology details (time of concrete placing, time of formwork dismantling, time and sequence of prestressing force application) were taken into account.

It was found that it exceeds 1 MPa at the bottom edge of the cracked 392

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cross-sections. So total computed normal stress at the cracked crosssections was likely to exceed the concrete tensile strength.

Analysis of Cracked Crosssections As mentioned earlier, the degree of longitudinal reinforcement near the

Vertical displacements of the structure were analysed similarly as for stress analysis. The analysis results are shown in Fig. 14 together with the span displacements recorded in situ.

The concrete superstructure of the bridge was completed in a formwork on scaffolds. The applied formwork profile, based on the initial prestressing design, is given in Fig. 14; initially, the designed prestressing was meant to lift the middle span by over 60 mm (the “formwork” line).

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Considering only time-dependent losses, it was found that the prestressing reduced the total deflection in the middle span to about 10 mm. This result was incompatible with the observed deflection of the middle span that was about 80 mm.

Bearing the above considerations in mind, the analysis had to account for immediate and time-dependent losses of prestressing forces. The assumptions were similar to those for stress analysis.

The results are shown in Fig. 14. It can be seen that prestressing is not able to lift the level of intended bridge profile in the middle span (the “prestressing” line against the “formwork” line).

Comparison of total span displacements computed on the basis of the described analysis (the “total” line, Fig. 14) and the current span displacements (“recorded,” Fig. 14) suggests that total prestressing force losses may be even greater than those assumed—the dots of the “recorded” set are located “below” the “total” line.

Conclusions The presented analysis shows that based on static strength computations, it is possible to determine the causes of observed defects of a structure (e.g. a bridge). In the considered case, the defects occurred due to underestimation of prestressing force losses, underestimation of the influence of concrete shrinkage on internal force distribution and inappropriate tendon layout which caused rapid changes in prestressing force.

The analysed case shows that continuous main girders are sensitive to cracking, especially in the zones where sagging and hogging of bending moments occur in turns. In such zones, symmetric prestressing should be applied.9 In the analysed case, the zones turned out to be critical in

terms of sensitivity to cracking. Zones of a prestressed member where positive-only or negative-only bending moments occur are easier to protect against cracking.

For long tendons, significant prestressing force losses may occur due to elastic member shortening, friction of tendons against the inner surface of the duct and concrete shrinkage. This had happened in the analysed case. In general, long tendons should be applied with care and the length of tendons prestressed at one end should be limited to 60 m.8 In the analysed case, cracks occurred within the zones where a considerable number of tendons were anchored.

Rapid variation of prestressing force occurred. Anchoring numerous tendons at the same cross-section should be avoided since it makes it more difficult to enforce stress limitations in concrete.8 Variations of structure deformations due to prestressing should be recorded and checked against the predicted ones (i.e. formwork profile). In the studied case, the insufficient lift of bridge spans after prestressing indicated the problem concerning influence of prestressing on internal forces of distribution.

The existence of the problem was confirmed later by the occurrence of cracks.

Acknowledgements

[2] Okumus P, Oliva MG, & Becker S. Nonlinear finite element modeling of cracking at ends of pretensioned bridge girders. Eng. Struct.

2012; 40: 267–275.

[3] Machelski C. Modelling of Bridge Prestressing (in Polish), Dolno sląskie Wyd Edukacyjne: Wrocław, 2011, 148 pp.

[4] Akl A, Saiid Saiidi M, & Vosooghi A. Deflection of in-span hinges in prestressed concrete box girder bridges during construction.

Eng. Struct. 2017; 131: 293–310.

[5] Ruan X, Shi X, & Li X. Failure analysis of tendon breakout on bottom slab of a prestressed concrete box gird bridge during construction. Eng Fail. Anal. 2012; 25: 291–303.

[6] Park H, Jeong S, Lee S-C, & Cho J-Y. Flexural behavior of post-tensioned prestressed concrete girders with high-strength strands. Eng. Struct. 2016; 112: 90–99.

[7] Tong T, Liu Z, Zhang J, & Yu Q. Longterm performance of prestressed concrete bridges under the intertwined effects of concrete damage, static creep and traffic-induced cyclic creep. Eng. Struct. 2016; 127: 510–524.

[8] Holst R, & Holst KH. Brücken aus Stahlbeton und Spannbeton Ernst&Sohn: Berlin, 2014, 811 pp.

[9] Leonhardt F. Vorlesungen über Massivbau, Sechster Teil: Grundlagen des Massivbrückenbaues Springer Verlag: Berlin, Heidelberg, New York, 1979, 225 pp.

[10] PN-91/S-10042. Bridges. Structures of Concrete, Reinforced Concrete and Prestressed Concrete. Design (in Polish) PKN: 1992.

[11] PN-85/S-10030. Bridges. Loads (in Polish) PKN: 1985.

[12] Knauff M (Ed.). Basics of Design of Reinforced and Prestressed Concrete Structures according to Eurocode 2 (in Polish) Dolno sląskie Wyd. Edukacyjne: Wrocław, 2006, 976 pp.

The research was completed at the Institute of Civil Engineering of Pozna n University of Technology, Pozna n, Poland.

[13] Madaj A, & Wołowicki W. Design of Concrete Bridges (in Polish) WKŁ: Warszawa, 2010, 672 pp.

References

[14] Flaga K. Shrinkage Stress and Subsurface Reinforcement of Concrete Structures (in Polish). Series Civil Engineering Technical University of Cracow: Cracow, 2011.

[1] Yapar O, Basu PK, & Nordendale N. Accurate finite element modeling of pretensioned prestressed concrete beams. Eng. Struct. 2015; 101: 163–178.

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[15] PN-EN 1992. Design of Concrete Structures – Part 1–1: General Rules and Rules for Buildings European Union: Brussels, 2004.

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Forensic Investigations of Collapses of Buildings in “Aggregate”: Lesson Learnt from a Meaningful Case Study Fabrizio Palmisano, M Eng., PhD, F.IABSE, Adjunct Professor, Politecnico di Bari, PPV Consulting, Bari, Italy; Pasquale Perilli, M Eng., PPV Consulting, Bari, Italy. Contact: fabrizio.palmisano@poliba.it DOI: 10.2749/101686617X14881932436933

Abstract

aggregate, mainly because of the following reasons:

The article describes the forensic investigation of a collapse that occurred in a building “aggregate” in Barletta (Puglia, South Italy) on the morning of 3 October 2011. According to Italian technical standards, a building aggregate consists of a set of adjacent buildings that is the result of an articulated and not unitary genesis due to multiple factors. Some recent collapses, such as the one that occurred in Barletta, have highlighted that the approach to be used in the relevant forensic investigation is not so obvious and simple. The aim of the article is to describe both the different approaches used in the investigations by different consultants and their different conclusions in order to trace general criteria that can be useful in similar cases.

• Difficulties in finding documents of the original project • Difficulties in finding historical documents describing the construction sequence • Difficulties in conducting surveys and tests in buildings of other owners • Difficulties in the assessment of the mechanical characteristics of old construction materials (e.g. natural masonry blocks, “poor” quality mortar) • Absence of national/international codes relevant to the vulnerability assessment of old existing constructions in building aggregates

Keywords: forensic engineering; building aggregate; masonry building; demolition; vulnerability assessment.

Introduction In recent years, the feasibility of demolition and reconstruction interventions in urbanised areas (especially in historical city centres) has become a topic of great interest from a scientific and technical point of view mainly for the following reasons1: • Ageing of existing buildings and relevant need for functional renovation • Urban planning and environmental policies that limit the use of land in the suburbs • Poor quality of existing buildings that makes retrofitting to current standards expensive (e.g. seismic resistance, energy savings, architectural barriers, plants) • High-value and commercial attractiveness of central urbanised areas (already saturated with buildings) However, the obvious demands of urban renewal collide with the peculiarities of urban centres: • Heterogeneity of buildings, often made in different times with Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: July 28, 2016 Paper accepted: February 3, 2017

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successive modifications/enlargements by using different materials and construction techniques • Absence of documents of the original project and of the construction process • The need to identify possible interactions between adjacent buildings in order to ensure structural safety during and after the renovation works In this scenario, the present Italian technical standards (i.e. Ref. [2,3]) define the so-called building “aggregate” (Fig. 1). According to these standards, the aggregate of buildings is delimited by an open space, and it is as a set of adjacent buildings that is the result of an articulated and not unitary genesis due to multiple factors (e.g. different construction and modification ages, materials, uses, owners).3 Italian historical city centres are full of building aggregates.

According to the Italian technical standards2,3, in the analysis of a building (or of a part of a building) that is an element of an aggregate, it is necessary to take account of possible interactions arising from the structural contiguity with adjacent buildings.

The problem is that when dealing with historical building aggregates, it is quite impossible to have a deep structural knowledge of the whole

In this scenario, this paper analyses the different approaches and conclusions of the technical consultants who were involved in the investigations of a recent building collapse in aggregate due to demolition interventions. The aim is to trace general criteria that can be useful in similar investigations.

Methodology for Investigations According to Ref. [4], the fundamental answers that should be given at the end of the investigation are the following: • Status of the construction at the moment of the collapse • External loads acting at the moment of the collapse • Internal stresses or forces acting at the moment of the collapse • Resistance of the structure • Place where the collapse originated • Primary cause of the collapse • Contributing factors to the collapse • Trigger of the collapse • People responsible Moreover, according to Refs. [5–7], in cases similar to that under study the collapse is often disproportionate to the original cause; therefore in the

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investigation, a clear distinction between the cause of the initial failure (i.e. the triggering cause) and the factors that transform an initial failure into a progressive collapse has to be made because it could involve different responsibilities.

To find these answers, Ref. 8 has proposed a methodology that draws inspiration from Ref. [9] and that is mainly composed of the following steps: • Historical-kinematic reconstruction of the collapse based on questioning, video-photographic documentation and visual inspections of the ruins • Acquisition of technical data necessary for the numerical analyses: documentation to be found in public offices; documentation on the ongoing works; on-site geometrical surveys (also during the works of debris removal); visual sample inspections, laboratory and on-site tests; analysis of video-photographic documentation • Recalculations • Identification of design and construction errors and analysis of their influence on the collapse dynamics

The Collapse During the demolition work of a building (called building A henceforth; Fig. 2) in a building aggregate in Barletta (Puglia, South Italy), a huge part above ground of the adjacent building (called building B henceforth; Fig. 2) collapsed on the morning of 3 October 2011, causing five fatalities.

Building A had an underground floor and two floors above ground, while building B had an underground floor

and four floors above ground (the upper floor is not visible in Fig. 2).

Both buildings had vertical masonry structures and various structural typologies for the floors (i.e. masonry vaults at the underground and ground floors, Fig. 3; composite RC/steel beam-blocks or wooden beams at the higher floors). The demolition of building A was necessary in order to build a new building. The two buildings had one vertical wall in common (orthogonal to the main façades, called common wall henceforth). This wall was 800 mm thick, composed of two leaves (each 200 mm thick) of calcarenite blocks with a rubble stone fill. It was not subjected to demolition as it was the bearing structure of the floors of both buildings.

Figure 2 shows the condition of the two buildings in 2008 (i.e. before the beginning of the demolition works), while Fig. 4 shows the buildings after the demolition of the part above the ground of building A, and the partial collapse of the building B that occurred in 2011.

The demolition work of building A started in October 2008. It was performed by hand in accordance with the demolition design.

It is worth mentioning that at the beginning of the demolition, the structural conditions the buildings (i.e. B and D; Fig. 2) adjacent to building A were significantly different. In building B, apart from some hairline cracks due to ageing, there were some cracks due to an imperfect connection of the main façade wall with the vaults and the orthogonal walls. Conversely, in building D, there were diffuse and wide cracks in the walls and floors, typical of great distress from service

Fi g. 1: Example of an “aggregate” of buildings (the yellow line indicates the boundary of the aggregate)

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loads. The different crack pattern was also the consequence of a significantly different structural layout and dimensions of the main structural elements.

In December 2008, the demolition work was stopped because of the enlargement of the existing cracks in building D. Hence, in February 2009, building D was declared unusable by the municipality. Figure 5a shows the condition of building A in January 2010.

On 21 September 2011, the contractor declared to start only the removal of the rubble in order to make some detailed investigations in order to plan the strengthening work of the wall between buildings A and D. Actually, the contractor resumed the demolition work by using heavy mechanical machines in contravention of the demolition design. On 30 September 2011, huge cracks appeared in building B; this is why technical inspections were performed that day, on 1 October and also at the moment of the collapse (which occurred on 3 October 2011).

Phases of the Investigations Many technical consultants were involved in the investigations (for the Public Prosecutor’s Office, the Judge for the Preliminary Investigations, the parties).

The pre-trial investigations divided into two phases.

were

In the first phase (from the collapse to January 2012), they were chaired by the Public Prosecutor. In fact, in Italy, the Public Prosecutor has a duty to initiate the criminal proceeding and to carry out investigations. In the case under study, the Public Prosecutors

F i g. 2 : Front façade before the collapse

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The second phase (from March to July 2012) was regarding the so-called special evidence pre-trial hearing.

According to the Italian Criminal Law, for witnesses or evidence that may not be available at trial, either the public prosecutor or the suspects’ defence can request this procedure, which allows for the hearing of testimony from a witness, thus preserving a witness’s testimony for future trial proceedings. The testimony is then included in a file for the trial. In this phase, the investigations were chaired by the judge for the preliminary investigations, who appointed experts (i.e. consultants) to carry out the technical investigations. Also in this phase, the suspects and the other parties involved (public prosecutor included) appointed their own consultants to take part in the investigation.

At the preliminary hearing, the judge evaluated all the evidence collected by both the public prosecutor and the judge for the preliminary investigations and decided to order the trial.

At the first grade trial (from July 2013 to December 2015), all the technical consultants (i.e. public prosecutors, preliminary investigation judges, other parties) were heard as witnesses.

F ig . 3: Axonometric and plan view of the ground floor of buildings B (red) and A (white).

Since the first phase of the investigations, the authors of this paper have been the technical consultants of one of the parties involved (i.e. the architectural designer of the building that had to be built in the place of the demolished one).

First Phase of the Investigations

F ig . 4: Front façade after the collapse

appointed experts (i.e. consultants) to carry out the technical investigations. As these investigations, for many aspects, could not be repeated (i.e. some operations of the investigations imply the modification of things 396

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and places), the suspects and the other parties involved (e.g. the victims’ relatives) could appoint their own consultants to ensure their right of defence.

During the search for survivors in the aftermath of the collapse, the rubble of the collapsed building was reshaped. Moreover, under the supervision of the public prosecutor’s consultants, many parts of building B, considered unsafe, were immediately demolished; all rubble was removed a few days after the collapse, and most of the wall between buildings B and C was strengthened by using shotcrete (Fig. 5b). Thus it was not possible to directly identify dimensions, typologies, location and connections of the structural elements.

At the beginning of the investigations, two hypotheses about the collapse were mainly formed by the technical consultants:

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(a)

(b)

Fi g. 5: (a) View of building A in January 2010; (b) view of buildings B and C few days after the beginning of the first phase of the investigations

1. Errors in the demolition design: the two buildings had a unitary structural behaviour, and hence, the demolition of one building caused the collapse of the other due to the loss of equilibrium (i.e. rotation of the wall between the buildings A and B, caused by the not-equilibrated thrusts of the vaults of building B).

2. Demolition errors: in the days immediately preceding the collapse, the demolition of building A, until then conducted by hand, was conducted by using heavy mechanical machines. Some video shoots proved the inappropriate and violent use of an excavator to demolish the last part of the ground floor of building A. At the end of the first phase of the investigations, the consultants of the public prosecutor made available to the other consultants a laser-scanner survey of the rubble and of the uncollapsed parts of building B, the results of some (only five) compression and unit weight evaluation tests on masonry specimens taken from the rubble and many photographs of the demolition work and of some cracks in building B during the demolition of building A.10 No numerical analyses were performed by the consultants of the public prosecutor’s office.

According to the final report10 of the public prosecutor’s consultants, the causes of the collapse were the errors in the demolition design (i.e. the above-mentioned hypothesis 110).

This conclusion was based only on the assumption that, taking into account that buildings A and B were adjacent, some structural interactions between the two buildings had definitely occurred in time, and hence, the demolition of building A caused the loss of equilibrium of building B.

Second Phase of the Investigations The consultants of the judge for the preliminary investigations chaired the second phase. The site condition at the beginning of their operations was severely changed because of some further strengthening work of the wall between buildings B and C. They could not perform tests as most of building B had been completely removed. They conducted further surveys (mainly in buildings C and D), analysed the report of the consultants of the public prosecutor and collected further documents regarding the compliance of the interventions with urban planning laws.11 No numerical analyses were performed by these consultants. Their judgement was mainly based on the detailed analysis of the acquired documentation (report of the public prosecutor’s consultants included).

According to the final report11 of the consultants of the judge for the preliminary investigations, the causes of the collapse were both design and demolition errors (i.e. the abovementioned hypotheses 1 and 211). In their final report, they did not provide elements in order to differentiate the contribution to the collapse of each of the assumed causes.

The Historical Investigation The authors started their analyses with an in-depth search for historical documents to better understand the construction phases of the building aggregate and hence the relevant possible interactions of the adjacent buildings.

Owing to this long and painstaking search in the State Archives, it was

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possible to understand the construction sequence starting from the edification in 1880 of the first building.12 It is worth noting that because of huge difficulties in finding historical documents, the final report12 was delivered only during the first-grade trial. According to this report, the aggregate was originally composed of six different buildings. Each of them was built in different years and underwent different and independent events of modifications, demolitions, reconstructions and enlargements.

Building A was the last to be built, around 1930 (in substitution of two buildings that collapsed in 1914), while building B reached its final configuration in 1901. According to the authors, this proves that building B had been auto-stable for more than 15 years without the hypothesised support of building A.12

The Numerical Analyses Although the data acquired (regarding geometry, materials, etc.) during the investigations were insufficient to perform reliable numerical analyses, during the first-grade trial, two (i.e. a twodimensional, 2D, and a three-dimensional, 3D) non-linear finite element analyses were carried out by one of the consultants of the judge for the preliminary investigations.13 Currently, as only the first-grade trial is finished, it is not possible to share the images of the 3D model because they are included in the report of a consultant of the judge for the preliminary investigations.13 However, Ref. [14] illustrates a model very similar to that shown in the first-grade trial.

The 3D model regards the first floor of building B before the collapse and was developed using the finite element code Strand7. In this finite element model, masonry and vaults were modelled as an equivalent homogeneous material (i.e. the model considered the two leaves plus the fill of the masonry wall as an equivalent homogeneous material).

In the 3D model, two different analyses were performed. In the first case, the Rankine constitutive law was adopted with a bi-linear stress–strain diagram (both in tension and in compression) with the following limits: • Compression yielding stress: 1.667 MPa • Compression yielding strain: 0.33% • Compression ultimate stress: 1.7 MPa Scientific Paper

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â&#x20AC;˘ â&#x20AC;˘ â&#x20AC;˘ â&#x20AC;˘ â&#x20AC;˘

Compression ultimate strain: 1.00% Tension yielding stress: 0.167 MPa Tension yielding strain: 0.03% Tension ultimate stress: 0.170 MPa Tension ultimate strain: 1.00%

As the ultimate stress values are very similar to the yielding ones, and the ultimate strain values are very high for masonry, the hardening branches were added for computational reasons and/or to appreciate (in a qualitative way) the plastic behaviour of the masonry.

In the second case, a Mohr-Coulomb constitutive law with 30 friction angle was assumed (in Ref. [13], there is no information about the assumed value for the cohesion, while in Ref. [14], it is equal to 29 kPa).

The model also included the transversal walls of building A that were present before the ďŹ nal demolition (i.e. those visible in Fig. 5a). The analyses were performed by gradually removing these transversal walls. The model shows that even if the demolition of the transversal walls causes an increase of the out-of-plane deformation of the wall between buildings A and B, no collapse in building B is reached.

The 2D model considered one vertical section of the ground ďŹ&#x201A;oor of building B and was implemented in the LSDYNA numerical code.13,15 Regarding the plan of the building complex before the last demolition (Fig. 6a), the hypothesis developed is that the critical element is the contrasting

transversal wall (B in Fig. 6a), which contrasts the horizontal thrust of the vault.

The 2D model represented only a typical transversal section of the ground ďŹ&#x201A;oor, and hence, no spatial effect of the vault was considered. Three hypotheses were made regarding the volume of the inďŹ&#x201A;uencing part that acts on the transversal contrasting wall. The length (L in Fig. 6a) of the part (i.e. the â&#x20AC;&#x153;pertinent volumeâ&#x20AC;? in Fig. 6a) was considered a variable parameter. The values assumed for L in the three cases are 10, 7.1 and 5 m, respectively.

In the 2D model, the bricks were modelled individually, but the mortar was modelled as a contact (with nonlinear properties) between the bricks (thus, the mortar was not explicitly modelled). The coefďŹ cients of static and dynamic friction were assumed to be equal to 0.4 and 0.2, respectively. The wall between buildings A and B was considered to be composed of two separate vertical leaves without any connection and ďŹ lled with an incoherent material (Fig. 6b). The brick density was assumed equal to 1650 kg/m3, while the density of the rubble masonry inside the wall and above the vault was assumed to be equal to 1400 kg/m3.

The walls were considered to be ďŹ xed at the base, and the wall between buildings B and C was considered to be horizontally ďŹ xed along its height.

Considering that this is a time-history analysis, and hence, the deformed shape changes in time, Fig. 7 shows the deformed shape of the structure in

(a)

the abovementioned three cases 0.9 s after the removal of the contrasting transversal wall.15 According to the model, the collapse occurs in all the three cases, but its velocity depends on the â&#x20AC;&#x153;pertinent volumeâ&#x20AC;? (as shown in Fig. 7).

According to one of the consultants of the judge for the preliminary investigations, the 2D model, together with the 3D model, proves that the triggering cause of the collapse was the ďŹ nal demolition of the transversal walls. In fact, due to this demolition, the wall between buildings A and B was subjected to out-of-plane actions (i.e. the thrusts of the ground ďŹ&#x201A;oor vaults) that caused it to reach its ultimate transversal strength.

After the abovementioned numerical analyses performed by one of the consultants of the judge for the preliminary investigations, the authors performed a simple equilibrium analysis under the same assumptions of the 2D model analysis.16 In this equilibrium analysis, on the wall between buildings A and B at the ground ďŹ&#x201A;oor, in addition to the vertical loads, the thrust of the ground wall vault was applied (Fig. 8).

In Fig. 8, P1 is the weight of the common wall above the ground ďŹ&#x201A;oor; P2 is the weight of the common wall at the ground ďŹ&#x201A;oor; and P3 is the weight of a half of the vault at the ground ďŹ&#x201A;oor and the load acting on it.

In order to underestimate the safety factor, the load acting on the common

(b) Collapsed part

Demolished part 1200

mm

2500 mm

Contrasting transversal wall 200 400 200

360

Spherical Vault

1200 Pertinent volume

Contrasting transversal walls

800

200 L

2300

3900 Spherical Vault

2800 4600

F ig . 6: (a) Layout of the ground ďŹ&#x201A;oor of building B used for the 2D model; (b) geometry of the 2D model (Unit: mm) 398

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P1 200 400 200

800

829

q = 2 kPa

L = 10 m

D C

360

P3

P2

C

S 200

P3

2300

N

3900

S

L = 7.1 m

E

h=1600

h1=1128

B N

4600

α

P3

A

(a) α = 22.97°

F Fi g . 8: Geometry and forces of the equilibrium model (Unit: mm)

impressed by the ground floor vault on the common wall, consistent with the shape and thickness of the vault, was considered. This value is relevant to the thrust line, indicated as a red line in Fig. 9 and evaluated in the above-mentioned assumptions of the limit analysis.

L=5m

Fi g. 7: Deformed shape of the 2D model 0.9 s after the removal of the transversal wall.

wall and relevant to floors 2–4 was neglected.

In the model, P3 was decomposed into its components N and S, acting on the common wall and on the other half of the vault, respectively (Fig. 8). For the equilibrium analysis, the limit analysis approach by Refs. [17–20] was used.

It is recognised in the scientific literature (e.g. Ref. [21]) that this formulation for the plastic (or limit) analysis of masonry arches and domes synthesises all the historical insights and provides a powerful and theoretically sound tool for the study of this type of constructions. According to this formulation, the limit theorems of plasticity can be applied to masonry structures, provided the following conditions are verified: (a) the compression strength of the material is infinite; (b) sliding between parts is impossible; and (c) the tensile strength of masonry is null.

In order to underestimate the safety factor (i.e. in a pessimistic scenario), the maximum value of the thrust

Assuming MO as the overturning moment of the horizontal component S of N with respect to point A (at the base of the common wall) and MS as the stabilising moment of P1, P2 and P3 (vertical component of N) with respect to point A, the safety factor SF (with respect to the rotation equilibrium of the wall between the buildings A and B) can be expressed as the ratio of MS to MO. In this model, SF was evaluated in the absence of the contrasting transversal wall and in two different situations: 1. The wall between buildings A and B behaves as a unique wall.

2. The two leaves of the wall between buildings A and B behave as two separate walls.

The safety factor is equal to 4.69 and 1.17 in situations (1) and (2), respectively. According to the authors, this proves that, different from what was concluded by the public prosecutor’s consultants, the collapse did not occur because of the loss of equilibrium of the wall between buildings A and B. This is also true also in the worst scenario (i.e. the total disconnection between the two leaves of the wall: scenario (2)). In fact, in this equilibrium model, the thrust of the ground

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floor vault could find equilibrium in the wall between buildings A and B thanks to the vertical loads acting on this wall.16

The Causes of the Collapse: Different Parties, Different Opinions As mentioned in the previous paragraph, the equilibrium analysis performed by the authors has numerically demonstrated that the collapse did not occur because of the loss of equilibrium of the wall between buildings A and B as assumed by the public prosecutor’s consultants.

Although the authors agreed with the consultant of the judge for the preliminary investigations that the 2D model shows the collapse kinematic exactly, the authors have a different opinion regarding the triggering cause.

According to the consultants of the judge for the preliminary investigations: • The rubble of the wall between buildings A and B was completely unable to connect the two leaves.

• There was no evidence of the effectiveness of the header blocks between the two leaves of the masonry walls.

• There was no evidence of the presence of ties that could reduce the thrusts acting on the wall after the removal of the transversal walls of building A. Hence, in their opinion, the 2D model is representative of the actual condition of building B during the demolition of Scientific Paper

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–0.50 3.00

2.50

2.00

1.50

1.00

0.50

0

0 0.50 1.00 1.50 2.00 2.50 3.00

F ig . 9: Thrust line (red) of the ground floor vault used in the equilibrium model (Unit: m)

building A because, in this model, the two leaves of the common wall are completely disconnected, and there are no horizontal ties that can limit the outof-plane action on the common wall caused by the horizontal thrusts of the ground floor vaults. This is why, according to the consultants of the judge for the preliminary investigations, the collapse would have also happened if the demolition of the last part of transversal walls of building A had been made by hand (i.e. the collapse did not depend on the demolition technique but was caused only by the incapacity of the common wall to bear the thrusts of the ground floor vaults).

the collapse did not occur because of the loss of equilibrium of the wall between buildings A and B. 3. The 3D model created by one of the consultants of the judge for the preliminary investigations has numerically demonstrated that by removing building A, no collapse of the building B is reached.

According to the authors, this happens because the 3D model correctly takes account of the connection between the two leaves of the wall and of the spatial effect (i.e. connection between orthogonal walls) that reduces the out-ofplane actions on the walls due to the thrusts of the ground floor vaults.

This means that according to (1) and (2), the collapse did not occur because of the loss of equilibrium but because the structure, for resistance reasons, could no longer sustain all the acting actions. At the same time, because of condition (3), if the two leaves of the common wall were properly connected, as well as orthogonal walls, building B would be able to survive the removal of building A.

Conversely, according to the authors:

It is worth noting that according to the authors, both the effectiveness of the header blocks between the two leaves of the walls and the presence of ties that could reduce the thrusts acting on the wall after the removal of the transversal walls cannot be excluded a priori. In fact, for example, some images taken before the collapse (e.g. Fig. 10) show the presence of many header blocks in the common wall. In these pictures, the distance between the header blocks appears too dense to think that they were completely ineffective in connecting the leaves of the wall. Moreover, during the removal of the rubble, many structural elements (e.g. RC and steel beams) were found that could play the role of ties in order to reduce outof-plane actions on the common wall due to the thrusts of the ground floor vaults.

1. The historical investigation has proved that building B had been auto-stable for more than 15 years without the hypothesised support of building A, and hence, it had the capacity to survive the removal of building A. 2. The equilibrium analysis performed has numerically demonstrated that

Hence, according to the authors, the causes of the collapse have to be found in something that has changed the condition of building B with respect to both the condition of the past (i.e. before the demolition and during the demolition by hand) and that assumed in the 2D model.

According to the investigations, the

Moreover, during the first-grade trial, one of the consultants of the judge for the preliminary investigations declared that even if, on the basis of the historical investigation, it is possible that there was a period in the early 1900s when building B was auto-stable, taking account of the characteristics and the ageing of the materials, it is possible that new structural relations have arisen in time that made building B take advantage of the support “offered” by the building A, especially with reference to the equilibrium conditions of the thrusts of the ground floor vaults.

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F i g. 10: Wall between buildings A and B before the collapse; the header blocks are highlighted with red dots

only cause of this fact can be the inappropriate and violent demolition techniques adopted in the last days. These techniques transferred huge dynamic actions to the wall between buildings A and B that: 1. Broke the connections of the common wall with the transversal ones 2. Broke the connections between the two leaves of the common wall 3. Disjointed the masonry blocks of the common wall In this scenario, the assumptions of the 2D model of the consultant of the judge for the preliminary investigations have become reliable, and hence, this model well describes the final condition and why the collapse occurred. It is worth highlighting that some video shoots of the demolition techniques used in the last days clearly show that the inappropriate and violent use of an excavator tended to crumble the mortar of the common wall and disaggregate the bricks. Hence, it caused the complete loss of wall cohesion.

This means that the causes of the collapse, according to the authors, are demolition errors and particularly the demolition techniques used in the last days (i.e. the above-mentioned hypothesis 2).

Conclusions This paper, starting from the collapse that occurred in Barletta on 3 October 2011, has shown that when dealing with the collapse of a building in historical aggregate, the forensic investigation process seems to be more complicated than in other cases.

While for a single building collapse, surveys, tests and numerical analyses are sufficient to individuate the causes of the collapse, when dealing with a

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building aggregate, the main aspect to investigate becomes the identification of possible interactions between adjacent buildings. This should be the starting point of the investigations, and hence, in the absence of this kind of information, each analysis becomes unreliable because it is based on unreliable assumptions (i.e. the hypothesised interactions).

This means that, in this context, the first step of the investigation (before either the total or partial removal of the rubble) should be a detailed historical investigation. As shown in the case under study, the understanding of the construction/demolition/collapse sequence of the buildings in the aggregate is necessary to better find all possible interactions, and hence, it should address the investigations.

The next step should be the detailed analysis of the rubble in order to understand not only the possible kinematic of the collapse but also the connections between different elements.

In this phase (before the removal of the rubble), very simple numerical analyses, based only on the acquired documentation, could be very useful in order to highlight the elements/ parts of the aggregate that play an important role in the overall stability and strength. These elements/parts are the main points on which it is necessary to focus the investigations, starting from the removal of the rubble (with detailed visual inspections during all the operations) and continuing with the above-mentioned methodology, also proposed by Ref. [8].

Unfortunately, this did not happen in the case of the Barletta collapse. In the first phase of the investigations, the public prosecutor’s consultants were too hasty in removing the rubble, losing evidence of structural interactions between buildings. This aspect was highlighted by the authors in that phase but without gaining any outcome by the public prosecutor’s consultants (who chaired this phase of the investigations). This is the main

reason that the conclusions of the authors and that of the consultant of the judge for the preliminary investigations, even if agreed on the kinematic of the collapse, are different with reference to the triggering cause and, hence, to the people responsible.

References [1] Palmisano F, Vitone A. Changes in codes following structural failures: an Italian perspective on mitigation of structural vulnerability.

Proceedings of the IABSE Conference 2015 – Structural Engineering. Providing Solutions to Global Challenges, Geneva, 23–25 September, 2015; 801–806.

[2] MIT (Ministero delle Infrastrutture e dei Trasporti). D.M. 14.01.2008. Approvazione delle Nuove Norme Tecniche per le Costruzioni [Italian Technical Standards], 2008 (in Italian).

[3] MIT (Ministero delle Infrastrutture e dei Trasporti). Circolare del Ministero delle Infrastrutture e dei Trasporti 2 febbraio 2009, n. 617.

Istruzioni per l’applicazione delle ‘Nuove Norme tecniche per le Costruzioni’ di cui al Decreto Ministeriale 14 gennaio 2008 [Commentary of the Italian Technical Standards], 2009 (in Italian).

[4] Ratay RT, & Peraza DB. Investigation and Analysis of Structural Collapses. In Encyclopedia of Forensic Sciences, Siegel JA, Saukko PJ (eds), vol. 2 Academic Press: Waltham, 2013; 461–465.

[5] Palmisano F, & Vitone A. A meaningful case of a collapse caused by hidden structural defects. Struct. Eng. Int. 2015; 25(3): 308–318.

[6] Palmisano F, & Vitone A. Partial collapse of one of the most important historical buildings in Salerno, Italy. J. Perform. Constr. Facil. 2015; 29(6): 04014164/1–13.

[7] Palmisano F, Vitone A. Lessons learnt from the partial collapse of a historical masonry building. Proceedings of ASCE Forensic Engineering 7th Congress, Performance of the Built Environment, Miami, Florida, 15–18 November, 2015; 163–172.

[8] Palmisano F, Vitone A, Vitone C, & Vitone V. Collapse of the Giotto Avenue building in Foggia. Struct. Eng. Int. 2007; 17(2): 166–171.

[9] Fédération internationale du béton (Fib).

Bulletin 17. Management, Maintenance and Strengthening of Concrete Structures Sprint-Digital-Druck: Stuttgart, 2002.

[10] Salvatori A, Aledda M. Procura della Repubblica presso il Tribunale di Trani. Relazione tecnica dei consulenti del Pubblico

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Ministero. Crollo edificio in via Roma 18 – via Mura di Spirito Santo 62 – Barletta [Final report of the Public Prosecutor’s consultants], 2012 (in Italian).

[11] Bontempi F, Gerundo R. Tribunale di Trani. Relazione [Final report of the consultants of the Judge for the preliminary investigations], 2012 (in Italian).

[12] Palmisano F, Perilli P. Tribunale di Trani.

Relazione dei Consulenti Tecnici di Parte [Report of the technical consultants of one of the parties], 2014 (in Italian).

[13] Bontempi F. Tribunale di Trani. Risposta alla richiesta del 29/10/2014 [Report of one of the consultant of the Judge for the preliminary investigations], 2014 (in Italian).

[14] Arangio S, Molinaro S, Bontempi F. Basic modeling for the forensic investigation of the collapse of a masonry structure. Proceedings of the Conference Built Heritage 2013, Monitoring Conservation Management, Milan, 18–20 November, 2013; 752–758.

[15] Olmati P, Gkoumas K. Simplified Fem modelling for the collapse assessment of a masonry vault. Proceedings of IF CRASC ’15, Rome, 14–16 May, 2015; 8 pp.

[16] Palmisano F, Perilli P. Tribunale di Trani.

Osservazioni dei Consulenti Tecnici di Parte [Remarks of the Technical Consultants of One of the Parties], 2015 (in Italian).

[17] Heyman J. The Masonry Arch Ellis Horwood Limited: Chichester, 1982.

[18] Palmisano F, & Totaro A. Load path method in the interpretation of dome behaviour. In Structures & Architecture (Proceedings of the 1st International Conference on Structures & Architecture, Guimarães, Portugal, 21–23 July, 2010), Cruz PJS (ed) CRC Press/ Balkema: Leiden, 2010; 1826–1833.

[19] Palmisano F. Interpretation of the behaviour of masonry arches and domes by simple models. Structural Studies, Repairs and Maintenance of Heritage Architecture XIII (WIT Transactions on the Built Environment 131.

Proceedings of STREMAH 2013 – 13th International Conference on Structural Studies, Repairs and Maintenance of Heritage Architecture, New Forest, UK, 25–27 June 2013), WITpress: Southampton, 2013; 233–244.

[20] Palmisano F. Assessment of masonry arches and domes by simple models. Int. J. Struct. Eng. 2014; 5(1): 63–75.

[21] Roca P, Cervera M, Gariup G, & Pela L. Structural analysis of masonry historical constructions. classical and advanced approaches.

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Failure Identification: Procedural Causes and Corresponding Responsibilities Eric Brehm, Managing Dir., BREHM Bauconsult GmbH, Bensheim, Germany; Robert Hertle, President, Hertle Ingenieure, Gräfelfing, Germany. Contact: ebrehm@brehm-bauconsult.de DOI: 10.2749/101686617X14881937385160

Abstract When structural members fail, there must be a physical cause, such as missing reinforcement or failing supports. The physical cause itself is often triggered by deficits in the design, in the execution and in the chosen reliability management procedures. Consequently, these deficits determine the level of responsibility for the failure of the involved parties. Human error generally plays a major role in the development of procedural issues; thus, the focus of avoidance in a reliability management system is of special importance. This paper defines the relevant aspects of a reliability management system, introduces practical examples of failure investigations and outlines the related measures to avoid human error that may have contributed to the failure.

Keywords: case study; failure cause; forensic assessment; reliability management system; design supervision.

Introduction The protection of human life and its physical integrity are fundamental human rights and, consequently, a crucial part of a nation’s legal framework.

In this paper, the term “structural failure” is used to describe a catastrophic event that may cause severe injuries and loss of human life as well as damages to surrounding structures and the environment. Failure in fulfilling limit states other than the ultimate limit state (e.g. serviceability, durability) is not within the scope of this paper as the reliability management measures applied in most countries do not cover these issues. In case of infrastructure, the impact of structural failure on society and economy is especially significant. Total safety corresponding to a complete absence of structural failures is impossible. To provide the utmost safety without making structures inefficient and unaffordable, different approaches for building control have been chosen by various countries. In design, sufficient structural integrity is thought to be achieved through the application of partial safety factors that are deemed to define the necessary margin between the design value of the Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: August 5, 2016 Paper accepted: December 27, 2016

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actions and the design value of resistances. Safety factors are derived from prediction models and stochastic assessment, that is, the design problem is formulated under uncertainty to account for variations in so-called (random) basic variables such as load magnitude, material strengths, geometrical deviations, uncertainty in the prediction models, etc. From there, the reliability of a structural component can be determined through the use of advanced algorithms (see Refs. [[1, [2] for further information). Reliability is a characteristic property of a member that can be compared to the target reliability that is given in design codes (e.g. in Europe EN 1990)[3]. From the reliability analyses, deterministic approaches to account for the reliability of structural members are derived, with the concept based on partial safety factors being the most widely used. In other words, partial safety factors account for uncertainty related to exceeding actions and material as well as geometric structural properties falling below the reference levels. This explains why the number of structural failures due to extreme actions in combination with significant deficits in structural strengths is fairly low (see Responsibilities). However, probabilistic models describe uncertainty under a set of preset conditions.

Reasonable deviations from the nominal values applied in design are thought to be included in this uncertainty.

Human errors, however, likely lead to entirely new conditions as these are not

covered by the mentioned probabilistic models. Thus, the academic structural system “without errors” requires a totally different verification than the real-world structural system “with errors”, that is, the partial safety factors commonly applied in design do not account for human errors. In a conventional forensic engineering assessment, the physical cause for a failure is derived and assessed with an emphasis on the correct application of design codes. It is, however, more crucial to be aware of possible fields of human errors and the required mechanisms and methods to avoid them as structural failure happens mostly due to human errors (see next section). These mechanisms and methods have to be known and understood to provide a comprehensive forensic engineering assessment.

Failure Causes Structural failures have been documented since the recording of events.

In ancient times, structural failures were considered to be acts of God as the events were often disastrous and unforeseeable. Nowadays, structural analysis allows for scientific verification of the structural integrity. However, structural failures still occur.

Researching the structural failures in the past shows that the causes can primarily be categorised by: A. Failures due to unforeseeably high actions or insufficient structural strength B. Failures due to human error Cause A refers to structures with appropriate design according to the valid design standards and codes at the time of construction. Failure then occurred due to extremely high loads that exceed the characteristic value of the actions according to the appropriate codes in conjunction with low material strength. Failure due to this cause is unlikely and normally covered by safety concepts (such as partial safety factors), as mentioned in the Introduction section.

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Cause of failure

Total damages

Ignorance, carelessness

37%

Insufficient knowledge

27%

Underestimation of influences

14%

Forgetfulness and mistakes

10%

Unjustified reliance on others

6%

Objectively unknown situations and influences

6%

Table 1: Causes and distribution of structural failures according to Ref. [4] (published in Ref. [10])

Cause B is responsible for failure in almost every case—failure commonly includes some kind of human error.

Errors in the development of design guidelines and rules are excluded from the definition of human error here. Table 1 shows the typical causes for structural failures in a more detailed manner. The table is based on the findings of Ref. [4]. In the corresponding analysis, only 6% of failures were found to be caused by unforeseeable events. The by far larger portion of failures is caused by human error that could have been prevented by proper measures.

Human error can occur at every stage of planning and construction, not only during design. Therefore, it has to be taken into account and avoided during design and execution as well as during service. According to Ref. [5], human error is divided almost evenly between design and construction phases. Additionally, the often referred to “calculation flaw” has proven to be only a minor reason for a serious design error. Failure does not necessarily happen within the first years in the service life of a structure, even though it is more likely. As failure is linked to human error in almost every case, the forensic engineer’s main responsibility is the identification of such errors after a case of failure.

To be able to do this, the forensic engineer needs to be aware that human errors of the types provided in Table 1 do not happen without a reason. The commonly mentioned reasons according to Ref. [6] are: • Time pressure, too low engineering fees • Pressure to minimise the costs of the structure to be built

• Insufficient coordination of the design • Black-box type of design software • Lack of detailing • Large number of new standards and design rules It becomes clear that human error, and consequently structural failure, is caused to a large extent by systematic and cultural issues (pay grids, time pressure), and thus, the identification of structural failure must also happen systematically.

The two main causes for structural failure, as mentioned above, must be tackled efficiently by the structural codes. While cause A—failure due to unforeseeably high actions or insufficient structural strength—can be counteracted by design concepts in structural codes, cause B—human error—requires an organisational set of measures, referred to as quality or reliability management system (QMS or RMS). Possible QMS or RMS may be provided in design codes, such as EN 1990,[3] Annex B. The presented sets of measures mostly include design supervision and site inspections, which have proven to be effective methods to identify and prevent human errors (see Ref. [7]). While the aforementioned cultural issues need to be assessed by lawyers, the fulfilment of due diligence and the correct application of an RMS need to be assessed by the forensic engineer.

Case Studies General The application of an RMS and the corresponding assessments will be explained and detailed by introducing two case studies of real forensic assessments.

Case Study 1: Roof Collapse Part of the roof of a structure that was used as combined storage and a community centre collapsed after heavy snowfall in Alberta, Canada. The snowfall was not untypical and had been experienced before over the previous 30 years of the building’s service life. The one-storey building’s dimensions were approximately 50 m × 30 m at a height of about 8 m.

The outside walls were reinforced concrete blocks. The roof structure consisted of two single-span timber girders resting on the outside walls on

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one end and a centre glulam beam on the other.

One half of the roof collapsed and deformed the rear walls when it fell.

The other half of the roof structure remained in its place and did not appear to have shifted or been damaged (Fig. 1). The glulam beam also appeared unharmed. About 4 weeks after the first collapse, the rest of the roof collapsed suddenly—about 10 min after a site inspection. The following investigation proved that the design of the trusses was flawless.

However, the installation of the roof trusses was not. Some of the trusses were damaged initially and were also installed incorrectly, resulting in larger buckling length of the top chords.

Over time, the damage increased until the massive snow load broke one truss and caused progressive failure to the roof. The collapse of the first part of the roof caused a displacement in the remaining trusses, which then was further followed by progressive failure of the rest of the roof structure.

Thus, the actual failure cause happened at installation. An experienced engineer would likely have noticed the issue at installation. However, the RMS that was valid at the time of construction did not include an independent design and execution inspection. The designing engineer, often an employee of the truss manufacturer, signs off on the design and shop drawings. Another check of the installation is not required. Prevention of the failure would have been possible only if the execution had been checked.

The Alberta building code requires a coordinating professional (architect or engineer) to consolidate documentation from the involved consultants.

However, in the end, the owner has to sign and validate that site inspections and substantial compliance with the code is provided. This is equal to a repressive system (see Repressive System section).

However, larger projects require a second seal for the design check on the drawings. Smaller projects normally do not. Residential projects, if engineered at all, only require one seal. The seal can be provided from another registered engineering professional who is not required to be a specialist in structural engineering— any licensed professional engineer, for example, electrical engineers, can sign off. It can be seen that the Scientific Paper

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requirement for design checking is strongly linked to the type of project, and the corresponding failure consequences. Thus, in part, Alberta applies a preventive system depending on the kind of structure. The overall system is a combination of repressive and preventive systems.

After an extensive investigation, the incorrect installation of the trusses leading to a heavily reduced buckling resistance of the top chords was determined as the cause of failure. In the case at hand, neither the design engineer nor the contractor was required to check the execution. Site inspections would have required a special assignment by the owner. Therefore, the responsibility lies with the owner who would have to prove negligence or false work by the involved design engineer or contractor. To evaluate the responsibilities, the forensic engineer needs deep knowledge of the Alberta system, needs to be able to evaluate the kind of structure and needs to be familiar with the corresponding procedure.

Case Study 2: Falsework Collapse A steel bridge, constructed in the early 1900s, was being replaced by a new reinforced concrete bridge. The collapse happened when the approximately 120 m long reinforced concrete bridge section was being shifted from the location where it was being produced to the final position. The falsework had been used for shifting as well as for the transfer of the loads due to the concreting (Fig. 2).

The forensic assessment showed that the failure happened due to a combination of a number of causes. First, the

bracing of the framework was not designed to act as a true load-bearing member that would require full structural verification. Second, during the shifting process, the structural system of the falsework changed and became kinematic. These two mentioned causes point out that the planning of the shifting as well as the design of the framework and falsework were contradictory. This had not been discovered by the design review. Another cause was the much early cut-off of the cantilevering framework, which had been connected to a braced section.

The failure occurred due to a sophisticated combination of causes. The causes for the failure in this case were triggered by miscommunication between the parties involved.

Regarding the forensic assessment, an obvious cause that could have triggered the entire failure could not be determined. All mentioned causes contributed. This makes the forensic assessment more complex in terms of evaluating possible responsibilities.

The fact that the application of an RMS does not only make design errors less likely but also makes the forensic assessment more difficult and time-consuming, which should be considered right from the start in a forensic assessment. If a working RMS was applied, which should be the first aspect to be checked, the forensic assessment will likely not deliver simple answers.

Responsibilities General The RMS is often part of a legal framework.

Thus, the actual

F ig . 1: Remaining roof structure from case study 1 404

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responsibilities will and have to be determined in a court of law. However, the responsibilities in the application of an RMS can be assessed from the technical point of view as it consists of procedures and methods that should be applied correctly. The duty of the forensic engineer in assessing a case of failure is thus to also check whether these methods and procedures have been applied correctly. From practical experience, the authors would like to mention that in many cases, forensic engineers limit their assessment to the mechanical aspects of a failure. As an example, the overload of a beam may be the physical cause for failure, although the actual reason may be the negligence of a site inspection as stipulated in the RMS. The forensic assessment will then often produce a report that states the overload as a reason, and consequently, it will be up to counsel to determine whether an RMS was applied correctly or not.

If the assessment of the RMS became part of the forensic engineer’s assessment, the complete investigations could be more efficient. A reason for this is that forensic engineers understand RMS as another technical aspect of achieving reliable structures.

RMSs are different internationally.

The respective state administrates this responsibility by granting building permits. The RMS is consequently linked to the building permit. Every state has different legislation so that the RMSs exhibit minor or major differences. However, the RMS can be divided into two general types depending on the main approach to the avoidance of structural failures.

F i g. 2: Debris of formwork from case study 2

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Repressive System A typical version of a repressive system is explained by an example system applied in France. For the issuing of a building permit, the building authorities only check zoning aspects and the building masterplan. Structural integrity is within the responsibility of the parties involved with the construction: architect, contractor and owner who are legally obligated to build up to the relevant codes. However, fulfilment is not to be checked by the authorities. The parties are also legally obligated to take out insurance policies to cover risks associated with the construction.

The advantage of this system is the financial safety in case of failure and damages as the insurance will cover possible claims. However, disadvantages are substantial—reliability management does not happen to prevent failure in the first place, and the structural cost increases significantly due to the cost of the insurance policies.

Design reviewing and site inspections are only performed if an involved party specifically requires these, for example, the insurance company might grant better policy conditions for the owner if design reviewing is performed. Compared with the above identified causes for structural failure, the following must be stated.

Cause A (unforeseeably high actions or insufficient structural strength) and Cause B (human error) are not necessarily covered as there is no legal requirement to review the design before the building permit is granted.

A building permit can be obtained without design and execution supervision. However, design and execution supervision often happens under private law.

Preventive System Contrary to the repressive system, the preventive system tries to avoid structural failure in general by consequent design supervision, before the permit is granted, and on-site inspections for checking the execution. A typical example for a preventive system is the one applied in Germany, presented in Fig. 3. In this system, a full design review has to be performed by a highly qualified and chartered design review engineer and reported to the building authorities before the building permit is granted. During construction, the design review engineer

Owner Grants permit

Protects

Applies for permit

Building Authority

State

Assigns Checks qualification and grants license

Assigns Verifies accuracy of design Controls

Structural Design Engineer

Prepares structural design documents Reviews design and provides advice

Design Review Engineer Inspects

Fi g . 3: Preventive system using the example of the German system

is responsible for checking all the relevant design drawings as well as for performing sample inspections of the execution (e.g. reinforcement in situ).

The disadvantages of this system lie within the possibly larger inertia in the design and execution phase as an additional, economically independent party is involved. However, this additional player introduces advantages into the design and execution process as he provides the prevention of failures and cost-efficiency as additional insurance policies are not required.

Compared with the identified causes for structural failure, the following must be stated.

The first and second causes are covered as there is a legal requirement to check the design before the building permit is granted, for example, in Germany, the detailed report of the design check is a requirement for granting a permit. A building permit cannot be obtained without design and execution supervision. After completion, the design review engineer has to provide another report.

The preventive system is sometimes criticised for the possible relying of the design engineer on the design review engineer, leading to a flawed design in the hope that the design review engineer will find mistakes.

From the author’s experience as a consulting engineer, however, this is generally not the case because the design engineer understands the liability issue (the liability lies with the design engineer, not the design review engineer), and a flawed design will harm the firm’s reputation. Additionally, the design review engineer does not solve possible mistakes for the

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design engineer; he will require additional documentation and verification.

This again leads to more effort. Thus, relying on “design aid” by the design review engineer will only cause an inefficient design and disadvantages for the design engineer.

Influence on Forensic Engineer’s Assessment From the above explanations, it becomes obvious that in a repressive system and a preventive system, assessments of the forensic engineer have to be different. As the building permit already proves that a design check has been conducted in a preventive system, the likelihood that an error in design led to the failure at hand is lower than in the case of the situation in a repressive system. Here, whether a design check was required and conducted has to be verified first. Thus, the forensic engineer’s assessment starts at a different level. It can be stated that: 1. In a preventive RMS, the forensic assessment should start with an assessment of the execution works.

Only if a comprehensive cause cannot be determined should the design be revisited. If a cause in the execution was determined, however, a limited check of the conformity of execution and design for the respective member should be conducted.

2. In a repressive system, the assessment should start with the design and then be extended to the execution. Alternatively, design and execution should be checked in parallel.

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failure causes in design. By having greater confidence in the design due to the performed design check, assessments of failure cases after a long period of time post completion are also more efficient as design reviews are very difficult to perform in these cases due to missing design documentation and drawings.

Aspects of an RMS General From the previous case studies and findings from the Responsibilities section, the following sections will attempt to create a comprehensive, optimised proposal for an RMS. In this section, the aspects that should be covered in every RMS are derived and explained. All these aspects should also be considered when assessing the application of an RMS. Efficiency The effort for reliability management must be adjusted to the task. As an example, it is clear that a nuclear power plant requires more detailed supervision than a town house. Therefore, a general requirement for an RMS is efficiency by means of adjustable requirements. In the forensic engineer’s assessment, it must be evaluated whether the applied RMS was sufficient and adequate for the task at hand.

Failure Consequences and Complexity of the Task To achieve efficient progress in design, the required level of design checking and review must be related to the severity of the failure consequences and the complexity of a structure; that is, structures with a severe impact on society in the case of a structural failure (infrastructure, large number of possible fatalities, etc.) as well as highly complex structures (shells etc.) require higher levels of checking and review than simple structures or structures that will not cause severe failure consequences.

Relevant Levels of Design Checking and Review and Site Inspections Design checking and review and site inspections are very important as they tackle both the previously identified causes for structural failures (see Failure Causes). By assuring that the design and execution have been 406

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performed according to the relevant design codes, the probability of failure due to unforeseeably high actions or insufficient structural strength can be minimised. However, the main task in design review and checking is the discovery of human error. In an RMS, the required level of design supervision and site inspections must be detailed. In general, both can happen at different levels: 1. Self-check Plausibility checks of design are performed by the (design) engineer himself or herself as every engineer does every day, for example, by using a pencil and an eraser.

2. Internal review The design is reviewed by another person within the same design organisation. The possible spectrum of this method can range from discussion with a colleague to a check procedure run by a specialised department. The effectiveness increases with the chosen procedures and the independence— technically as well as economically—of the reviewing body.

3. External review The design and execution is reviewed by an specially assigned organisation that is completely independent from the design organisation. This is a very effective method due to the lack of shared interest between the design and review organisation.

4. External review by an independent, chartered expert The effectiveness of the review increases with the qualification of the reviewer. In some states, reviewers who had to undergo elaborate licensing examinations (chartering) perform the design review by fulfilling the responsibilities of the public administration. These individuals provide different angles, insight and extensive experience and often not only support efficiently in eliminating errors but also help improve the original design. This is considered to be the highest level of design review.

Possible criticism to licensing and chartering is the fact that the sole volume of required design checking exceeds the capacities of single (licensed) design review engineers.

Because of this, the design review engineers will normally require a team of engineers working for him or

her. However, the superior qualification of the design engineer is necessary to point out weaknesses in the design and pose the “right” questions when provided with the results of the design review. It should be within the responsibility of the design review engineer to choose the project team wisely and with the highest level of trust. The composition of the team and the required leadership abilities are another reason why the licensed design review engineer must be exceptionally qualified. Pure design proficiency is not enough.

In addition to the design review, site inspections are required to verify the correct transfer from the abstract design level to the actually built state.

The inspections can be performed by different parties and at different stages comparable to the aforementioned ones.

Self-check and internal review will hardly be documented or available to the forensic engineer. If an external check had not been performed or assigned, it is almost impossible to prove the required design supervision, which will shift the responsibility towards the client.

Qualification of Designer and Design Review Engineer Owing to the varying complexities of structures, the required qualification of the design engineer plays a major role in providing high-quality design.

Qualification of the design engineer is often understood as the background in the form of academic degrees and relevant experience. Additionally, the capacity of the organisational background in which the engineer acts may have an impact on the capability of the engineer. For example, an excellent engineer working alone on a large-scale project may be overwhelmed, while an inadequate engineer will not necessarily succeed because of the support of a large engineering firm. There are many examples of small engineering firms that excel in large-scale projects due to the organisational efficiency, while large companies may be slow in processing due to the organisational overhead and vice versa. The link of engineering qualification solely to the organisational size, processes and capacity is not purposeful and will not provide sufficient answers. Furthermore, qualification must be based and assessed

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on a three-dimensional level: personal, technical and organisational qualification. This is especially important when it comes to the qualification of the design review engineer. The design review must not only be understood as a search for errors and flaws—it also opens up the chance of an independent second opinion with a possibility of improvement and optimisation of the design at hand.

When assessing the correct application of an RMS, the design qualification must also be assessed. Often, the design qualification has to be proven before assignment by providing references.

However, this is not impartial as references can be modified and are hard to check. The forensic engineer should communicate closely with counsel to determine whether an engineering firm was actually qualified to provide the assigned services. This is not only related to engineering competence but also to the available resources.

A Proposal for an Optimised RMS Allowing for Improved Forensic Assessment

required to minimise the cases of failure. It is the forensic engineer’s responsibility to provide the lessons learned from structural failures and introduce them into the technical discussion to improve the built environment for future generations. Thus, a proposal for an optimised RMS will be provided in this section.

Different approaches exist depending on national legislation and culture.

The aspects mentioned in Aspects of an RMS section define an RMS and must be addressed in the forensic engineer’s assessment. The design qualification criteria and the requirements in terms of design and site supervision especially need to be addressed in an RMS and require thorough assessment by the forensic engineer.

Preventive System As mentioned in Repressive System section, a preventive system covers the identified causes for structural failure efficiently and should be preferred. This generally means that the building permit should be linked to design and execution supervision.

General Failure must be prevented. Total failure prevention is impossible due to human error. A working RMS is

Proposed Classes From the elaborations above, the following classes should be introduced

and applied consequently within an efficient RMS to cover the identified required elements from Aspects of an RMS section.

• Consequence classes (CC) The consequences of failure must be used to determine the required effort for design and site supervision. Consequence classes are defined in, for example, EN 1990[3] and presented in Table 2.

• Structural classes (SC) Complexity as a measure for errorproneness must be combined with the relevant consequence class. In Germany, structures are grouped into different structural classes in state law (see [8]; also see Table 3). Also, the engineering fee from the national pay grid (see [9]) depends on the complexity of the structure.

• Design competence classes The possible link between CC and design supervision level (DSL; Table 4) can be found in Table 5.

• Design supervision levels Design supervision is an effective tool to tackle human error. It can be assumed that it may be especially effective when executed by an

Consequence class

Failure consequences

Example of buildings and civil engineering works

CC3

High consequence for loss of human life or economic, social or environmental consequences very great

Grandstands, public buildings where consequences of failure are high (e.g. a concert hall)

CC2

Medium consequence for loss of human life; economic, social or environmental consequences considerable

Residential and office buildings, public buildings where consequences of failure are medium (e.g. an office building)

CC1

Low consequence for loss of human life, and economic, social or environmental consequences small or negligible

Agricultural buildings where people do not normally enter (e.g. storage buildings, greenhouses)

Table 2: Consequence classes Structural classes

Level of difficulty

Characteristics and examples

SC 3

Structures with a high level of difficulty

• Complex statically indeterminate structures • Structures with non-trivial load scenarios and action effects • Highly complex structural systems requiring, for example, non-linear calculations or dynamic effects to be considered • Complex structures requiring new design techniques or design assisted by testing • Pre-stressed and post-tensioned structures • Difficult stability considerations required

SC 2

Structures with medium level of difficulty

Difficult statically determinate or statically indeterminate regular structures built with common construction techniques

SC 1

Structures with a low level of difficulty

Simple statically determinate structures built with common construction techniques

Table 3: Structural classes

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Design Design supervision class supervision level DSL3

External independent design check by a chartered design review engineer

DSL2

External independent design check, strongly recommended to be performed by a chartered design review engineer

DSL1

Self-check

Table 4: Design supervision levels

Consequence class

Design supervision level

CC3

DSL3

CC2

DSL2

CC1

DSL1

The higher DSL determined from Tables 5 and 6 is governing.

Table 5: Link of consequence class and design supervision level

Structural class

Design supervision level

SC3

DSL3

SC2

DSL2

SC1

DSL1

The higher DSL determined from Tables 5 and 6 is governing.

Table 6: Link of structural class and design supervision level

independent, external (i.e. not a member of the designing office to guarantee objectivity) engineer fulfilling the responsibilities of the public administration. In case of CC2 and CC3 (see Table 2), namely, when substantial damages and fatalities are possible, it

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is recommended that the design review be performed in such a way.

For structures that yield only minor failure consequences, a self-check by the design engineer or the corresponding organisation is sufficient. A possible definition of DSL can be found in Table 4.

an international setting. However, it becomes clear that the forensic engineer should start his assessment, once all data are collected, with the assessment of the RMS and application issues. From there, the level of detailing can be estimated, and the focus of the assessment can be set.

The relevant DSL can be determined from the complexity of the task through the use of the SC and the severity of the failure consequences in terms of CC as shown in Tables 5 and 6. Note that the higher DSL determined from either CC or SC is governing.

To prevent failures and avoid the loss of human life associated with structural failures, the necessary aspects of an RMS are evaluated and provided along with a proposal for an optimised RMS.

• Site supervision levels (SSL)

References

The idea behind SSL is the same as for DSL. Human error must be avoided, and this requires inspection.

Conclusions Sophisticated skills in mechanics and physics will not necessarily provide the required answers in terms of the determination of responsibilities after a structural failure. However, these skills form the basis for any assessment. Once the physical cause has been identified, responsibilities can be determined. Note that the term “responsibility” refers to the responsibility of a person to perform in a certain way from a technical point of view. It does not refer to responsibilities in the form of liability—the determination of these is the purpose of a court of law, that is, a judge. The case studies showed that a deep understanding of reliability management is essential for the determination of responsibilities—and this requires the forensic engineer to be fully aware of the legal environment and the relevant RMS, especially in

[1] Rackwitz R. Zuverlässigkeit und Lasten im konstruktiven Ingenieurbau. Technical University of Munich, München, Germany, 2004.

[2] Brehm E. Reliability of Unreinforced Masonry Bracing Walls. Doctoral Thesis, Edition Massivbau, TU Darmstadt, Germany, 2011, ISBN 978-942886-02-4.

[3] CEN/TC 250, EN 1990: Basis of Structural Design, Brussels, 2002.

[4] Matoussek M, Schneider J, Untersuchungen zur Struktur des Sicherheitsproblems, IBKBericht 59, ETH Zürich, 1976.

[5] Ellingwood B. Design and construction error effects on structural reliability. ASCE J. Struct. Eng. 1987; 113(2), 409–422.

[6] Vereinigung der Prüfingenieure für Baustatik in Baden-Württemberg, Bauschäden und Bauüberwachung, VPI BW, 1995.

[7] Schneider J. Sicherheit und Zuverlässigkeit im Bauwesen, vdf Hochschulverlag: Zürich, 1994; 13.

[8] Proske D, Lieberwirth P, & van Gelder P. Sicherheitsbeurteilung historischer Steinbogenbrücken, Dirk Proske Verlag: Wien Dresden, 2005.

[9] HOAI. Honorarordnung für Architekten und Ingenieure, Bundesrepublik Deutschland, 2013.

[10] Proske D. Catalogue of Risks, Springer: Berlin-Heidelberg, 2008. ISBN 978–3642098482.

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Requisite Diversity of Expert Witness Testimony and Issues of Objectivity, Bias and Advocacy Joshua B. Kardon, Principal, Joshua B. Kardon+Company Structural Engineers, Berkeley, California, USA. Contact: jbkse@jbkse.com DOI: 10.2749/101686617X14881932436852

Abstract The United States’ court systems allow each litigant to retain qualified expert witnesses to provide opinion testimony under oath, in support of the litigant who retained the expert.

The Court qualifies and admits the witness, and the expert testimony presented by each side is subject to cross examination. According to some legal systems in other countries, and also in some circumstances in the US, the Court itself may retain a single expert to provide technical evidence independent from any litigant’s perspective. However, because of an individual’s biases and because of inherent threats to objectivity, a single expert may not possess or be able to bring to bear the diversity of knowledge, experience, opinion or judgment to adequately evaluate all reasonable sides of a dispute. The subject of this paper is objectivity and bias of experts, and advocacy by the expert witnesses for their client. The objectives of the paper include improvement of readers’ appreciation of the concepts of objectivity, bias and advocacy, and the need to emphasize the importance of diversity of opinion in the resolution of disputes.

Keywords: forensic engineering; dispute resolution; expert witnesses; bias; objectivity; advocacy.

Introduction Error and uncertainty are inseparable from engineering and construction.

Error arises from the natural variability of the materials used in construction, the assumptions, modeling and analysis methods used in the design process, the limitations on time and resources available to the engineer or builder, and the environmental conditions the built facility will be exposed to during its intended service life.

Error also occurs as accidental deviations from an intended process or practice. Uncertainty originates in the incomplete knowledge or understanding of conditions, materials, processes, forces and loads. Engineering design

methods and practices, including the development and application of building codes, acknowledge the existence of inherent error and uncertainty in several ways, such as prescription of default or minimum load values, the inclusion of load and resistance factors and the probabilistic characterization of extreme environmental loads.

The Role of Expert Testimony in Dispute Resolution Error can result in failure, which may cause injury or damage and may lead to disputes. After other means of dispute resolution are exhausted, litigation may be undertaken to explain the causes and effects of a failure, to define compensation for injury, to describe appropriate repair or mitigation, or to assign responsibility to pay for the consequences of the failures.

Technical aspects of the issues in dispute may be beyond the areas of knowledge of decision makers such as judges or juries. In that event, expert witnesses may present testimony on those technical issues to aid the trier of fact.1 Experts are people, and by their very nature, people have biases.

The premise of this article is that for a trier of fact (a judge or jury) to arrive at a decision, that trier of fact must be presented with a diversity of opinion presenting the technical issues from a wide perspective and from various points of view.

Does Litigation Stifle Productivity?

A pre-publication reviewer of this article noted that some observers may believe that law suits—and the presence of many expert consultants involved for various sides of a dispute—hamper productivity. Do societies where technical disputes concerning engineering and construction are resolved by litigation stifle innovation or productivity more than other societies where focus may be thought to be more on solving technical problems rather than fighting over them? This paper does not address that important question in its

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entirety, but examines the limited aspect of expert witnesses being retained by each litigant independently to present technical evidence on behalf of that litigant as opposed to a single expert retained not by the litigant but by the trier of fact.

Objectivity One can define “objectivity” as the expressing or dealing with facts or conditions as perceived without distortion by personal feelings, prejudices or interpretations. To be objective is to be uninfluenced by feelings and to make judgments based on observable phenomena, uninfluenced by emotions or personal prejudices.

The expert can maintain objectivity by analysing the data, arriving at opinions and presenting opinion testimony dispassionately, and by assuring that the expert’s perceptions are not clouded by prejudice or unduly influenced by unacknowledged bias. A possible source of a loss of objectivity on the part of an expert witness lies in the difference between the roles of the attorney and of the expert the attorney retains.

About this difference, psychiatric expert witnesses2 wrote: The goals and the ethical mandates of the psychiatric expert witness and of the retaining attorney differ in essential ways: these differences always create an enduring tension between the parties. The attorney is ethically obligated to embark on zealous, vigorous and partisan advocacy on behalf of the client.

While objectivity may govern case selection, the attorney need not be objective in presenting the case in court. In contrast, the expert is committed to honesty and to striving for objectivity throughout, even when those goals are accomplished at the cost of disappointing the retaining attorney by, in essence failing to be sufficiently partisan.

Engineering expert witnesses face the same situation, and also have to keep Technical Report

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in mind their professional obligation to serve the client.

The historian cited in Ref. [3] wrote about his view of objectivity of historian expert witnesses,3 and points out that the expert’s objectivity can be influenced by the expert’s own values, tastes, talents and circumstances. The expert’s exposure to and appreciation of currently popular trends or ideas in the profession can influence objectivity. Engineering experts may have personal preferences for particular analysis methods or evaluation criteria; building code sections may be open to various interpretations or assessments of applicability; the types of projects with which the engineer has been involved over the course of the engineer’s practice, and the outcomes of those projects, can differ among experts. All these factors and more can result in differences of engineering judgment or opinion. Those differences might be interpreted as indications of a loss of objectivity if one assumes that there is only one possible interpretation of evidence developed by various experts’ investigations. However, the reason a dispute goes to trial is that there are differences of opinions and shadows of doubt concerning the facts and circumstances surrounding the incident or condition that led to the dispute.

The experts for each party can retain objectivity while serving the retaining party by maintaining awareness of differing views and by understanding and appreciating the positions of the experts helping other parties to the dispute.

Another threat to objectivity is the fact that the expert is providing and being paid a fee for professional services and may want to continue doing so for the attorney in future cases.

An expert’s opinion contrary to the position of the retaining attorney might be misinterpreted as being against the client’s best interest. The expert should realize, however, that the value the expert provides is in carrying out a competent and thorough investigation. If that results in an opinion that the attorney’s position is weak or incorrect, it may be in the client’s best interest for the expert to inform the attorney of that opinion so the attorney can take a position for the attorney’s client based on the evidence presented, or pursue settlement in the most favorable terms for the client.

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The attorney who retains the expert can also influence the objectivity of the expert by framing the question at the root of the dispute. Controlling the framing of the expert’s area of concern, the attorney might be selective regarding information provided to the expert. The opposing attorney could bring up any “framing” of the issue for the expert in an effort to undercut the expert’s testimony if that framing is inappropriate or deceptive.

These and other influences can emerge as a loss of objectivity, or a bias on the part of the expert, but, as Ref. [3] points out, although that bias may be a “predisposition” or “proclivity,” it need not be an “unalterable presupposition.”

Bias Bias is a tendency of people to utilize cognitive shortcuts.4 It is the subject of many areas of intellectual pursuit, including psychology, statistics and economics. There are several types or sources of bias which can affect the testimony of an expert witness, including but not limited to the following: Confirmation bias: The tendency to search for, interpret, focus on, and remember information in a way that confirms one’s preconceptions.5 Curse of knowledge: When betterinformed people find it extremely difficult to think about problems from the perspective of lesser-informed people.6 Hindsight bias: The inclination to see past events as being more predictable than they actually were; also called the “I-knew-it-all-along” effect.

Outcome bias: The tendency to judge a decision by its eventual outcome instead of based on the quality of the decision at the time it was made.

Reactive devaluation: Devaluing proposals only because they purportedly originated with an adversary.

Selective perception: The tendency for expectations to affect perception.

Semmelweis reflex: The tendency to reject new evidence that contradicts a paradigm.7 Random error results from sampling variability and decreases as sample size increases. Bias, however, is independent of both sample size and

statistical significance. The source of many biases may be inherent in the normal human functioning of the mind and consciousness.4 Because biases may be normal and expected tendencies of human consciousness, the forensic expert should be aware of the possible influence of bias on any observation or opinion (the expert’s own observation or opinion, or another’s observation or opinion). The expert should be able to detect biases in his own or in others’ observations and opinions, acknowledge their presence, accommodate, overcome or compensate for them.

Advocacy Advocacy can be defined as the act or process of supporting a cause or proposal. The purpose of expert witness testimony in a trial setting is to “aid the trier of fact”1 in answering the ultimate factual questions in dispute, where knowledge necessary to effectively answer those questions is beyond that of normally competent jurors or judges. Because each party of a dispute may present expert testimony in support of its own position, it should not be unexpected that experts for different parties can arrive at differing opinions, and testify in support of different positions.8 To be helpful and admissible, the expert testimony should be valid and reliable. If the expert witness testifies without a valid and reliable basis for the opinion testimony, the court may exclude that testimony, or the opposing side may discredit the testimony on crossexamination.

If thorough, objective investigation by the expert results in the expert’s opinion that the party who retained him is on the “wrong” side of the dispute, or results in the expert developing a credible argument, the opposing side might also arrive at without there being a valid counter-argument, it is the expert’s responsibility to inform the retaining attorney. The weakness of the client’s case may result in losses which the expert may help mitigate, by, for example, suggesting the attorney discuss with the client a consideration of settling the case prior to trial.

This is a kind of “advocacy” that is within the purview of an expert. The expert provides that advice in support of the client, for the purpose of reducing the risk of a large decision against the client.

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The Importance of Diversity of Opinion For a judge or jury to make a decision regarding the facts in dispute, where those facts are beyond the normal scope of knowledge or experience of normally competent judges or juries, the trier of fact hears technical evidence from experts acknowledged by the court to have that knowledge or experience. Because of the normal, expected influence of bias, and because of threats to objectivity, a single expert’s testimony may not expose the trier of fact to as broad a range of possible opinions regarding the facts in dispute. A diversity of opinion is valuable in the process of decision making.

Models of Decision Making Judges and juries serving as triers of fact are assumed to make their decisions logically, rationally and unemotionally. Current science of cognition, behavioral psychology and economics, however, shows that the rational choice theory is inadequate to explain the entire process of decision making.

It is necessary to understand how a decision-maker sees and appreciates “competing priorities and available choices.”9 Judicial decisions of law are based on doctrine, statute and precedent, but also rely on discretion and other nondoctrinal factors. When precedent cases differ slightly, a judge is more likely to rely on “a very slight preponderance of feeling, rather than of articulate reason....” (Holmes, 1881 cited in Ref. [9]).

Just as critical discretion and other nondoctrinal factors affect decision making regarding questions of law, questions of fact as well are answered by the court with input other than purely rational deliberations.

Decision-making rules based on rational choice generally rely on both expected utility theory and Bayes’ theorem.9 Decisions can be made based on rational choice when the choices are highly restrained (Clark, 1997 cited in Ref. [9]). Such restraint does not exist when several sides of a factual argument are presented.

When multiple perspectives and viewpoints are allowed to be considered, the restraints supporting one particular choice are absent. In that situation, decision making must include an appreciation and evaluation of not only the position argued

or facts offered in testimony, but also on the “scaffolding” or restraints implied or assumed by each side’s argument. “In order to assess the impact of some piece of evidence, an individual must know the theoretical alternatives and how likely each is on prior evidence.” (Richardson, 1998 cited in Ref. [9]) Effective decision making in complicated or complex contexts requires perception and understanding of a broad range of possible answers or solutions. Complicated contexts are those that lend themselves to multiple correct answers, and where the relationship between cause and effect is obvious. Complex contexts are those where correct answers may not be fixed or discernible, where the conditions are dynamic and unpredictable. The circumstances and conditions that lead to a failure resulting in litigation are often both complicated and complex. In those contexts, an effective decision maker must hear conflicting expert advice, encourage dissent and diversity, and allow for interaction of competing perspectives.10

Conclusions When a forensic engineer expert witness testifies regarding technical aspects of a failure, the witness should be attentive to maintaining objectivity and be aware of the presence and influence of bias. Threats to objectivity inherent in the expert–attorney relationship, and threats arising from the expert’s own particular preferences and experience, can be controlled and accommodated.

Confirmation bias may slant the expert’s opinion testimony toward a position developed prior to subsequent evidence which objectively could support a different position.

Hindsight bias may blind the expert to the situations and circumstances that may reasonably have influenced the defendant at the time of the alleged negligence. Reactive devaluation and selective perception may act to prevent an expert from reasonably appreciating the other side’s argument simply because they originate from the opponent and differ from the expert’s opinion. Biases can be overcome or be accommodated by thorough investigation and vetting of opinions and conclusions.

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If a court retains a single expert, unaffiliated with any particular side of a dispute, bias is not eliminated nor is objectivity guaranteed. On the other hand, if each side retains its own experts, and each expert is allowed to present truthful technical evidence from the perspective of the side that retained the expert, opposing experts and the trier of fact can assess the presence of bias and objectivity and their influence on the experts’ testimony. This supports the trier of fact in its proper function, which is to answer the ultimate question of fact concerning the dispute.

The role of the expert is not to answer the ultimate question—that is the job of the trier of fact. A single, court-appointed expert may not be aware of inherent, natural biases, or may not possess the broad objectivity needed to see all sides of a dispute.

A trier of fact making a decision in the absence of diversity of opinion that would be presented by experts for each side of a dispute, may not be able make an informed decision or answer the ultimate question of fact in a dispute.

Reference [11] points out that several authors conclude “[an] aggregate of several individual judgments is more accurate than judgment of [an] individual.” It is the range of opinions, provided by competent, qualified experts for each side, that give the trier of fact the requisite diversity of opinion needed in decision making.

Further Information Holmes O. The Common Law Little Brown: Boston, 1881.

Clark A. Economic Reason: The interplay of individual learning and external structure. In Frontiers of the New Institutional Economics, Drobak J, Nye J (eds) Academic Press: London, 1997.

Richardson R. Heuristics and satisficing. In A Companion to Cognitive Science, Bechtel W, Graham G (eds) Wiley: Hoboken, 1998; 566–67.

References [1] U.S. Government Printing Office. Federal rules of evidence. Washington, DC; 2001.

[2] Gutheil T, & Simon R. Attorneys’ pressures on the expert witness: Early warning signs of endangered honesty, objectivity, and fair compensation. J. Am. Acad. Psychiatr. 1999; 27(4): 546–553.

[3] Kousser J. Are expert witnesses whores?

Reflections on objectivity in scholarship and expert witnessing. Publ. Hist. 1984; 6(1): 5–19.

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[4] Kahneman D. Thinking, Fast and Slow Farrar, Straus and Giroux: New York, 2011.

[5] Oswald M, & Grosjean S. Confirmation bias.

In Cognitive Illusions: A Handbook on Fallacies and Biases in Thinking, Judgement and Memory, Pohl F (ed) Psychology Press: Hove, 2004; 79–96.

[6] Ackerman M (Ed.). Sharing Expertise Beyond Knowledge Management MIT Press: Cambridge, 2003.

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[7] Edwards W. Conservatism in human information processing. In Formal Representation of Human Judgment, Kleinmuntz B (ed) Wiley: New York, 1968; 17–52.

[9] Drobak J, & North D. Understanding judicial decision-making: the importance of constraints on non-rational deliberations. Wash.

Univ. J. Law Po. 2008; 26: 131–152.

[8] Kardon J, Schroeder R, Ferrari A. Ethical dilemmas of technical forensic practice, 3rd Forensic Congress, Technical Council on Forensic Engineering, American Society of Civil Engineers, San Diego, California, 2003.

[10] Snowden D, Boone M. A leader’s framework for decision making. Harvard Business Review, 2007.

[11] Woudenberg F. An evaluation of Delphi.

Technol. Forecast. Soc. 1991; 40: 131–150.

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Problems Associated With Curtain Walls c Tovarovi c, Asst. Prof.; Nenad Šekularac, Assoc. Prof.; Jelena Ivanovi c-Šekularac, Assoc. Prof., Department of Jasna Ciki Architectural Technologies, Faculty of Architecture, University of Belgrade, Belgrade, Serbia. Contact: cikic.tovarovic@gmail.com DOI: 10.2749/101686617X14881937385322

during the construction and implementation phase. The paper identifies potential causes that restrict the use of this facility. The main research method is expert analysis, carried out through on-site observations and detection of explicit evidence, as well as through direct access to the project’s technical and photo-documentation from the contractors. The results of the paper describe the construction measures to be applied to some parts of the glass façade of the house, in order to prevent further deterioration and reduce the negative impact on the users.

Keywords: glass; façades; errors; water leakage; thermal expansion.

Abstract The curtain wall design involves a complexity of elements responsible for performing different functions, which is the key prerequisite for holistic approach to construction and implementation of systems. This paper deals with specific problems associated with curtain walls, as presented in the case study of a new, single-family home of a villa design in Belgrade. The aim of this paper is to present effective ways to overcome the problems of water leakage and negative thermal impacts of solar radiation on curtain wall, which is caused by common mistakes made

E F

F'

G

H

2

atrium

Introduction Effective enclosure depends on the following factors: static stability, waterproofing, air permeability, thermal and sound insulation, solar protection by shading or coating,1 energy efficiency, aesthetics, durability and maintainability.2 Despite widespread use of curtain walls and significant technological development in the industry of glass, profiles and seals, there are still unsolved problems with glass façades and constructions. Ref. [3] provides an almost complete list of potential problems. Some of them are: nickelsulphide in annealed glass, negative thermal effects, corrosion and chemical effects, incompatibility between glass and other materials, water leakage, structural failures, building movements, lack of additional safety measures and maintenance. Some authors provide the guidelines for design and construction of curtain walls, summarizing the types of failures faced in curtain wall systems, with the focus on water-induced damage as the consequence of over-reliance and thermal stresses. This paper lists the most effective prevention techniques for curtain wall failures.4 So far, there has not been any extensive research on the phenomenon of sudden sounds in curtain wall systems.

D

The research deals with the case study of the new building, its glass façade

1%

1%

DV2 2⨯U20

III

B

C

IV

DV5

DV1

A

II

2 0

2

3

4

5

Fi g. 1: Ground floor plan and Section 2-2. (Units: [mm])

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and two specific issues related to the curtain wall system: 1. Water leakage and deformation of the lower horizontal façade element occurring while opening the sliding door, in the curtain wall toward the atrium area 2. Disturbing, unpleasant sounds of unknown origin in the curtain wall area typically occurring in the morning, on sunny days, and acoustically resemble an abrupt “metal–metal” bang

2012. Detailed analysis of the physically observable current situation and interviews with the users of the building provided useful information regarding the building itself and related issues.

The technical documentation of a project consists of the design (floor plans, sections, elevations) and details of the façade. Photo-documentation during the construction of the building and installation of its façade was done by the contractor, from September to November 2011.

The paper presents: - Observations of the problem situations by on-site visit and from the desk.

- Analysis of potential causes of water leakage into the interior of the finished building through the glass façade system and disturbance in acoustic comfort of the occupants of the building.

- Necessary construction measures in order to rectify the detected failures and the identified design errors made during the curtain wall implementation.

This study consists of two hypotheses.

The first hypothesis states that possible causes of the first problem are that moisture damages, at the bottom of the curtain wall, is caused by either inadequate hydro-insulation or condensation. The second hypothesis is that the origin of unpleasant sounds is associated with movements of glass façades and lack of spatial tolerance between different elements of construction and/or façade and primary structure.

Methodology The following methods and approaches were used in the investigations process: • Collection and analysis of data on previous research in relation to glass façade forensic engineering and design problems of glass structures • On-site observation of the building • Conducting of interviews with the occupants of the building • Desk study of project’s technical and photo-documentation created during the project realization • Recommendations to upgrade the current condition of the building concerning the necessary construction measures The house was completed and occupied in April 2012. On-site observation was carried out three months after the completion of the building, on 5 July 414

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Observations General On-site Observation of the Project 1. It is a free-standing family villa on Mladena Stojanovica street in Belgrade, a semi-enclosed atrium-type structure that consists of a basement, ground floor and first floor; the roof is flat.

2. The building has a standard type of curtain walls with supporting vertical aluminium mullions, with low eglazing.

3. The glass façade is predominantly exposed to the sunlit east and south-east.

4. There is no exterior shade on the glass façade; the building has deep eaves that may cause uneven heating of glass façade.

5. There is no visible damage to the glazed unit and there are neither cracks nor any other defects.

6. There is no perceptible damage to the aluminium mullions, inside or outside; there are no relocated aluminium caps.

Observations of Project Documentation about the Design and Construction Choices 1. The glass façade was well designed as a standard curtain wall (specifications: FNJ 60 + HI, ANJS 75HI, ASS 70HI) with fixed transparent areas, maximum size of 2850 × 3179 mm2; triple-glazed (6 + 16 + 6 + 16 + 6 mm); lift-andslide door system, maximum size of 1767 × 3179 mm2. This type of door system is suitable for large glass areas. All the elements and the curtain wall system are well designed and properly used.

2. Non-transparent attic segments are up to 2850 × 800 mm2 and filled with glass on the outside. The size of the vertical supporting bar is 60/200 mm.

3. The analysis of the main and detailed design (floor plan, sections and elevations) does not reveal any defects and design errors that could negatively affect the curtain wall system (see Fig. 1).

4. The glass façade was installed towards the atrium, spreading over two floors. Its total height is 7802 mm. The fence on the roof is made of glass.

5. The curtain wall is placed 30 mm from the primary concrete structure (DV2), the height of the façade is 7940 mm, which means that vertical movements of 22 mm are to be considered (see Fig. 2).

6. Movable anchor system of the curtain wall is well designed.

7. As shown in the photo documentation, the completion of the curtain wall was properly performed, in accordance with DV4, which means that the thermal insulation of the vertical profiles on the upper side was properly carried out, thus preventing thermal bridge and heating of aluminium profiles of the façade at the point of separation between the profiles and the transition zone from “warm to cold” façade.

8. As for the visual arrangement and design elements, the glass fence on the flat roof is a unified whole blending into the façade, but it is structurally separated in accordance with the discontinuity concept of “warm and cold” façade.

On-site Observation Problems 1. There is a perceptible bending of a lower horizontal profile of the sliding door, caused by relatively small loads.

2. There are no dilatation gaskets (Ger. stoβdichtung) between the vertical supporting and horizontal bars, up to 2850 mm long.

3. There is no perceptible exterior dilatation between the aluminium caps.

4. There is perceptible moisture on the inside, at the bottom of the sliding door, on the ground floor between the dining rooms and the atrium, and on the first-floor bedroom towards the balcony.

Archival Photo-documentation and Observation of Problems During Installation The lack of continuous rigid supports at the bottom of the sliding positions/ project implementation details (DV5 and DV11) may cause bending of the

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Analysis Since it is not possible to perform a critical examination of the entire façade system, it is also not possible to determine all potential causes of the unwanted events without disassembling the existing façade elements.

(a)

As for the analysis of the first problem, water leakage, there are two potential problem causes: water infiltration due to inadequate hydro-insulation and condensation (a)

(b)

60

30

Curtain wall 125

180

Since there is no perceptible rigid support for the lift-and-slide door (the built structure is not in accordance with the design details), it is evident that water leakage is the direct consequence of low profile bending, due to lack of inappropriate support at the bottom of the glass door and displacement of the applied insulation, which occurs when the profile without support bends while being exposed to external pressure.

No other problem regarding condensation was detected, although some studies did deal with that very problem in their works.4 Air seal was placed on the outside, in accordance with the previous experience and recommendations,5 which reflects the

296 200 155

30

120

1. Angular metal sheet for closing the gap between the curtain wall and the primary structure is not included in the design.

2. The detail DV1 detects a thermal discontinuity, below the thermal insulating glass (Fig. 4).

3. The detail DV11 detects an incorrectly installed hydro-insulation, on the firstfloor balcony, which may cause water leakage inside the house (Fig. 4).

due to thermal discontinuity (as explained in previous section). Based on water being localized over the entire depth of the façade and due to substantial quantity of water leakage, it can be concluded that moisture damage is caused by water infiltration from the exterior.

600 125 120

lower horizontal bar, i.e. the frame of the sliding door (Figs. 3 and 4).

135 30

Distance from primary concrete structure

(b)

Metal sheets

Primary structure

outside terrace Horizontal mullion

Horizontal mullion

F i g. 2: Intermediate floor details: drawing (a) and photo-documentation (b).

(Units: [mm])

Inside Bending

well-designed elements of this part of the façade, without causing condensation problems in this area.

Rigid insulation Rigid insulation

Restoration of the existing façade elements is recommended: disassembly of the balcony floor and installation of the steel profile with supports.

Hydro-insulation problems

300 420 200

DV5

60

590

120 170

57 390

Thermal bridge

1%

180 120

60

0.00

60

300

250

DV1

155

180 120 60

120

140.9

0.00

160 49 98

150

30 0.00

147

324

47 136.3

Support the bottom

98

Fi g. 3: Rigid insulation is under lowest horizontal mullion

347

DV11

Fi g. 4: Details of DV1, DV5, DV11. (Units: [mm])

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Underpinned Horizontal profile of glass facade 130 mm

4

4 120

10

270

50

100

4

A

100

100 B

Holes 12.5 mm filled with composite mortar

Anchor screw M10

F ig . 5: Reliance of existing curtain wall on the concrete construction–proposal for improvement. (Units: [mm])

Underpinning is carried out by adding elements made of durable material (stainless steel), since this is a conThe distance cealed location.4 between these elements is ≈ 800 mm, in accordance with the submitted design illustration (Fig. 5). Also, the elements of aluminium structure should be separated from new elements in order to prevent the consequences caused by incompatible materials. It is necessary to pay special attention to the thermal insulation design (to avoid thermal bridge and condensation) and to install continuous vapour-tight foil on the inside and hydro-insulation on the outside of the balcony.

The second problem—the occurrence of sudden sounds indicates the movements of certain elements and lack of correctly designed dilatation and separation of the elements. In light of the evidence, it appears that the sounds come from inside the building, considering the high acoustic performance of the installed triple-glazed glass (35 dB).

There are many potential causes of the current problem. First, the sudden sounds could be the result of any system error that occurred during 416

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installation, i.e. rigid, immovable joints were created between the curtain wall and the primary structure during the application of angular metal sheet in order to cover the gap between the curtain wall and the primary intermediate and/or roof ceiling system. Since both the curtain wall and the building itself are potentially susceptible to seismic movements, this might produce sudden sounds at the connection point between the screws in metal sheet and the primary structure. This statement can be justified by numerous studies dealing with the effects of seismic activities on curtain walls.5–7 Second, it could result from the contact between the aluminium façade mullions and the plaster board profile in the intermediate floor area; the profiles for plasterboards are attached to the primary concrete structure, and also to the curtain wall.

Third, due to thermal expansion of the profiles in the early hours of summer mornings, the cooled façade might be directly exposed to solar radiation, which could cause sudden noises, caused by the impact of horizontal bar and movements at the contact place between the horizontal mullion/vertical mullion/transom and the anchor. According to Ref. [8] the problems arise when there is no possibility to accommodate free movements that are caused by thermal extension, due to absence of glass façade. In this case, it is an obvious problem, considering the length of the horizontal transoms, i.e. 2850 mm.

Since the sudden noises occur in various places, depending on the exposure of the façade to solar radiation, it is evident that the sounds are caused by thermal effects and failure to provide necessary dilatation during the project realization. It is certainly not due to the system errors of the applied glass façade. The following measures are suggested in order to solve the current problem: • gradual disassembly of the existing glass panels and cover, pressure and horizontal profiles for curtain walls; • disassembly of the angular metal sheet in the area of the intermediate floor ceiling; • disassembly of the non-transparent segments of the curtain wall in the area of the intermediate floor ceiling;

• installation of dilatation gaskets between the horizontal and vertical mullions of the curtain wall; • gradual assembly of intermediate floor segments of the curtain wall, including thermal insulation of sheet metal cassettes, additional sealing of the space between the curtain wall and the primary structure by applying insulation and installation of a vapour barrier along the edges to prevent the penetration of moisture; • reassembly of the angular metal sheet; • reassembly of the curtain wall.

All the obtained evidence confirmed that it is possible to undertake specific, locally managed construction measures in order to prevent water leakage in the curtain wall area, such as underpinning the lower horizontal elements of the glass façade, thus preventing potential bending and deformation that could affect the insulation layers of the façade at the lowest segments. However, if the façade problem is caused by either lack of significant elements or by system failures during the assembly, the solution requires disassembly and reassembly of the entire façade cladding, which means that the building is not to be used during this period of construction rework.

Conclusions Based on the conducted analysis of the case study of the new family house and its glass façade, two problems were noticed: water leakage and disturbing sounds. Desk study of the project’s technical and photodocumentation and on-site observation, led to the conclusion that the first problem—water leakage in the lowest parts of the curtain wall system, is caused by inadequate support of the glass façade elements and physical deformations of the façade, along with inadequate waterproof and steam-tight barriers. The problem can be solved by locally undertaken construction measures and underpinning. Also, significant mistakes regarding the project realization were detected—the absence of expansion gaskets between the vertical aluminium mullions and horizontal transoms crossbars on the façade.

Inability to manage temperature expansions causes disturbing sound effects, due to the mullion-transom impact, during early hours in the

Structural Engineering International Nr. 3/2017

morning when the façade is exposed to solar radiation. This problem can be corrected only by disassembly and reassembly of all the horizontal aluminium transoms.

The study points out potential problems regarding water infiltration into the peripheral area of the curtain wall as well as issues related to the lack of thermal expansion in the curtain wall systems, which could be relevant to other structures and preventive as well, considering the frequency of appearance.

[1] Kaymiercyak K, & Neeb D. Transitions: how to design facade interface. J. Build. Enclosure Des. Summer 2007: 20–23.

[5] Quirouette RL. Building envelope design using metal and glass curtain wall systems.

Building Practice Note no. 37, Institute for Research in Construction, National Research Council Canada, 1982.

[2] McCowan BD, Kivela JB. Lessons Learned from curtain wall failure investigations. RCI Symposium on Building Envelope Technology, San Antonio, 8–9 November, 2010.

[6] Kazmierczak K. Movements and toleranceconsiderations for curtain wall and gladding design. The Construction Specifier, October 2008; 46–58.

[3] Loughan P. Falling Glass: Problems and Solutions in Contemporary Architecture Birkhauser: Basel, 2003.

[7] Chan SL. Basic structural design considerations and properties of glass and aluminum structures. Proceedings from Symposium Advanced Facade Engineering and Technology, Hong Kong, 2006.

References

[4] Schwarz T. Curtain wall fundamentals.

APT Bull. 2001; 32(1): 37–45.

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Investigation of the Cause of Failure of the Omo River Bridge Girma Zerayohannes, Dr. -Ing Civil Eng.; Esayas Gebreyouhannes, Dr. Civil Eng.; Adil Zekaria-Abdullahi, Dr. -Ing, School of Civil and Environmental Engineering, Addis Ababa University, Addis Ababa, Ethiopia. Contact: adilzekaria@yahoo.com DOI: 10.2749/101686617X14881937384729

Abstract The reason for the failure of the Omo River Bridge that collapsed and partially sank into the river during construction was investigated using standard, structural and analytical finite element model (FEM) and design review procedures according to Eurocode 3. The bridge was initially constructed by a launching construction technique. The span design was based on a CallenderHamilton bridge type with Warren trusses of the D8M type classification.

In all 40 stages were required to put the bridge in place. The bridge collapsed and partially sank into the river during construction, between stages 34 and 35.

From in situ inspection, it was observed that the main bridge did not show any significant distress. Moreover, the mode of failure suggested a weak link between the main bridge and the launch nose.

The results of the investigation showed that the diagonal link angles in the critical region where the main truss was connected with the launching nose were found to be the most critically loaded members. These members were stressed to their ultimate limit state, causing them to buckle and trigger the sudden collapse of the bridge at the end of launching stage 34.

Keywords: girder truss bridge; launching construction technique; battened built-up members; launch nose; buckling failure; failure analysis.

constructed from standard-type steel truss bridge parts through a process similar to actions involved in stages 1 to 11 prior to the launching of the main bridge. To connect the main bridge and the launch nose during the launching operation, special erection link angles were provided.

The Ethiopian Road Authority had commissioned the Addis Ababa Institute of Technology (AAiT) and flown senior faculty members to the site on 15 May 2011 to investigate the cause of failure of the bridge. In situ inspection during the site visit indicated that the main bridge did not show any significant distress. Furthermore, the mode of failure, which resembles a mechanism of a simply supported beam with a plastic hinge at the critical section, suggested a weak link between the main bridge and the launching nose.

The focus of the investigation was therefore directed towards checking the resistances of the members in the critical region.

The investigation was based on the material strengths determined from laboratory tests, with average yield strength of 517 MPa and action effects for load case unfactored self-weight plus additional permanent loads used as counterweight.

Description of Launching Sequences A typical launching cycle is described in the erection manual1 using launching stage 12 as an example. The same is repeated below to highlight the steps undertaken during each cycle and their significance with respect to the required structural models (more than one) for any given cycle. A typical launching cycle consists of: • Adding a D8M panel (D8M is the designation used to identify one particular type of bridge with a depth of 9414 mm and 8000 mm center to center distance of the joints where the main diagonal elements meet the top and bottom chords) • Jacking up the nose and moving the launch skate to the next node • Pulling the bridge forward by half a panel length • Jacking up the bridge and nose and moving the skates to the next node points • Pulling the bridge forward by an additional half a panel length so that at the end of each cycle, the bridge is moved by a distance equivalent to span of one D8M panel as illustrated in Fig. 2 The construction was proceeding successfully until launching of stage 34.

Introduction The Omo River Bridge was intended for construction in the Southern Nations, Nationalities and People’s Region of Ethiopia, approximately 800 km southwest of the capital Addis Ababa. In 2011, the bridge collapsed into the river while still under construction (Fig. 1). The launching method was used for the construction and the main bridge was constructed behind a light-weight “launch nose” and pushed or pulled into place, the bridge being supported on rollers or sliding skates. The launching nose was 418

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Main bridge Launch nose

F i g. 1 : Picture of the collapsed Omo River Bridge near Omarate

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D8M Bridge

Erection Link Steel

B15M Launch

(1)

(2)

Investigation of the Cause of Failure A three-dimensional (3-D) finite element centreline model was developed for the investigation of the cause of failure of the bridge during construction.

All members of the main bridge and the launch nose are uniform battened builtup members with different crosssections, as shown in Table 1. The link angles that are used to connect the main bridge with the launch nose are also uniform battened built-up members.

Therefore, substitute truss members with effective stiffnesses have been used to develop the structural model.

Stage 11 Completed

Stage 12 (i): Add Panel of D8 Bridge

Stage 12 (ii): Jack up Nose at (2) and move Launch Skates to next node point

Stage 12 (iii): Pull forwards

Stage 12 (iv): Jack up Bridge and Nose at (1) and (2) and move Skates to next node point

Skates

Launch Beams

Stage 12 (v): Pull Forwards

Fi g. 2: Illustration of the launching procedure during stage 121

1

2

4

3 Stage 15 Stage 16

. . . .

Stage 17

Stage 30 Stage 31

The result of the analysis showed that for the load case self-weight plus the additional permanent load applied as a counter-weight, the maximum responses in the erection link members (link angles, diagonal link angles and top chord link angles) occurred at the end of launching stage 34. Moreover, the design of these members were checked according to the recommendation by EN1993-1-1:2005;2 the results showed that the most critical member had reached the ultimate limit state of buckling at the end of launching stage 34.

Stage 32 Stage 33 Stage 34 Stage 35 Stage 40 Stage 41 -Complete

Fi g. 3: Summary of the launching sequence1

Figure 3 shows successfully executed launches up to stage 34 and the remaining unfinished launching sequences as illustrated in the erection manual.1 Grid lines 3 and 4 represent the locations of the pier heads.

As the bridge had collapsed during construction, the governing load case is selfweight plus additional permanent load used as a counter-weight at launching stage 34. Thus, the assessment of selfweight and stiffnesses of truss members as accurately as possible was critical for obtaining more reliable values of the action effects in the structural members.

This was facilitated by available design documents in general and the schedule of parts for the main bridge and nose3 in particular. The self-weights of the plates and bolts were taken from the schedule and distributed as corresponding nodal loads in the structural model, while the self-weights of truss members were calculated using the software. The recommended counter-weight in the original design for launching stage 34 is 25 t per panel applied on the outer four panels as shown by the bold lines in Fig. 3.

To provide a clear launching path for the main bridge and the launching nose, the concrete side-span abutments were not constructed until after the main bridge was in place, supported by the pier heads.

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Figure 4 shows the action effects in the critical region at launching stage 34 for load case unfactored self-weight plus unfactored permanent load of a magnitude of 25 t per panel, applied in the outer four panels of the main bridge as a counter-weight. The most critically loaded members in the critical regionâ&#x20AC;&#x201D;link angles, top chord link Technical Report

419

Truss

IT

OT

IT

OT

IT

OT

IT

OT

No. L sections

1

1

2

1

2

2

3

2

Top chord

Diagonal Bottom chord

Truss

IT

OT

IT

OT

IT

OT

No. L sections

3

3

4

3

4

4

Top chord

Diagonal Bottom chord

IT, internal truss; OT, outer truss; make-up, number of L-sections used to build up the sections for internal or outer trusses.

Table 1: Built-up sections3

angles and diagonal link anglesâ&#x20AC;&#x201D;are shown in Fig. 4 with the member designations used in the schedule of parts.3 Table 2 shows the magnitudes of the action effects, the types of the cross sections, the lengths of the members and the batten spacings in the critical members.

Discussion of Results More than one structural model is required for a single launching stage because a typical cycle of the launching process consists of a number of substages, as shown in Fig. 2. The models for the different stages have been developed and analysed to determine the state of stress in the critical members during the initial stages of launching prior to stage 33. The analysis results showed action effects that were insignificant compared with those at critical stages 33, 34 and 35. Internal forces are more critical in other members during the initial stages of launching. However, the member resistances are all adequate at the initial stages. Thus, the launching continued without a problem until stage 34, when it suddenly collapsed after stage 34 was completed and stage 35 was about to be implemented.

At the end of launching stage 34, there were five panels (Main) to the left of pier 3 and nine panels (Nose) to the right of pier 4. Analysis of the intermediate and final steps during launching stage 35 -with five panels (Main) to 420

Technical Report

the left of pier 3 and ten panels (Nose) to the right of pier 4 for intermediate step and four panels (Main) to the left pier 3 and 11 panels (Nose) to the right pier 4 for final step-, resulted in internal axial forces in the diagonal link angles that were less than the corresponding value at the end of launching stage 34. Fig. 5 shows the internal forces in the critical members for launching staged 32 up to 36.

The following were observed:

plus counter-weight at the end of stage 34.

(2) Design checking using the Eurocode 3 (EN 1993-1-1:2005)2 and the action effects for the abovementioned load case and material strengths determined from laboratory tests (Ď&#x192; m = 517.0 MPa) showed that the diagonal link angle member reached the ultimate limit state of buckling about the y-axis at the critical action effect of (P/ M = â&#x2C6;&#x2019;1661.6 kN/2.9 kNm).

(1) The analysis result showed maximum values of the axial compressive forces for load case self-weight

Therefore, the diagonal link angles had been stressed to their ultimate limit state, causing them to buckle

Link angles

Top chord link angles Diagonal link angles

F i g. 4 : Action effects in the weak link region and the critical members

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Member

Cross-sections 750 mm

Link angle

Length of member and batten spacing (mm)

Action effects P(kN)/M33(kNm)

l = 7680 a = 2560

−3087.3/−25.4

Conclusions

y

z

double angle flanges made of 2 × L 200/200/20 750 mm

Top chord link angle

l = 4572 a = 1524

−3159.5/14.4

l = 5112 a = 1704

−1661.6/2.9

y

z

4-angle and 3-angle flanges made of 4 × L 150/150/10 and 3 × L 150/150/10 750 mm

Diagonal link angle y

z

double angle flanges made of 2 × L 150/150/10 P, Axial; M33, Moment about the major axis.

Table 2: Critical members’ cross-sections, lengths and action effects3 3500 Linkangle

3000

Diaglinkangle Topcrdlinkangle

2500 2000 1500 1000 500 0 32

elements, causing all of them to fail suddenly and, as a result, the collapse of the entire bridge.

33

34

35

36

Fi g. 5: Internal axial compressive forces (kN) in the critical members for launching stages 32–36 Members

P (kN)

M33 (kNm)

M22 (kNm)

Link angle

−28 866.8

2338.8/−2053.7

32.8/−38.8

Link top chord

−2638.2

1566/−430.9

Link bottom chord

−2704.8

−1795.5/2405.4

−16.2/2.8 2.5/−8.4

Detailed analytical investigation was conducted to determine the cause of failure of the Omo River Bridge during construction. Of the three highly stressed members (link angles, top chord link angles and diagonal link angles) in the region where the main truss was connected with the launching nose, the diagonal link angles were found to be the most critically loaded members.

The investigation is repeated for loading and boundary conditions corresponding to launching stage 33 to be able to determine the state of stress before collapse and reflect on available strength reserve and/or signs of imminent failure at that stage. The results of the investigation showed that the most critical member, that is, the diagonal link angle, had not reached its ultimate limit and possessed some strength reserve at launching stage 33.

It was thus concluded that the Omo River Bridge had collapsed during construction because the buckling resistance of the diagonal link angles, determined on the basis of the mean value of the yield strength and according to the Eurocode 32 for built-up sections, was less than the action effects in the members determined using unfactored loads coming from the self-weight and other permanent loads applied as a counter-weight at launching stage 34.

The result of the study indicated that the failure of the bridge could have been prevented with proper routine engineering analysis prior to the launching of the bridge.

References [1] Callender-Hamilton Bridges, Erection Manual for 128.0 m Span Omo River Bridge and 82.3 m Launch Nose, Ethiopia, 1988.

Table 3: Action effects in critical members after buckling of diagonal link angle member

[2] CEN. Eurocode 3: Design of steel structures— Part 1–1: General rules and rules for buildings, European Standard EN 1993-1-1:2005, European Committee for Standardization, Brussels, 2005.

and trigger the sudden collapse of the entire bridge. The adverse effect of the failure of the diagonal link elements can be illustrated by assuming

[3] Callender-Hamilton Bridges, Drawing Nos. AO-CHB-8701-2/3/4; AO-CHB-S112-15, Schedule of Parts for 128.0 m Span Omo River Bridge and 82.3 m Launch Nose, Ethiopia, 1986/87.

(conservatively) that these elements become ineffective once they have buckled. Internal forces increase many fold (Table 3) in all link

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Damage Investigation of a Tunnel Subjected to an Unplanned Surface Load Through Non-Linear Analysis Oriol Arnau, Researcher; Climent Molins, Assoc. Prof.; Antonio Marí, Prof., Department of Civil and Environmental Engineering, Universitat Politecnica de Catalunya, Barcelona, Spain. Contact: oriol@oriolarnau.com DOI: 10.2749/101686617X14881937384882

Abstract A description of the deformations and damage that occurred in a segmental tunnel lining as a consequence of an unplanned surface load is presented, as well as the numerical analyses performed for its safety assessment. The tunnel in study is located in soft soil conditions and presents a low overburden. Few months after tunnel drilling, a new bridge was constructed at surface level, placing an access embankment over the tunnel path. Monitoring points were installed along the affected section which recorded the deformation of the tunnel caused by the embankment weight. More worryingly, despite no additional loads were introduced, the deformation of the lining continued increasing significantly along the next months, indicating the existence of soil consolidation phenomena. As a consequence, structural cracks emerged along the affected section. Non-linear finite element models that realistically simulate the behaviour of the non-linear segment joints and the concrete cracking were used to characterize the actual response and strength of the lining. The results showed that most significant damages should occur at the non-visible extrados side, and characterized how they evolve along the soil consolidation process. Finally, the adopted tunnel strengthening procedure is described.

Keywords: forensic engineering; segmental tunnel lining; soft soil; structural damage; non-linear analysis; soil–structure interaction; structural strengthening.

Introduction When an abnormal response or a damage is detected in a structure it is of paramount importance to determine the causes and the possible Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: August 8, 2016 Paper accepted: December 16, 2016

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implications. Moreover, it is essential to answer key questions such as: whether the situation is stabilized or still evolving, if there exists a real risk of collapse and whether exceptional measures have to be taken in order to assure the structural stability.

The engineering approach for finding an appropriate response to these essential questions can be based on the analysis of the sensors data that reported the irregular response and/or the observed damage. Usually, an additional and more specific monitoring set-up is placed in the affected zone, about which recorded data could lead to a better explanation and determination of the problem. But, depending on the type or configuration of the affected structure, it could exist in several parts that are not visible or cannot be inspected, thus making impossible to adequately define the actual state of the structure. This situation is common in underground structures where it is not possible to directly evaluate the state of the extrados on direct contact with the ground, or in flooded structures in which service cannot be 100% interrupted like sewage tunnels or dams.

Therefore, in these cases it would be necessary to estimate the possible damages in the un-surveyed zones in order to appropriately assess the actual state of the structure.

The interest of obtaining the maximum possible information is clear: a better knowledge about the causes for the abnormal response and a more complete definition of the actual situation in terms of damage would allow taking more appropriate decisions about the necessity and urgency of structural interventions. Moreover, better suited interventions could be proposed in order to solve the disorder.

Nowadays, modern modelling techniques can be used to study and comprehend the complex response of underground structures, to determine the causes of the problem, and to estimate the damage at non-visible sides.

Advanced non-linear finite element

models allow to realistically reproduce the main phenomena involved in the actual response of the structures, being capable of considering the cracking and plasticity of reinforced concrete, the response of the joints between different members, or the influence of the soil– structure interaction. Consequently, they can be used to simulate complex damage processes, evaluate the stress redistributions, estimate the actual state and the remaining capacity of the structure, and to adequately design intervention or retrofitting processes.

Additionally, their repeatability permits easy analysis of different scenarios and conditions, allowing to determine the most influencing parameters, or to test different parameter values in order to cover possible uncertainties (materials, loads, etc.).

This paper presents the study and analysis of a Barcelona L9 tunnel section that presented significant movements that originated following the placement of a new bridge embankment on the tunnel surface (Fig. 1). Initial deformation measurements continuously became larger over time despite no other loads were introduced on the surface, indicating a soil consolidation phenomenon. Consequently, large deformations were experienced by the segmental tunnel lining, also presenting significant cracking at the intrados of the tunnel.

The situation at the time of study is initially exposed, and the movements and damages registered are then described. Later, two-dimensional (2D) non-linear numerical models that appropriately considered the joints, the soil–structure interaction, and the inelastic response of constituent materials, are defined in order to reproduce the structural response of the segmental rings at the damaged section.

Actual measurements and damage observations are used to validate the adopted modelling strategies and values. Numerical model results are presented and discussed, showing how they can contribute towards comprehending the causes of damages and the main governing mechanisms,

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Case Study The studied section is located at Barcelona’s Line 9 metro tunnel. The tunnel was constructed by means of a tunnel boring machine (TBM) and presents an internal diameter of 8.43 m. A 320 mm thick precast concrete segmental tunnel lining was used as a unique resisting structure, which was composed by six conventional segments plus one key segment. The construction of the tunnel was finished without remarkable incidences.

Few months after the tunnel drilling, a new road was planned in the case study zone, including a new bridge and a new access from an existent road. As it can be observed in Fig. 1, this new infrastructure implied the placement of embankments over the tunnel path, which achieved a maximum height of around 7 m over the original level at the bridge access.

In order to record the movements of the tunnel during the embankments construction, five different sections were initially instrumented by means of three topographic prisms (sections S1 to S5 in Fig. 1), located around the crown and at the mid-height point of the tunnel (see Fig. 2 indicators). The monitoring process started on 18 December 2008 and the embankments were erected in January 2009 (as seen in Fig. 2). Two more prisms were added to the bottom part of the monitored sections (starting at 19 March 2009) in order to obtain a more complete information about the increasing deformation of the lining.

Displacement records clearly showed that, despite no further loads were applied on the surface, the deformation of the lining was continuously increasing in the following months. As can be observed in Fig. 2a, maximum vertical crown movements increased from around 18 mm at the time of end of construction to above 80 mm 8 months later, thus defining an increase of over 300% in the lining deformation without performing any action. Slightly higher deformation increments were observed in the lateral deformation (Fig. 2b), increasing from around 4 mm to over 35 mm, and defining an increment above 700%.

The segmental tunnel lining presented a significant cracking at the crown intrados, as can be observed in Fig. 3 for the rings compressed between the

Tunnel

At the studied section, the tunnel runs through a geological layer of silt and clay (QL3), presenting an average depth of 25 to 30 m (see Fig. 1). The top of the tunnel is placed at the interface to a sand layer (QL2), which is preceded by a thin silt–clay layer (QL1) and an initial fill layer (F). The total original overburden of the tunnel was around 15.3 m, whilst the groundwater

table was located at 4 m depth. The basic properties of the soil layers are listed in Table 1.

0–7

to characterize the actual state of the lining, and to estimate the damages at the non-visible extrados side. Finally, the adopted tunnel strengthening procedure is described, and main conclusions are presented.

The crack patterns registered also showed different intensities and location depending on the position of the longitudinal joints, thus indicating a clear influence of the ring configuration in the damage generation.

Studies performed by local geological specialists indicated that the most feasible cause for the long-term deformations experienced by the lining could be found in a consolidation phenomenon experimented by the soil, since no significant additional loads were placed at the surface. According to the consolidation principle, when a saturated stratum like clayey soil is subjected to a stress increase (like the one caused by the embankment weight, or by the tunnel’s own deformation), the excess of pore water pressure is slowly dissipated due to the relatively low permeability of the

New embankments

4

S1

instrumented sections S3 and S4 (see Fig. 1). Maximum crack widths were always registered at the sections weakened by the lifting and grouting holes of the segments (indicated with circles in Fig. 3), achieving a maximum of 1 mm at ring 237 (the crack runs across the whole segment in the longitudinal direction of the tunnel).

Along the complete sections, crack widths of 0.1 and 0.2 mm systematically appeared, achieving 0.3 mm width in the majority of the rings along the affected zone (see rings 232 and 235 in Fig. 3 as example).

The maximum value registered at a complete section was around 0.5 mm, as can be observed at the crown central part of ring 235 (Fig. 3).

QL1

S2 ist

S3

Ex

QL2

11

ro

1.3

ad

t en

9.07 Tunnel

S5

8.43

New bridge

S4

Tunnel

QL3

Fi g. 1: Description of the case study: top view and soil section (Unit: m)

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Layer

Modulus of elasticity, E0 (N/mm2)

Total or submerged density, γ0 (kN/m3)

Poisson coefficient, ν

Cohesion, C0 (kN/m2)

Friction angle, φ0 ( )

Fill (F)

3–5

18–19 (T)

0.3

–

25–27

Silt clay (QL1)

9–15

10.1 (S)

0.3

5

29

20

10.1 (S)

0.3

5

31–32

12.0 (S)

0.35

5–15

25–32

Sand (QL2) Silt clay (QL3)

12–30 (depth depending)

tunnel linings as a consequence of the consolidation phenomenon.

Numerical Analyses This section gives the numerical reproduction of the structural response of the Line 9 segmental tunnel lining following the positioning of the new embankment and the progressive reduction of the lateral earth pressure caused by the consolidation process.

Table 1: Basic characteristics of the soil at the site1

soil, may also causing adjustments in the internal structure of the soil particles. As a consequence, the pressure increment caused by the embankment weight over the top of the ring could become gradually higher than in the lateral sides, thus producing a more unbalanced loading pattern as time goes by.

conditions, bending moments and deformations emerge. Segmental tunnel linings are multiple-hinged structures that require the reaction of the surrounding soil to find the equilibrium.

For soft-soil conditions, significant soil deformations are required in order to produce the reaction forces, thus originating significant displacements in the rings even for low hydrostatically unbalanced loads. Therefore, it is of paramount importance to find out the structural implications of the ovalization distortion increase (from circular to oval form) suffered by the segmental

For perfectly hydrostatic load conditions (i.e., constant radial pressure applied at the tube section), tunnels should only be subjected to compression forces, but out of these ideal

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–10

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Section no. 3 - Laterals

Section no. 4 - Crown

Section no. 4 - Laterals

Section no. 5 - Crown

Section no. 5 - Laterals

–90

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(b)

A 2D plane stress model is used for such purpose, which can be seen in Fig. 4. The height of the segments is divided in seven regular quadrilateral elements, defining a complete model presenting 4186 elements. The concrete properties and inelastic response are adopted from the Model Code 903 through an automatic option available in the software,4 assuming a C40 class concrete and using the Total Strain Crack Model.

A Von Misses plasticity model is used to reproduce the embedded steel reinforcement, adopting a modulus of elasticity of 200 000 N/mm2 and a yielding stress of 500 N/mm2.

Perfect bond is considered between both materials.

The unilateral response of the 206 mm in height joints is modelled through interface elements (shown in violet in Fig. 4) that do not present tensile stiffness, as proposed in Ref. 5. The soil– structure interaction is considered through radial and tangential springs related to each one of the external nodes of the mesh, as shown in Fig. 4.

When external loads are applied to the numerical model, they produce the movement and deformation of the tunnel since the ground reaction is provided by the springs. In order to avoid the unrealistic pull of the tunnel to the ground that can be obtained at certain tunnel zones, non-linear radial springs with no tensile stiffness are used. The radial stiffness value is determined from the modulus of elasticity of Es = 15 N/mm2 specified in the project document for the soil layer containing the tunnel at the axis depth (Table 1), using the equation of a circular beam in elastic media (Eq. 1), and assuming a tangential stiffness of 1/3 of the radial stiffness6 (Eq. 2).

10

5

Kr =

E R ð1 + νÞ

ð1Þ

Kt =

Kr 3

ð2Þ

0

F ig . 2: Vertical and horizontal displacements recorded at the instrumented sections 424

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convergence criterion of 0.05% in energy.

Ring 238 Section 4

Ring 237

Ring 236

Ring 235

Ring 234

Ring 233

Ring 232 Section 3

Fi g. 3: Survey of cracking of the rings between instrumented sections S3 and S42

In the absence of experimental evidences of the actual soil and water pressures acting on the studied tunnel section, the initial loading of the ring was determined according to the ground section and characteristics presented in Fig. 1 and Table 1, respectively. The calculated vertical and horizontal soil pressures are introduced in the model through nodal forces applied at the outer surface of the tunnel. The original conditions are introduced in the first load step, whilst the pressure created by the embankment is

applied later. The initial lateral pressure produced by the embankment is determined by using a lateral earth pressure value of K0 = 0.58, associated to a 25 friction angle soil. To reproduce the ground consolidation phenomenon, the K0 applied to the embankment vertical load is progressively reduced up to 0, according to what has been called “fictitious lateral earth pressure coefficient” (K0f). The analyses are performed in a finite elements software, using a regular Newton Raphson iterative procedure with a

K

r

K

t

Interface elements

Fi g. 4: Segmental ring: numerical model

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According to the definition of universal ring type used in this section of Line 9 tunnel, the ring presents a certain conical form that produces variable width of the ring. This geometry allows defining the tracking of the tunnel by modifying the relative position between the rings. In this particular case, 13 different ring placements are possible in order to create turns and depth variations. This originates different possibilities of joint locations in respect of the load position, as it can be observed in Fig. 5.

Joints present a significant influence in the structural response of segmental tunnel lings and, therefore, different structural responses can be presented for different joint positions.

A numerical study using linear material properties was carried out over the most common configurations presented along the damaged section.

Position 13 (Fig. 5a) emerged as the joint position configuration that represents a stiffer response of the ring, thus providing the smallest displacements but the highest bending forces.

Position 12 (Fig. 5b) behaves in the opposite way, thus defining the most flexible option for the same loading.

Consequently, non-linear analyses were performed over both configurations in order to cover a wide range of possible responses.

Figure 6a and c presents the vertical displacement at the crown and the maximum horizontal displacement numerically obtained for joint positions 13 and 12 (Fig. 5) when reducing K0f. As can be observed, ring joint position 12 presents a more flexible response to the applied load than position 13, increasing 12% the vertical displacement at the crown (114 mm in front of 102 mm) and 57% the maximum horizontal displacement (63 mm in front of 40 mm) when K0f turns 0. Consequently, for the stiff ring configuration (Fig. 6a), the registered vertical displacement at the crown (Fig. 2) would be achieved at K0f = 0.2, whilst maximum horizontal displacement at K0f = 0.05. For the flexible configuration (Fig. 6c), these points are found at K0f = 0.31 and K0f = 0.27 respectively, thus indicating that the proposed load configuration for reproducing the complex consolidation phenomenon tends to slightly Technical Report

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As can be observed in Figs. 6d and 7b, the inner reinforcement is the one in most demand in terms of stress in the flexible ring configuration. It rapidly achieves a value of about 300 N/ mm2 (K0f = 0.4), but later it presents a stable response around this value up to the complete elimination of K0f. The external reinforcement does not present significant activity since K0f approaches 0.35 (hence denoting the initiation of concrete cracking at the extrados), progressively increasing to 170 N/mm2 (Fig. 6d) for K0f = 0. This response could indicate that, for such particular configuration, the deformation of the ring would be based more on the rotation of the longitudinal joints than on the opening of the cracked sections, as can be seen in Fig. 7b.

The concrete and reinforcement stresses at the most unfavourable sections are used to determine the crack width (wk) at both intrados and extrados sides according to Eurocode 2.10 As can be observed in Fig. 6d, the crack width expected at the intrados side (crown) should be in the range of 0.2 to 0.3 mm, which perfectly fits the actual measures in the survey presented at the case study description (Fig. 3).

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(b)

Position 13 (stiff configuration) K

Position 12 (flexible configuration)

–30.11

–19.32

C

C

27.69°

A1

K °

A1

B

–27.99

A2

B

+27.99

38 55.

+15.81

These results clearly show the significant influence that the position of the longitudinal joints (between two adjacent segments of a same ring) has on the structural response of segmental tunnel rings. Concentrated rotations occur at longitudinal joints when rings deform and, depending on their relative positions in front of the load, the structural system can present a stiffer or more flexible response (presenting higher or lower deformation). This phenomenon could be especially relevant in soft-soil conditions, where larger ring deformations can be presented until the soil provides the necessary reaction to equilibrate the acting loads. Additionally, they also point out the difficulty to estimate the actual response of a tunnel section as a compendia of different ring positions, more so when the interaction between adjacent rings and the threedimensional response of the lining is not considered.7–9 Therefore, a complete range of K0f varying from 0.31 to 0.05 has to be studied in order to cover the possible scenarios affecting the actual tunnel state at the last registering date (Fig. 2).

(a)

–15.81

underestimate the actual lateral displacements of the ring.

A2

A4

13

A4

A3

+30.11

A3 12 +19.32

F i g. 5 : Analysed joint position configurations1

When analysing the maximum reinforcement stresses obtained for the stiff ring configuration (Fig. 6b), it can be seen how both internal and external reinforcements easily scale up to 300 N/mm2 (K0f = 0.4). From this point, the decrease of K0f does not produce a relevant stress increase at intrados reinforcement at the crown, but significant increments occur at the extrados reinforcement at the haunches, achieving the steel yielding stress for K0f =0.05 (view also Fig. 7a). Consequently, it has to be considered that, for the last registered ring movements, it would be possible that the extrados reinforcement could have achieved the plastic regime or could be close to.

These stress states define intrados crack openings at the tunnel crown as around 0.25 mm for all values of K0f under 0.4 (Fig. 6b), which equals the flexible configuration results, and are also completely consistent with the damage survey presented at the case study description (Fig. 3). On the other hand, maximum extrados crack opening significantly increases with the decrease of K0f, achieving an opening of around 0.4 mm for the last registered situation.

This response indicates that, for the stiff ring configuration, the deformation of the ring demands local rotations at the cracked sections of the haunches, thus concentrating the damage at the extrados. Figure 7a clearly shows wide concrete damage zones at the haunches (orange–red), as well as the ascension of the section neutral axis, and the increment of the compression stress at the intrados. Consequently, maximum steel reinforcement stresses and crack openings could be located at the non-

visible side, which is in contact with the soil, and do not allow appropriate survey or repair. In consequence, external reinforcement could be potentially affected by durability problems along its service life, fact that has to be considered when designing the strengthening of the lining.

Proposed Strengthening The objective of the strengthening of the existing lining was to provide enough structural capacity to resist long-term scenario, whilst improving the stiffness in order to avoid excessive deformations that could affect the serviceability of the tunnel. For adequately defining the long-term damage scenario, it was considered that the effect of the corrosion on the reinforcement at the cracked extrados subjected to the aggressiveness of soil and water in contact with the tunnel could lead to a reduction of 50% of the reinforcements’ section. The authors carried out verifications to ensure that the flexural capacity of the enlarged section, assuming the 50% loss of steel reinforcement, was equal or greater than that of the original intact section. The constructive design of the strengthening intervention was carried out by an engineering consulting company, on the basis of the presented conclusions and recommendations, using more conventional tools for the structural analyses and mandatory code checks.

The strengthening consisted essentially of enlarging the depth of the lining of the tunnel, by adding a continuous layer of 200 mm of reinforced sprayed concrete connected to the existing tunnel structure, in order to provide enough stiffness and capacity to the

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• Non-linear advanced numerical models, and their capacity for reproducing the actual inelastic response of the constitutive materials and joints, emerge as necessary tools in order to determine the causes of the damages, comprehend the main governing mechanisms, and to evaluate the actual state of the structure. In the present work, the results of such models helped to understand and characterize the influence of the joint positions in the structural system stiffness, and in the damage evolution process. In this case study, maximum variations of 12 and 57% were obtained at the vertical displacement at the crown and maximum horizontal displacements by varying only the ring joint positions.

• In the case study, numerical simulations demonstrated that the maximum structural damage could be located at the extrados side, being practically impossible to be directly evaluated or repaired. For the stiff ring configuration analysed, average crack opening at the intrados side should be located around 0.25 mm,

The strengthening was applied in 2010 and since then no new cracks and no relevant increments of deformations have been observed.

lining, preventing further deformations and cracks in the long term. The characteristic compressive strength of the sprayed concrete was 30 N/mm2. To prevent early-age cracking of the lining due to thermal and shrinkage strains, circumferential joints were placed every 7.5 m or less, corresponding to five segments, in the longitudinal direction. This was consistent with the designed behaviour of the strengthening as a ring.

Conclusions

A single layer of reinforcement was placed close to the intrados with a cover of 50 mm, composed by circumferential 10 mm rebars at 150 mm and longitudinal 8 mm rebars at 250 mm.

The reinforcement also included connections made of steel dowels to improve the shear-friction capacity, distributing two dowels of 10 mm diameter (forming one U-shaped rebar) every 400 × 400 mm. This U-reinforcement was inserted 200 mm into the existing segmental lining for a good anchoring. Sand blasting was also applied in the intrados surface before spraying concrete to reduce rebound and to improve shear-friction capacity.

• The increasing deformations and damages observed at the interior of the lining were mainly caused by the construction of a new embankment over the tunnel path, and the related consolidation phenomenon experienced by the soft soil surrounding the tunnel.

• Internal surveys and measurements could not be enough to adequately define the actual situation of an underground structure presenting an abnormal response and/or damages, since the structural state at non-visible sides (extrados in contact with the soil) cannot usually be directly characterized.

The study and analysis of the abnormal response and damages that appeared in the tunnel section in this case study pointed out the following:

Ring joints position 13 (stiff configuration) (a)

(b)

100 80 60 40

Crack opening, Wk (mm)

120

Max. numerical horizontal Max. registered horizontal

20

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0.45 Displacement (mm)

Numerical crown vertical Max. registered crown vertical

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Reinforcement stress (N/mm2)

140

0

0.6

0.5

Value of k0f

0.4 0.3 0.2 Fictitious lateral earth pressure, k0f

0.1

0

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Ring joints position 12 (flexible configuration) (c)

(d) Numerical crown vertical Max. registered crown vertical

60 40 20

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0.05 0

0 0.6

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Fictitious lateral earth pressure, k0f

Fi g. 6: Structural response of the lining when reducing K0f for the two simulated configurations

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(a) Concrete stress (N/mm2)

Concrete segments

4 1.9 –.19 –2.29 –4.38 –6.48 –8.57 –10.7 –12.8 –14.9 –17 –19 –21.1 –23.2 –25.3 –27.4 –29.5 –31.6 –33.7 –35.8 –37.9 –40

Ring joints position 13 Steel reinforcement

(stiff configuration)

(b) Reinforcement stress (N/mm2)

Concrete segments

500 467 433 400 367 333 300 267 233 200 167 133 100 66.7 33.3 0 –33.3 –66.7 –100 –133 –167 –200

Ring joints position 12 (flexible configuration) Steel reinforcement

F ig . 7: Circumferential stress state for K0f = 0.1; deformation factor = 10

which is fully consistent with the information reported in the damage survey. At the extrados side, the average crack opening for the last registered deformations could be located around 0.4 mm, obtaining a maximum reinforcement steel stress close to its yielding value. Therefore, for the last registered ring movements, extrados reinforcement steel could have achieved the plastic regime or could be close.

• The most complete knowledge about the causes of the damage and about the actual state of the structure achieved by using advanced numerical models allows taking appropriate decisions about further actions to take, more so to design better interventions and retrofitting processes. In this case study, the conclusions about the possible wide crack openings at the extrados side led to appropriately consider the 428

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cracking and the long-term corrosion of the reinforcement in the design of the strengthened section.

Acknowledgements The present work was carried out in the framework of the contract subscribed between the Technical University of Catalonia (UPC) and the public company Infraestructures de la Generalitat de Catalunya SAU for research and innovation in the construction of Line 9 of the Metro of Barcelona.

References [1] Line 9 Metro of Barcelona. Project documents (in Catalan).

[2] COTCA. Damage categorization of a Line 9 tunnel section (in Catalan), 2009.

[3] CEB-FIP. CEB-FIP Model Code 1990. Thomas Telford: London, 1993.

[4] TNO. Diana user’s manual. Release 9.4, 2009. http://www.tnodiana.com

[5] Arnau O, Molins C. Experimental and analytical study of the structural response of segmental tunnel linings based on an in situ loading test. Part 2: Numerical simulation. Tunn.

Undergr. Space Technol. 2011; 26: 778–788.

[6] Molins C, Arnau O. Experimental and analytical study of the structural response of segmental tunnel linings based on an in situ loading test.

Part 1: Test configuration and execution. Tunn.

Undergr. Space Technol. 2011; 26: 764–777.

[7] Arnau O, Molins C. Three dimensional structural response of segmental tunnel linings.

Eng. Struct. 2012; 44: 210–221.

[8] Arnau O, Molins C, Blom CBM, Walraven JC. Longitudinal-time dependent response of segmental tunnel linings. Tunn.

Undergr. Space Technol. 2012; 28: 98–108.

[9] Arnau O, Molins C. Theoretical and numerical analysis of the three dimensional response of segmental tunnel linings subjected to localized loads. Tunn. Undergr. Space Technol. 2015; 49: 384–399.

[10] CEN. EN 1992-1-1, Eurocode 2: Design of Concrete Structures, Part 1-1: General Rules and Rules for Buildings CEN: Brussels, 2004.

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Investigation of the Collapse of a Building in the Final Stage of Its Construction Miodrag R. Grbic, Dr. Teaching Assistant; Neda Dzombic, Teaching assistant; Milan T. Glisic, Prof., Dr. Faculty of Architecture–Architectural Technologies, University of Belgrade, Belgrade, Serbia. Contact: grbic.miodrag@gmail.com DOI: 10.2749/101686617X14881937384927

Abstract In 2009, the Municipal Court of the

City of Cajetina, Republic of Serbia, issued an order to a Commission of court-appointed experts (Commission) to perform an analysis of the collapse of a building in the final stage of its construction. The Commission was appointed by the Faculty of Architecture, University of Belgrade. The study reveals the steps taken in the process of determining the cause of collapse and the problems that have led to it, the conclusions and findings. Poor design and bad execution led to the collapse of the building. The causes of collapse were structural defects in calculations and rebar detailing, disregard of the procedures related to technical supervision and the procedure of obtaining a building permit, violation of building codes and good practice. The cause was determined using the following methods: analysis of documentation, analysis of regulations and legislation, in situ observations, interviews with participants on the process, and structural analysis through three dimensional modelling. The study also shows that safety factors in the built construction elements were low and inadequate. This work summarizes the principles, conclusions and findings of forensic investigation.

Keywords: assessment; column; concrete; structural analysis; expert report.

Introduction On 15 July 2009, the Municipal Court

of the City of Cajetina, Republic of Serbia, issued an order to a Commission of experts to draw up an expert report on a building that collapsed in the final stage of its construction (Fig. 1).

Fi g . 1: Collapsed building in final stage of construction

was a reinforced concrete frame construction, and the segment above the ground was a solid masonry wall structure. Only the façade had been unfinished before the building collapsed.

After the collapse, the building suffered extensive damage. The underground frame structure was crushed under the weight of the upper floors (Fig. 3b).

Steps to Be Taken in Order to Establish the Facts and Causes of Collapse The Commission used the common steps1 that are necessary for an expert investigation process. The goal of these activities is to establish the following:1 the mode and sequence of the failure; the demands acting on the facility at the time of the failure; the capacity of certain components or of the entire facility at the time of the failure.1 Establishing the capacity of components includes determining the asdesigned and as-constructed state of the facility in cases involving fatigue, fracture or deterioration.1 The hazards that lead to collapse can take many different forms, ranging from design actions and accidental actions, deterioration processes, to human errors associated with the design, execution and use of the structure.2

Description of the Building Before collapsing, the building structure had included an underground garage, ground floor, two upper floors and an attic. Gross floor area was 1245.83 m2. The underground segment

The Process of Determining the Cause The Commission used the following methods: analysis of documentation,

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in situ observations, interview with participants on the process, analysis of regulations and legislation, and structural analysis.

Provided Documentation • Final design, report on the technical supervision of the final design, certificate for materials incorporated into the structure of the building issued by the supplier (provided by the court).

• Final design, report on the technical supervision of documentation, and construction contract (provided by the representative of the investor).

• Construction drawings, structural analysis and calculations without rebar detailing (provided by the design office, not certified by the project designer).

Documentation Submission Procedure The final design was dated August 2008 and had a positive technical supervision report. A confirmation of documentation submission (final design) was issued on 15 June 2009. It was issued based on the approval for construction issued by the department of construction land, roads, investments, urbanism and property affairs on 18 August 2008. This raises the question: how is it possible that the execution of work (September 2008) started before the documentation submission was confirmed? General documentation included the final design with corresponding authorizations inadequately certified by seal and date. Confirmation of documentation Technical Report

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submission, issued on 15 June 2009, did not include data from the authorized supervising engineer and authorized contractor responsible for the same.

was in accordance with the current legislation. It can be concluded that technical supervision was not carried out professionally and responsibly.

Structural Design

Overview of Certificates

Basic construction design logic applied for the part of the building above the ground does not correspond with the basic construction design logic applied for the underground part of the building. The design of the underground part of the building involved only a partial structural analysis and no analysis was done for the whole underground structure. There was no structural analysis for the columns that transfer the load to the foundation. There were no beams and columns to support the semi-prefabricated ceiling or the masonry walls of the upper part of the structure (Fig. 2).

The quality of the concrete determined by the certification was above the values required in the documentation. Certificates for steel rebars were provided, and they confirm that the quality satisfies the design requirements and standards.

Report on the technical supervision of the documentation was certified by the authorized engineers, which should mean that the documentation

Observations in situ Through in situ observation, the commission determined that the building which was never tested with live load collapsed into the underground garage area (Figs. 1 and 3b). The damage extent of the building was total.

Peripheral walls of the underground floor were made from concrete hollow blocks filled with concrete. The walls were strengthened with vertical tie columns and horizontal ring beams.

A hydrocarbon waterproofing

membrane interrupted the monolithic connection between the walls and the foundation. Peripheral walls were deformed, substantially displaced and were not able to resist horizontal pressure from the soil.

Only four rebar anchors were left for each connection between tie columns and beams. As a consequence, no rigid joints could be formed. Rebar anchors from the beams were anchored in the peripheral walls and inside the columns (Fig. 3a) with a length of only 100 mm. The anchoring quality contradicts Articles 149 and 150 of Technical Norms (PBAB â&#x20AC;&#x2122;87).

Columns at the underground level (reinforced with only four rebars 14 mm) were crushed under the load.

Interviews with Participants on the Process The interviewed participants were the architectural design engineer, construction design engineer, supervising architectural engineer, the authorized technical supervision engineer, and the investor.

F ig . 2: Disposition of elements (Units: [cm]) 430

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(a)

(b)

Fi g. 3: (a) Collapsed column and (b) collapsed building with construction machinery

Experience based on eyewitness accounts of failures is that the accounts are often not accurate or reliable.1 This is not because eyewitnesses are dishonest, but because most of them do not have the necessary technical training, and when involved in catastrophes, it is often that they are distracted by other issues and come away with a distorted view of the events.1 An exception to this is the information obtained from some construction workers, when failure occurs during construction1.

All interviewed participants confirmed that the construction site documentation was inadequate, and no valid decisions were made as to who would be the supervising engineer or authorized contractor, so the authorized technical supervision engineer led the process of construction. The authorized technical supervision engineer gave contradictory information and stated that he participated in the preparation of the final design. He stated that he presumed the concept drawings were used in construction, and he did not provide any additional drawings in the documentation on the site or give any instructions/ orders for the execution of work. This contradicts his initial statement, confirmed by all the interviewed parties, which indicates he was in charge of the process of construction.

Considering this Article and sections “Provided Documentation” and “Interviews with Participants in the Process”, the authorized technical supervision engineer did not perform the delegated job responsibly, including the verification of compliance between documentation and regulations.

Article 114 defines the steps for construction commencement.

As mentioned in “Provided Documentation”, the execution of works started before the confirmation of documentation submission was issued, that was in violation of the Article 114, paragraph one. All interviewed participants confirmed that there were no valid decisions on

the appointment of the responsible authorized supervising engineer and responsible authorized contractor.

Article 118 defines the obligations of the contractor.

In violation of Article 114, the contractor was not registered, which led to violation of Article 118 reflected in the fact that an unregistered contractor was unable to provide the name of an authorized contractor.

Article 120 defines the obligations of supervision.

The confirmation of documentation submission did not include data regarding the supervising engineer that should be provided by the investor.

Overview of Valid Regulations and Legislation in the Republic of Serbia Related to the Identified Problems Mx [kNm/m] –203.16 –135.44 –67.72 –0.00 112.24

Law on Planning and Construction (see Ref. 3) Article 111 defines the rules for technical supervision.

F i g. 4: 3D structural model

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432

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Table 1: Load capacity estimate for columns C1, C2 and C3, at the underground level

N = nominal normal force; M = nominal bending moment; Nul = ultimate normal force; Mul = ultimate bending moment; MB = concrete quality; l = column length; b, d = dimensions of cross section; a = protective layer of concrete; n = number of rebar; Nr = real normal force at degradation; M = real bending moment at degradation; kr = real safety factor; k = safety factor required by regulations (PBAB ’87).

1171.56 C3, B and II

13.66

2284.54

26.64

25

3.2

0.3

0.3

0.04

14

4

1886.21

21.99

161

1.61

1.95 0.6

0.77 77

60 45.64

43.11 990.27

519.71 4

4 14

14 0.04

0.04 0.3

0.3 0.3

0.3 3.2

3.2 25

25 109.18

148.34 1689.05

2507.82 55.99

Mr (kNm) Nr (kN) n Rebar ( mm) a (m) d (m) b (m) l (m) Mul (kNm) MB (MPa)

866.18

76.07

k kr

Real

Required by regulations (PBAB ’87)

Safety factor

Percentage of nominal force that element can withstand

1286.06

The consequences of structural failures as a result of hazards can take several forms: from material/structural damage, human injuries/fatalities and/or environmental impact.2 This collapse did not result in human injuries/fatalities, nor had a direct environmental impact. Environmental impact is reflected in the emission of greenhouse

C2, C and IV

Structural analysis confirms the logical discrepancies in the structural system, as identified in Structural Design section. The structure of the building was unable to transfer the loads to the ground. The above-ground segment of

Damage Assessment

C1, C and II

Load on structure was determined according to regulations.8

The consequence of an initial fracture is the progressive collapse of the entire building. Column calculations missing from the documentation are a crucial fault that led to the collapse.

Nul (kN)

Structural analysis of the collapsed building was performed using data collected from the available documentation and determined in situ. The structural analysis was done using a 3D modelling software (finite element analysis) for determining the effects of loads on the building.

M (kNm)

Structural Analysis–Three Dimensional (3D) Structural Modelling

All of these columns should withstand about 195% of nominal load before collapsing (regarding the commonly used safety factor of 1.95 [PBAB ’87]).

Calculation shows that the column in axes C and IV should be the first one to collapse followed by the rest of the structure (Table 1).

N (kN)

Articles 149 and 150 define the minimal length of rebar anchoring, and the absolute minimum is 150 mm; if the quality of concrete is taken into account as ls,min, it is 220 mm. Dimensions were determined in situ (Section Observations in situ) with the length of 100 mm.

Safety factor for reinforced concrete elements under compression, defined

by Norm (PBAB ’87) Article 80,6 is 1.9 for permanent load and 2.1 for live load (universal factor commonly used is 1.95). Usually, factors of safety and the general resilience or redundancy of structures provide sufficient margins to offset these.7

Pos

Article 41 defines the obligations of the concrete supplier (regular testing for the concrete they produce, frequency of testing for concrete compressive strength, etc.), and they were fulfilled by the supplier for concrete (confirmed by the certificates).6

Column dimensions

Technical Norms PBAB ’87 (see Ref. 6)

Comparative analysis of the graphical documentation and the in situ observation shows structural degradation of the column in axes C and II, B and II, and C and IV, which resulted in the redistribution of load. The cause of structural degradation of the remaining columns was redistribution of excessive load.

Ultimate (safety factor 1.95)

Chapter XVI and Article 94 of these regulations define the rules for solid masonry structures in seismic areas.5 Combination of vertical load-bearing concrete elements with structural masonry walls is not allowed, not even on different levels of the building.

Nominal

Regulations on Technical Standards for Construction in Seismic Areas (see Ref. 5)

Calculations show that columns in the underground segment had dimensions that are below the required values and the reinforcement was not sufficient (Figs. 4 and 5).

Real safety factor at degradation

Integral parts of the final design were missing, such as calculations for all the structural elements of the building, material specifications, deformation calculation, formwork plans for concrete structures, reinforcement plans and rebar detailing.4

There must be a continuous and sensible load path from the point of application of a load all the way to the foundation. Although this seems evident, load path issues do arise when the design analyses and detailing lose sight of the overall structural concept and become inconsistent with it. They also arise when the structural concept itself is inconsistent with the loads it needs to resolve and resist.7

Executed structural properties

Article 29 defines the numeric and graphical content included in the documentation of the final design.

the building was randomly set on the underground segment.

3D model analysis

Regulations of the Content and Preparation Method of Technical Documentation for Building Construction (see Ref. 4)

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1.58

1.60

3.30

2.60

3.30

1.60

1.58

Load combination: I+II+III

2 –494.64

2.54

–227.67 –156.81 –29.40 b/d=30 / 50

1

–55.99

350.72

–129.83

–145.36

–107.80 –13.66

0 52.21

b/d=25 / 30

347.50

b/d=30 / 30

b/d=30 / 30

b/d=25 / 30

93.24

51.04

–54.55

b/d=30/35

30.84

203.87

7

6

5

4

3

2

1

33.64 0

2.70

242.35

Columns C1 and C3 Frame: V 2 Bending moment: max M3= 350.72 / min M3= -494.64 kNm

1.58

1.60

3.30

2.60

3.30

1.60

1.58

Load combination: I+II+III

2

2.54 198.74 1 –18.40

b/d=25 / 30

b/d=30 / 30 b/d=30 / 35

7

–1286.06 4

3

2

1

0

2.70

–111.4

–1171.56

6 –153.64

–87.07

5

0

b/d=30 / 30

b/d=25 / 30

b/d=30 / 50

Frame: V 2 Columns C1 and C3 Normal stress: max N1= 198.74 / min N1= -1286.06 kN

Fi g. 5: Diagram showing bending moment and normal stress on columns C1 and C3

gasses due to the removal of debris and construction of a new building.

The extent of damage in this case is total damage. The prices in this estimation are based on the average prices of construction. Total estimated price is €521 323.70.

Conclusions This paper outlines the steps followed while performing an investigation into the collapse of a building, which was carried out in accordance with the principles listed in Steps to Be Taken

in Order to Establish the Facts and Causes of Collapse. Reasons for the collapse were structural defects resulting from inadequate calculations and rebar detailing, disregard for procedures related to technical supervision and the procedure of obtaining a building permit, and disregard of building codes and good practice.

An adequate load path was not provided in the design, so the load could not be transferred to the foundation without degradation of the building structure.

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Column calculations missing from the documentation are a crucial deficiency that has led to the collapse. Ignorance of this fact and adoption of unverified data and details in the building execution process are also major violations.

Safety factors were negligible and completely inadequate. Progressive collapse occurred as a result of column degradation due to miscalculations in the final design, and disrespect to procedures.

The collapse is directly related to the accumulation of errors such as use of incomplete, inadequate, inconsistent, Technical Report

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unapproved drawings and violation of construction standards.

Building Construction. Official Gazette of the Republic of Serbia, No.34/2006; 2008.

The underlying source of a failure is sometimes found in ignorance, incompetence, negligence and avarice, the “four horsemen of the engineering apocalypse”.9,10

[5] Regulation on Technical Standards for Construction in Seismic Areas. Official Gazette of SFRJ, No. 52/90; 1990.

References [1] Bell GR. Civil engineering investigation. In Forensic Engineering, Carper KL (ed) CRC Press LLC: Boca Raton, 2001; 231–232, 239.

[2] Janssens V, O’Dwyer DW, Chryssanthopoulos MK. Assessing the consequences of building failures. Struct. Eng. Int. 2012; 22: 99–104.

[3] Law on Planning and Construction. Official Gazette of the Republic of Serbia, No.34/2006; 2006.

[4] Regulation of the Content and Method of Preparation of Technical Documentation for

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[6] Technical Norms for Concrete and Reinforced Concrete PBAB ’87. Official Gazette of SFRJ, No. 11/87; 1987.

[7] Abboud NN, Jain A, Levy M. Design errors, construction defects, and project miscommunication. In Forensic Structural Engineering Handbook, Ratay RT (ed), 2nd, par. 8.2 edn. The McGraw-Hill Companies: New York, 2010.

[8] Standard for Design of Building Structures, Live Loads in Residential and Public Buildings.

Standard JUS(SRPS) U.C7.121. Official Gazette of SFRJ, No. 49/88; 1988.

[9] FitzSimons N. An historical perspective of failures of civil engineering works. In Forensic Engineering: Learning from Failures ASCE: New York, 1986.

[10] Carper KL. Learning from failures. In Forensic Engineering, Carper KL (ed) CRC Press LLC: Boca Raton, 2001; 19.

SEI Data Block Owner: Begining of work execution Date of Building Collapse: Percent of executed works (%) Damage extent Gross floor area (m2) Concrete(m3) Masonry (m3)

Private owner September 2008.

Estimated cost (€)

521 323.70

June 2009.

90 total 1245.83 347 212

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Improved Structural Health Monitoring of London’s Docklands Light Railway Bridges Using Digital Image Correlation Jan Winkler, Dr., Chief Specialist, Atkins – Highways and Transportation, Copenhagen, Denmark; Chris Hendy, Head, Bridge Design and Technology, Atkins – Highways and Transportation, Epsom, England.

Contact: jan.winkler@atkinsglobal.com DOI: 10.2749/101686617X14881937384648

Abstract As our transportation infrastructure ages and the challenge of keeping it serviceable grows, the need for improved condition information on which to make good cost-effective maintenance decisions becomes ever more vital. Gathering this condition information requires structural health monitoring and inspection on a grand scale and, for it to be useful, it must be accurate, inexpensive, easy to interpret and avoid interfering with traffic flows—whether rail or highways. Digital image correlation (DIC) is a noncontact photogrammetry technique that can be used for monitoring by imaging a bridge periodically and computing strain and displacement from images recorded at different times or operating conditions. This paper discusses the use of DIC for determining the in situ stress state for steelwork details on the London’s Docklands Light Railway bridge structures, including a fatigue-sensitive existing detail with a stress concentration and an impact damaged area, with a view to avoiding the need for strengthening.

Application of DIC technique in asset management proved to be cost effective. Throughout the monitoring, the structures were left untouched and were in full service operation.

Keywords: bridge monitoring; performance and damage assessment; digital image correlation; strain and deformation measurement; asset and fatigue management.

such as the need for external power and cabling/antenna for data transmission, high data acquisition channel counts and the limitation of only measuring at discrete points or along a line, so it is necessary to have an idea of where to expect damage when placing the instrumentation. These sensors can be used effectively to continuously monitor for abnormalities that indicate damage, but the type and severity of the damage can still be difficult to identify from discrete point measurements.

interventions defined if they are exceeded. This means that the structural engineer needs to identify what the most likely deterioration mechanisms are for the particular bridge and design bespoke monitoring that can be used to specifically measure performance in such a way that an acceptable level can be set and checked, e.g. movement at a bearing or joint, tensile crack lengths/widths at particular discontinuity regions, wire breaks in a cable, fatigue cracking at a steelwork detail, tilt of a column or pylon or deflection of a deck where a bridge is very sensitive to creep variations and other material parameters. This may require the use of many different systems.

Digital Image Correlation Digital image correlation (DIC) represents a photogrammetry technique used for accurate measurements of surface deformation. The digitized images (e.g. of a bridge deck) are compared to match facets from one image to another by using an image correlation algorithm (Fig. 1a and b). Image

Instruments such as strain gauges, accelerometers, fiber optic sensors and displacement transducers are becoming increasingly common in structural health monitoring. These types of sensors can however possess drawbacks

CAM

(a)

(b)

Undeformed state

Deformed state

1mm

Introduction Bridge Monitoring—Challenges The principal difficulty with adding several monitoring systems to bridges is that they produce vast quantities of data.

Bridge operators typically do not know what to do with this data unless there are very clear trigger levels defined associated with this data and clear

DIC

(c)

facet: 15 x 15 pixels

(d)

y

y

x

x

Fi g . 1: Digital image correlation method (Unit [−])

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analysis involves capturing a reference image of a bridge component surface in its nondeformed state. As the load is applied (e.g. truck load), additional images are collected. The algorithm (which can either run in real-time or post-process), involves a stage-wise analysis, in which each stage consists of one image resulting in a description of displacements occurring on the surface of the bridge component. The evaluation of a correlation measurement results in coordinates, deformations and strains of the surface (Fig. 1c and d). DIC method allows high-precision surface deformation measurement that can reach the accuracy of a few micrometers.

DIC measurement can provide information about strain (all directions), vertical and horizontal displacement, crack size, rotation and acceleration. DIC is independent of scale (local and global monitoring) and material (concrete, steel) and is especially helpful in monitoring of bridge components with a difficult access. Real-time operation, off-line analysis, remote access to equipment and live reporting of results is also possible. As the technology stands, DIC can be operated with as little as one individual and does not have to be in contact with the bridge, thereby avoiding any potential conflicts with traffic or difficult geography. Also, DIC takes a fraction of the time to setup compared to traditional instruments.

measurement techniques and DIC is an easier way to measure displacement of multiple points simultaneously. DIC was also proposed as a method to assess dynamic characteristics of suspension bridge hanger cables.14 In this study, a noncontact sensing method to estimate the tension of hanger cables by using digital image processing based on a portable digital camcorder was proposed. Moreover, DIC technique has been used to record the strain on a concrete girder during a full-scale bridge failure test15 and for the measurement of the displacement field on a cracked concrete girder during a bridge loading test.16 In both cases the photogrammetry method was able to detect a change in loading condition and locate cracks. Finally, DIC was used in fatigue testing of monostrands for bridge stay cables. Here, the visionbased system allowed for the measurement of the interwire movement (fretting fatigue) being the governing mechanism responsible for the fatigue life reduction in modern staycable assemblies.17,18

DLR Bridges—Fatigue Assessment

This paper discusses the use of DIC for determining the in situ stress state for steelwork details on the Docklands Light Railway (DLR) bridge structures, including a fatiguesensitive existing detail with a stress concentration and an impact damaged area, with a view to avoiding the need for remedial measures.

Measurement Methodology

DLR has implemented the running of three-car trains on the DLR network between Poplar station and Stratford station, known as the north route.

The “DLR Capacity Enhancement & 2012 Games Preparation Project” report identified that the fatigue life of some of the underbridge structures should be reviewed. Atkins was commissioned to prepare fatigue management strategy for these underbridge structures. In case of the Wick Lane and Warton Road Bridges, located close to the Olympic Stadium (Fig. 2), DIC monitoring was used as a part of overall assessment strategy. Both bridge structures had their own site constraints and monitoring challenge.

DIC was used for each structure, one using a local monitoring approach to investigate stresses at an existing bracket detail and the other using a global monitoring approach to compare deflections to those expected.

The DIC system comprised of a camera mounted on a geared head on top of a surveyor’s tripod and the pan. Lens was adjusted to provide required field of view.The sampling frequency was 20–40 Hz and the system was set to have a resolution higher than 1/50th of a pixel. The normalized grey-scale

Image Analysis in Bridge Engineering Research activity on the application of photogrammetry in bridge-related projects has been minimal and widely dispersed within the last 25 years.1 Early applications of this technique include measurement of the deflection of a continuous three-span steel bridge under dead load2 and identification of the bridge deformation.3 DIC was also employed to measure the geometry of a suspension bridge4 and deformation of steel connection of a pedestrian bridge.5 It was shown that DIC is a complementary tool of the conventional measuring systems such as Linear Variable Differential Transformers (LVDT) and strain gauges. The last few years have seen an increased application of the DIC in the measurement of the vertical deflections of steel and concrete bridges due to traffic6–12 and train transit.13 In general, it was concluded the DIC system accuracy is comparable to existing displacement 436

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AQUATICS CENTRE

OLYMPIC STADIUM

WARTON BRIDGE

WICK LANE BRIDGE

F i g. 2 : Docklands Light Railway (DLR) bridges

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correlation tracking algorithms were used for the image analysis. Calibration was achieved by measuring a distance within the image. The existing features of the bridge surface were used as the tracking objects on the Warton Bridge while in case of the Wick Lane Bridge additional speckle pattern was applied.

No editing of the raw data has been undertaken. All “processing” was related to optimizing contrast levels and exact positioning of reference points for the virtual strain gauges.

(a)

Validation and Calibration of Image-based Sensors

(b)

1'3" 1'3" 2'0"

4'0"

4'0"

4'0"

4'0"

4'0"

4'0"

4'0"

4'0"

3'0"

2'0" 1'3"1'3"

GIRDER B

GIRDER A

11'4"

GIRDER A

GIRDER B

A comprehensive validation (crosscheck) of image-based sensors was conducted by the author and can be found in a previous publication.17 Two analyses were performed indicating good agreement between measurements made with traditional strain gauges and those made using image-based sensors.

Calibration is a measuring process during which the image (pixel) coordinate system is adjusted such that the dimensional consistency is ensured.

The field of view (the area of the structure captured on the camera’s image) was treated as a grid of discrete elements and the calibration in case of both DIC monitoring projects was done by specifying real-world distances between points in an image.

The relevant distances were taken from as-built drawings. Locations of interest (girder and fatigue detail) in a field of view were described as a system of continuously varying coordinates. Two-dimensional (2D) planar coordinate system was used as both the girder and fatigue detail had no movement in the third dimension compared to the two being measured.

3'0"

3'343"

4'821"

3'343"

Fi g . 3: Plan view (a) and cross-section (b) of the Warton Road Bridge (Unit [−])

(a)

(b)

WARTON BRIDGE

Displacement 1

Warton Road Bridge

(c)

Deck Displacement 1 0.2 0.1

Displacement (mm)

The structure is a single-span halfthrough simply supported bridge constructed in 1937. The span of the longitudinal girders is 13.56 m. A plan view of the structural layout taken from as-built drawing records is shown in Fig. 3a. The structure consists of two longitudinal girders and a concrete slab with encased steel beams. Each longitudinal girder is formed from flange and web plates that are riveted together using angle sections. In addition to the DLR track, the southern girder also supports a walkway cantilevered from the

0 –0.1 –0.2 –0.3 –0.4 –0.5 –0.6 –0.7 3

4

5

6

7

8

9

10

Time (s)

Fi g . 4: Warton Bridge (a), close up on main beam and DIC measurement (b), analysis of DIC data (c)

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(a) East abutment

38'5"

41'5"

Wick Lane Bridge

Central pier

38'5"

41'5"

West abutment

(c) Bolts

Notched rivets

Bolts

(b)

5'5"

5'5"

5'5"

F ig . 5: Plan view (a) cross-section (b) and notch detail (c) of the Wick Lane Bridge (Unit [−])

girder top flange. The typical crosssection is shown in Fig. 3b.

The structure carries a single DLR track and the alignment is effectively straight in plan. Three-car trains are in operation on this line. The abutments are brick abutments founded on spread footings.

Global Monitoring of a Bridge (Displacement Measurement) A fatigue management strategy for the Warton Road Bridge (Fig. 4a) involved detailed site-based works using nondestructive testing (NDT) to check for damage assessment. For this purpose DIC technique was employed to capture the deflection of the main beam (Fig. 4b), without requiring closure of the road underneath which was the main route used for a large nearby shopping centre, and to 438

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deflections measured with DIC system at midspan (0.6 mm) are of a similar magnitude to the assessment deflection (0.7 mm) for a fully loaded train.

It is more realistic that the train was less than 100% capacity loaded with passengers and that there was some dynamic magnification on the static loads.

compare this to the deflection calculated under the fatigue assessment loading. The camera was located 30 m away from the bridge and the measured peak deflections at midspan (Fig. 4c) were of similar magnitude to the deflections from the fully factored assessment loading assumed in the finite element (FE) analysis. In case of the Warton Bridge, it was possible to set up, record data on deflections, pack up and leave the site within few hours. This was possible due to a useful recognisance site visit to plan the monitoring in advance. The dynamic monitoring using DIC technology was performed in summer 2015.

For comparing deflections on main girders, the authors looked at midspan moments using the original simply supported line beam with traversing DLR three-car trains. The peak

The structure is a two-span bridge constructed in 1948. Each span is simply supported and the clear spans vary from 11.7 to 12.7 m. This is due to a kink in the north abutment. A plan view of the structural layout taken from as-built drawing records is shown in Fig. 5a. Each span consists of 17 longitudinal steel girders. Each longitudinal girder is formed from flange and web plates that are riveted together using angle sections. The typical cross-section of the longitudinal girder with transverse beams is shown in Fig. 5b. The central support is a trestle. The longitudinal girders sit on the trestle alternating between directly over the columns and in the centre of each crossbeam. The crossbeams of the central support trestle have a right-angle notch detail at their connection with the columns, as shown in Fig. 5c.

The structure carries six tracks, one of DLR and five of Network Rail.

Similar to the Warton Road Bridge the alignment is effectively straight in plan and three-car trains are in operation on this line. The abutments are brick abutments founded on spread footings.

Local Monitoring of a Bridge Component (Strain Measurement) Similar to the Warton Bridge, a fatigue management strategy for the Wick Lane Bridge (Fig. 6a) involved detailed site-based works to check, using NDT, for damage at details assessed by calculation as being the most critical for fatigue. Atkins used DIC system (close up lens and virtual gauges) to measure actual strains at the corner of the notch (Fig. 6b) to determine the stresses and compared them to those obtained from current design standards and the FE model.

The DIC monitoring focused on the critical area of the notch with the camera positioned 0.3 m away from the notch detail. The data obtained from the DIC measurement showed strain profiles (Fig. 6c) that appeared

Structural Engineering International Nr. 3/2017

(a)

monitoring, the structure was left untouched and was in full service operation. Overall costs related with the DIC measurement were significantly lower than those associated with the traditional method.

(b)

WICK LANE BRIDGE

The ability to capture a bridge’s behavior with DIC and calibrate a structural model with the collected data provides bridge designers and managers with an easy-to-collect objective measure of bridge performance. Application of the DIC monitoring technique for the overall fatigue management strategy proved to be cost effective. To the authors knowledge this is one of the first examples of the DIC strain measurement performed in outdoor conditions.

DIC camera

Strain 1

Strain [%]

(c)

Strain 1

0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0.02 0

50

100

150

200

250

Time [sec]

DIC is becoming very versatile and cost effective due to the dramatic improvement in the digital cameras and is therefore a promising solution to low-cost structural monitoring of existing infrastructures. A combination of DIC and other advanced monitoring technologies could finally give the information needed to make effective whole life management decisions that keep public assets in service in a cost-effective way.

Fi g. 6: Wick Lane Bridge (a), detail connection and DIC measurement (b), analysis of DIC data (c)

References to be consistent with the loading expected on the bracket along with high strains at the corner of the notch.

Setting up, capturing and packing away was completed within the same day.

The measurement using DIC technology was performed in summer 2015.

Conclusions The DLR structures of Wick Lane and Warton Road Bridges required site-based monitoring to contribute to the assessments of the structures.

Atkins proposed undertaking site investigation using DIC to measure actual strains in the bracket to determine stresses and actual beam deflections. The outcome of the DIC measurements informed the fatigue management strategy. In case of the Warton Road Bridge, the results from the monitoring of the impact damaged deck displacements obtained with DIC, quantified the actual behaviour of the investigated structure, leading to the conclusion that the defects could be tolerated. In case of the

Wick Lane Bridge, processing of the results from the monitoring of the fatigue sensitive detail, showed strain profiles consistent with the loading expected on the bracket together with high strains at the corner of the notch detail. The data supported the general conclusion suggesting that the notch should be inspected for fatigue cracking.

Monitoring of both bridge structures using traditional methods would require construction of an access platform, working at night and at a height with road and rail closure. There were also additional requirements related with completing the monitoring in a timely fashion, avoiding disruption to local stakeholders (adjacent construction site and large shopping complex) and avoiding changes to appearance of structures (survey targets). DIC measurement (setting up, capturing and packing away) was completed within few hours. All of the client’s requirements and concerns regarding closures, stakeholder disruptions, safety risks and structural appearance were avoided. Throughout the

Structural Engineering International Nr. 3/2017

[1] Jiang R, Jauregui DV, White KR. Closerange photogrammetry applications in bridge measurement: literature review. Measurement 2008; 41(8): 823–834.

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Structural Engineering International Nr. 3/2017

Assessment of Structural Steel’s Properties in a Railway Plate Girder Bridge Janusz Hołowaty, Dr., ZUT Szczecin, Faculty of Civil Engineering and Architecture, Poland. Contact: Janusz.Holowaty@zut.edu.pl DOI: 10.2749/101686617X14881937385124

Abstract The material properties of structural steel in a railway bridge constructed in 1872 were determined for the upgrade of the bridge. The riveted superstructure of the bridge consists of six simply supported spans of similar structure but constructed at different periods, with four spans replaced during service. The bridge spans have degraded at different rates, which was clearly observed in the extent of the corrosion. To find the material properties of the steel, laboratory tests were undertaken, with chemical composition, strength, hardness and toughness verified for each span. Surprisingly, despite their similar structure, the spans utilise different types of steel, which not only take into account strength but also toughness. The material testing undertaken allowed some uncertainties regarding the structural steels used in the bridge to be cleared up and the necessity for strengthening the two newest spans to be confirmed. As some upgrading work was expected to be performed through welding, the weldability of the steels was also checked.

Keywords: railway bridge; riveted bridge; structural steel; ageing of steel; durability; assessment; retrofitting.

Introduction The assessment of old riveted bridges requires the verification of their material properties as many types and grades of structural steel have been used.1–3 The majority of old bridges were constructed using mild steels, but other types and grades of structural steel were in use in bridge engineering before the introduction of the first technical requirements at the end of the 19th century. Old bridges Peer-reviewed by international experts and accepted for publication

by SEI Editorial Board Paper received: July 29, 2016 Paper accepted: November 10, 2016

undergo gradual physical and material degradation,4–9 which does not usually affect the steel strength too much; ageing effects are more pronounced for toughness. The long service life of railway bridges also requires the corrosion rates of their members and the integrity of the elements and joints to be checked.

Over the last decade, many railway bridges have been repaired and upgraded in Poland.4 The bridge discussed in this paper was also designated for upgrading and strengthening. The material properties of the bridge’s structural steel were assessed at the design stage as well as at the construction stage. As the bridge was to be upgraded, other items in need of refurbishment were also assessed. The four oldest spans were fabricated from mild steel, and the two newest spans were fabricated from low-alloy steel.

Calculation of the steel grades and corrosion losses allowed the bridge strengthening programme to be modified. An assessment of notch toughness was undertaken to check ageing effects of the steel on the railway plate girder bridge. It was found that the two oldest spans had the highest Charpy index. Some bridge upgrade works were scheduled to be undertaken using welding, so weldability factors for each steel grade were assessed to ensure acceptable welding procedures. In Poland, as in other countries, the use of welding in old bridge structures is prohibited unless the grade of steel and its weldability are verified. The ban originated in the 1950s when the weld quality could have been unsatisfactory. Improvements in welding technology and materials, in addition to qualified welders and experienced inspectors, have made the field of welding reliable, but there are still some uncertainties when welding on old steels.

Small samples of steel were taken for testing from each span to determine their chemical composition, carbon equivalent and other factors in order to assess their material properties.

Structural Engineering International Nr. 3/2017

At the design stage, the samples were cut from the end sections of the plate girders in non-critical zones, so they were small, and the number of specimens for testing was limited.

Hardness measurements were also used to enhance the assessed strength. Rusted upper sections of the steelwork were cleaned with sand prior to the repairs, to estimate levels of deterioration due to corrosion and the replacement of the damaged members. Larger samples of material for additional testing were taken at the retrofitting stage from dismantled steel plates.

Description of the Bridge The double-track railway line where the bridge is located was opened to traffic in 1872. The bridge is a typical structure of that time, with masonry supports and riveted plate girder spans of mild steel. The bridge presently carries the freight railway line from Pozna n to Skandawa (on the Russian border) over a narrow stretch above a water reservoir that was created in the early 1970s. The sub-structure and two spans belong to the line construction times, but four spans were replaced in service: two around 1930 and the other two sometime before 1950. The spans appear to be very similar, but there are differences in their structural details and in their steel types, as was discovered during material testing. The spans are simply supported with lengths of 18.9 m, and the total length of the bridge is 59 m (Fig. 1).

The web height of the plate girders is 1.80 m at mid-span and 1.45 m at the supports. The different types of spans are marked as types 1, 2 and 3. The two-girder spans were supported on sliding triple roller bearings and fixed with knuckle pin bearings (Fig. 2). The bridge timber sleepers were directly supported on the top flanges of the girders and fastened to them by angles; this proved to be one of the main areas of excessive corrosion in the top flanges of the four oldest spans.

Technical Report

441

Side view 18.9

59 m 18.9

18.9

Plan 1Sa,1Sw

1

3

1

2

3Sp2

3

3Sa,3Sp1

2Sa

2

2Nw,2Na

No. 2

No. 1 1Na,1Np

2Np

F ig . 1: South view and plan of the bridge superstructure; types of spans are marked 1, 2 and 3

Cross-sections of the three types of plate girders used in the bridge superstructures are shown in Fig. 3. The oldest spans (type 1 and 2) utilise four flange plates of 10 mm thickness at the mid-span sections, while in the newest spans (type 3), only one flange plate of 20 mm thickness is used. The top plates in the oldest spans were also assembled in two sections, as was recommended in old books and guides. Despite corrosion damage to the top flanges of the oldest spans, crevice corrosion was also observed between the thin flange plates, with excessive plate deformation caused by pack rust near the web stiffeners where one-sided rivets were used.

Referring to the cross-sections and web thickness, the span structures differed in the number of flange plates (1–4) and their thickness, the thickness of the webs (14, 16 and 20 mm) and the

sections of flange angles (L130 × 130 × 16 and L140 × 140 × 13 mm).

The spans also differed in the number of riveted splices and bracing arrangements. Different periods of erecting the components of the span assembly were confirmed by the different corrosion rates in the individual spans, enhanced by a nearby chemical plant and the direct support of the timber sleepers on top of the upper flanges. All four plates from the upper flanges of the two oldest spans (type 1) and the two top plates of the two spans dating back to about 1930 (type 2) suffered excessive pitting corrosion and were identified for replacement. The samples from the dismantled steel plates from the type 2 spans were used in additional material testing. The top gusset plates and a few members of the upper wind bracings were also severely degraded by pitting corrosion and were replaced during repair work.

F ig . 2: View of the support area on the pier with two types of old bearings 442

Technical Report

Programme for Material Identification General guidance for material assessment was given in the requirements of the contract to strengthen the weak spans to the required design loading class (α = 1.21) and to adjust the track structure for freight traffic. The track structure of the bridge was to be replaced by 60E1 rail type, and the centring shims under the sleepers were to be welded to the top flanges of the plate girders, taking into account the vertical alignment of the railway line as well as the structural deformations and camber. Welding of centring shims is a standard procedure for retrofitted bridges on Polish Railways, but it was necessary to verify the steel grade and weldability for each span.

While preparing the contract documents, only a visual inspection of the bridge structure was performed, and severe corrosion loss of the plates under the sleepers was not detected.

The steel was expected to be mild, and it was planned to strengthen type 3 spans (Fig. 3) upon steel grade verification.

To find the material properties of the steel, laboratory tests were planned, where chemical composition, strength, hardness and toughness were verified in each span. Chemical elements influence the strength, ductility and weldability of steel to a large extent. Chemical analysis allows the identification and indicates the quality and consistency in the microstructure.

Verifying phosphorous and sulphur is essential in the assessment of ductility and weldability, while the amount of

Structural Engineering International Nr. 3/2017

1

320

2

3 20

10 10 14

10 10 1800 mm

320×20

160×10

10

160×10

L130

26

L130

26 16

14

L140

26 20

Fi g. 3: Cross-sections of plate girders and upper flanges at mid-span

carbon indicates strength, hardness and weldability. The material samples were planned to be taken from noncritical areas on the plate girders, and the sample was to be relatively small so as not to cause structural damage.

The number and size of the sample did not allow for statistical evaluation.

Despite a similar structure, the bridge spans utilise different steel types, which vary not only in strength but also in toughness. The material testing undertaken allowed some uncertainties regarding the structural steel used in the bridge that had to be cleared up and the necessity for strengthening the two new spans to be verified. As some upgrading work was expected to be conducted by welding, the weldability of the steel also required assessment. Due to the small number and size of the samples for tensile tests, Brinell hardness tests were opted for conducting on each sample.

Toughness was not specified in the contract documents as it is not a necessity in the assessment of old steel. However, toughness was checked for qualitative comparison and for the assessment of ageing. The tests chosen for steel assessment allowed their grades to be confirmed along with the mechanical properties for upgrading purposes, taking into account the experience gained during tests on similar bridges.

Besides the steel, the concrete and brick grades were also to be assessed.

Chemical Analysis Chemical composition of older steels provides guidance on the type of the

steel and its weldability.9 To identify the steel type, the data for carbon, silicon or manganese, phosphorus and sulphur are necessary. Chemical analysis was carried out in the laboratory on 10 samples cut mechanically from the ends of the plate girders (Fig. 1).

The samples were taken from webs (w), top flange plates (p) and top flange angles (a). For these samples, the chemical analysis was performed using a wet method giving the content of the five basic elements (Table 1).

During the retrofit works, another sample was cut out from a dismantled flange plate from type 2 span (2Np), and the chemical analysis was conducted by optical emission spectrometry. For this sample, the five basic elements are also shown in Table 1, while the other elements are Al, Cr, Ni, Mo, B = 0.00%; V = 0.01% and Cu = 0.229%. The values of these elements are average concentrations for four measurements (burns).

The results of the chemical analysis (Table 1) were compared with the chemical composition of wrought and mild steels used in the 19th century, at the time of the first repair (St38-3), and contemporary non-alloy weldable structural steels S235 and S355 (EN 10025). There are large differences in carbon content between the plates and the angles in the spans of types 1 and 2. For the angles, the carbon content is around 0.03%, and in the web and flange plates, the carbon content is 0.11% 0.15% and 0.025% 0.15%, respectively. Low carbon content is a disadvantage as the ageing rate increases when the carbon content is lower than 0.10%. The phosphorus and sulphur (impurities) contents are

Structural Engineering International Nr. 3/2017

generally below 0.05%, a value that is considered harmful. Nitrogen is present in all the mild steels and, commonly found with all impurities, and worsens the steel weldability. Nitrogen was not measured during the initial tests on mild steel due to the low carbon content and impurities, which indicate that the steels are easily weldable. The weldability of this particular steel was assessed by relevant factors, and three of the steel components (carbon, phosphorus and sulphur) in particular must be taken into consideration as they seriously affect the steel weldability.

High nitrogen content may be expected in Thomas steel, which was introduced around 1882, shortly after the bridge was constructed.

The chemical compositions of the different types of steel used in the bridge are given in Table 1; it gives the conclusion that the bridge spans were manufactured from early structural steels of two grades: 1. The oldest spans of types 1 and 2 are from a mild steel (Flußstahl) with a limited content of carbon, manganese and silicon. For the type 1 spans, the phosphorus content is lower than 0.1%, and the manganese content is higher than 0.1%, which indicates a mild steel, not a wrought steel. Silicon is low, indicating rimmed steel. Wrought steel is very seldom, if ever, detected in old existing railway bridges in this region. Mild steel was introduced in 1856 and quickly dominated the rail bridge market2 as there were no problems with phosphorous, as in the original Bessemer steel.

Technical Report

443

Element (%)

Type of span, sample designation* and element (mm) 1

2

3

C

Si

Mn

P

S

1Sw

tw = 14

0.11

0.01

0.59

0.058

0.038

1Np

320 Ă&#x2014; 10

0.15

0.01

0.48

0.038

0.026

1Sa

L130 Ă&#x2014; 16

0.03

0.004

0.39

0.039

0.017

1Na

L130 Ă&#x2014; 16

0.03

0.008

0.30

0.039

0.021

2Nw

tw = 16

0.15

0.01

0.43

0.045

0.034

2Np

320 Ă&#x2014; 10

0.0250

0.0006

0.400

0.0405

0.0203

2Na

L130 Ă&#x2014; 16

0.025

0.006

0.33

0.037

0.021

2Sa

L130 Ă&#x2014; 16

0.03

0.004

0.30

0.041

0.044

3Sp1

320 Ă&#x2014; 20

0.19

0.32

1.23

0.051

0.020

3Sp2

320 Ă&#x2014; 20

0.17

0.30

1.22

0.048

0.018

L140 Ă&#x2014; 13

0.20

0.40

1.27

0.016

0.022

3Sa

Wrought steel

0.018 0.30

0.10 0.33

traces 0.33

0.02 0.46

0.01 0.06

Mild steel

0.03 0.35

traces 0.18

0.04 0.75

0.004 0.16

0.004 0.115

St38-3

0.12 0.18

0.12 0.30

0.40 0.65

max. 0.040

max. 0.040

S235

max. 0.17

â&#x20AC;&#x201C;

max. 1.40

max. 0.025

max. 0.025

S355

max. 0.20

max. 0.55

max. 1.60

max. 0.025

max. 0.025

S, N: the south or north plate girder according to Fig. 1. w: web. p: ďŹ&#x201A;ange plate. a: ďŹ&#x201A;ange angle.

Composition of specimen 2Np by emission spectrometry; the rest of the specimens by wet method. *Designation: 1, 2, 3 â&#x20AC;&#x201C; type of span.

Table 1: Chemical composition of steels

2. The spans of type 3 are made from German low-alloy structural steel, marked by the symbol St52, or Polish steel used in the ďŹ rst half of the 20th century, designated as 6.1.20W52, with properties similar to Polish bridge steel 18G2A, used up to the end of the 20th century and replaced by S355 and S460.

diameter of 10 mm for each of the four samples marked 1Sw, 1Np, 2Nw and 3Sp1. Table 2 shows the mechanical properties measured: fy â&#x20AC;&#x201C; yield strength, fu â&#x20AC;&#x201C; ultimate tensile strength, A5 â&#x20AC;&#x201C; percentage elongation on a ďŹ vefold specimen, Z â&#x20AC;&#x201C; percentage reduction in area and KV â&#x20AC;&#x201C; Charpy index (impact energy). The specimens were in a post-service state (naturally aged). At the time of testing, the steel from the type 1 spans had been ageing for 139 years, steel from the type 2 spans for 87 years and that from the type 3 spans for 62 years. Hardness tests were also carried out on all samples for strength enhancement: fB â&#x20AC;&#x201C; strength according to Brinell hardness tests. Tensile tests were performed using a simple WPM tensile test

Mechanical Properties The samples extracted at the design stage were relatively small with only a limited amount of material available for destructive tests, for example, tensile and toughness tests. Standard tensile tests were conducted on one circular ďŹ vefold specimen with a base Type of span, sample and element (mm) 1 2 3

machine without the possibility for continuous registration of results. The small quantity of material also limited the toughness tests, with only three standard Charpy V-notch impact specimens machined from each sample and tested at a temperature of 20 C. The results are shown as average values in the last column of Table 2.

At ďŹ rst glance, the toughness test results were a surprise because, usually, the older the mild steel is, the lower the toughness is. It may be presumed that the higher toughness values in the oldest steel are due to heat treatment,3 perhaps an early kind of normalising.

The average of the results of Brinell hardness tests are summarised in

Tensile test

Charpy test

(mm)

fy (MPa)

fu (MPa)

A5 (%)

Z (%)

KV(20 ) (J)

1Sw

tw = 14

10

265

389

31

49

53

1Np

320 Ă&#x2014; 10

10

277

396

31

55

50

2Nw

tw = 16

10

257

405

31

52

30

2Np

320 Ă&#x2014; 10

7

254

372

22

69

36

3Sp1

320 Ă&#x2014; 20

10

363

585

30

56

22

tw = 14

10

265

389

31

49

53

1Sw

Table 2: Mechanical properties of steels 444

Technical Report

Structural Engineering International Nr. 3/2017

Type of span, specimen and element (mm) 1

2

3

Tensile test

Brinell hardness test

fy (MPa)

fu (MPa)

HB

fuB (MPa)

fyB (MPa)

1Sw

tw = 14

265

389

111.5

384

276

1Np

320 Ă&#x2014; 10

277

396

113.5

390

281

1Sa

L130 Ă&#x2014; 16

â&#x20AC;&#x201C;

â&#x20AC;&#x201C;

116

396

285

1Na

L130 Ă&#x2014; 16

â&#x20AC;&#x201C;

â&#x20AC;&#x201C;

114

388

279

2Nw

tw = 16

257

405

113.5

390

281

2Np

320 Ă&#x2014; 10

254

372

98

339

244

2Sa

L130 Ă&#x2014; 16

â&#x20AC;&#x201C;

â&#x20AC;&#x201C;

109

378

272

2Na

L130 Ă&#x2014; 16

â&#x20AC;&#x201C;

â&#x20AC;&#x201C;

110

382

275

3Sp1

320 Ă&#x2014; 20

363

585

174

583

420

3Sp2

320 Ă&#x2014; 20

â&#x20AC;&#x201C;

â&#x20AC;&#x201C;

170

569

410

L140 Ă&#x2014; 13

â&#x20AC;&#x201C;

â&#x20AC;&#x201C;

170

569

410

3Sa

Table 3: Tensile properties of steels

Table 3. Hardness was measured in standard conditions using a hardened steel ball with a diameter D = 10 mm as an intender under test weight P = 29 430 N for a period t = 15 s.

The tests were performed by a B3Ctype hardness tester, and average values from three measurements for each sample are given. Taking into account the HB (Brinell hardness) values, the ultimate tensile strengths fuB were assessed according to EN ISO 6596-1, taking the ratio fy/fe = 0.65 0.80. At the same time, the values of fyB were calculated using the average ratio Îą = fyB/fuB = 0.7. The estimated values of fuB and fyB are similar to fu and fy, which were determined in tensile tests on four single specimens (Table 2). The values of fuB and fu are consistent with the literature data for particular steel grades and were used to determine the characteristic strength fyk and the design strength fd.

At the time of the bridge retroďŹ tting, Polish bridge codes were in use for the load-bearing capacity assessment.

The load code is consistent with the international railway union (UIC) recommendations, and bridges are assessed using the UIC 71 rail trafďŹ c model multiplied by Îą = 1.21 (electriďŹ ed line). According to the tests, the minimum values of yield strengths fy = 240 and 360 MPa, respectively, were adopted for the bridge analysis with the material partial factors Îł s = 1.15 for steel of fy â&#x2030;¤ 355 MPa and 1.20 for steel of 355 â&#x2030;¤ fy â&#x2030;¤ 460 MPa according to the Polish bridge standard PN-S-10052:1982 for the design of steel bridges when relevant tests are not undertaken. The material partial

factors should also be increased by 5% for steel types not covered by the code. The calculated design strengths for the steels fd = 200 and 290 MPa are just the same as those given in the code for Polish bridge steels: nonalloy (St3M) and low-alloy (18G2A), respectively. The Polish bridge codes are rather conservative; they use the limit state method with partial factors, but a elastic range for analysis and veriďŹ cation of sections is allowed.

Later, during the retroďŹ tting work, additional material was available from the upper ďŹ&#x201A;ange of the type 2 spans for wider tensile and toughness tests.

The ďŹ&#x201A;ange plate sample (2Np) of a section measuring 320 Ă&#x2014; 10 mm and the specimens are shown in Fig. 4. For the standard tensile tests, circular ďŹ vefold specimens with a base diameter

of 7 mm were used. Four specimens were machined in their actual state, that is, naturally ageing for 87 years, and four specimens were normalised at 930 C for an hour and then air cooled. An attempt to test the normalised steel was undertaken to obtain the steel properties at the time of delivery (construction). After such a heat treatment, the lowest possible steel grain size is obtained, and the microstructure is more uniform.

During the heat treatment process, a homogenous austenite phase is produced. However, it is not known what type of ďŹ nal heat treatment was applied to the steel that was used.

The tensile test curves for the naturally aged specimens are shown in Fig. 5 and for the normalised specimens in Fig. 6. The tests were

Fi g. 4: Sample ďŹ&#x201A;ange plate 2Np (type 2 span) and specimens after toughness and tensile tests

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structural degradation processes. This point of view is conďŹ rmed by the fracture toughness test results (Table 5).

It is presumed that the most susceptible factor for the ductility property transition is fracture toughness. The toughness tests allow the estimation of the transition temperature of the ductile to brittle failure modes. The transition temperature is the temperature at which the minimum impact energy to fracture a standard Charpy V-notch specimen is not lower than 27 J in the case of modern weldable steels. The impact energy KV(T) was measured on two types of specimens: naturally aged and normalised, having a standard size 10 Ă&#x2014; 10 Ă&#x2014; 55 mm. The specimens were tested for a temperature range of â&#x2C6;&#x2019;20 C to +20 C (Fig. 7).

(a) 500

Stress (MPa)

400 300

3124

200 100 0

0

5

10

15

20

25

30

Elongation (%)

(b) 500

Stress (MPa)

400

1

300

3 2

4

200 100 0

0

1

2 Elongation (%)

3

4

F ig . 5: Stressâ&#x20AC;&#x201C;strain curves from tension tests for naturally aged steel (2Np): full range (a) and yield plateau (b)

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(a) 500 400 Stress (MPa)

Over several years of service, degradation changes to the steel microstructures might be observed. According to the literature, the major material property changes are related to an increase in immediate strength, yield strength and hardness and at the same time a worsening of ductile properties.

The worsening of ductile properties is related to a lowering of elongation, necking and toughness and an increase in transition temperature to a brittle state. A comparison level for these changes is supposed to be a simulation of the initial structure of the material through normalised annealing. The test results for the 2Np specimen (see Table 2) indicate a different value for the strengths, with a noted lowering of the ultimate tensile strength fu by 17.5% and the upper limit of the yield strength fy by 19.9%.

The shape of the stressâ&#x20AC;&#x201C;strain curves (Figs. 5 and 6) and the results of the static tensile tests for naturally aged and normalised steels are proof that

they are not evidence for degradation bounds in the ageing material. The tensile curves exhibit a yield phenomenon that, in the case of mild steels, is questioned in some literature.8 The strengths following normalisation are about 20% higher than after service.

These differences may be caused by

300 2

200

4

3 1

100 0

5

0

10

15

20

25

30

Elongation (%)

(b) 500 400 Stress (MPa)

performed on a Zwick/Roell Z600 tensile machine, and the results are summarised in Table 4. All the tensile tests displayed the common characteristics of a ductile metal, that is, elastic region, yielding, plastic region, strain hardening, necking and failure. The curves are typical for mild steels, with the appropriate yield plateaus indicating low impurity contents.

The tested steel demonstrated a relatively low fracture toughness, which is natural for old mild steels. This refers to both types of specimens, aged and normalised. The measured impact energy KV(0 ) is 12.0 J (naturally aged at temperature T = 0 C). The presumed toughness at the delivery stage (at the time of span construction) is also not sufďŹ cient; at temperature 0 C, impact energy is KV(0 ) = 22.7 J. The actual temperature of the ductile to brittle phase transition is T27J = +12 C, and it is much too high to be accepted according to the modern requirements for weldable steels. Brittle fracture

2

1

300 200

3

4

100 0

0

1

2

3

4

Elongation (%)

F i g. 6 : Stressâ&#x20AC;&#x201C;strain curves from tension tests for normalised steel (2Np): full range (a) and yield plateau (b)

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Type of specimen Naturally aged

Number

E (GPa)

fy (MPa)

fyL (MPa)

fu (MPa)

A5 (%)

Z (%)

1

205

274

264

372

25

38

2

212

278

265

377

15*

39

3

193

244

242

376

24

39

4

187

253

245

365

22

46

199

262

254

372

23

40

1

189

357

328

455

28

32

2

202

325

311

448

28

32

3

216

338

316

452

27

31

4

192

325

316

451

29

31

200

336

317

451

28

31

Average value Normalised

Average value

*Specimen fracture at the end of the gauge, one-sided elongation, not included in the average value.

Table 4: Mechanical properties of mild steel (sample 2Np â&#x20AC;&#x201C; type 2 span)

resistance decreases with a reduction in temperature. The measured impact energy at temperature â&#x2C6;&#x2019;20 C for naturally aged steel is only KV (â&#x2C6;&#x2019;20 ) = 5.6 J and for normalised steel KV(â&#x2C6;&#x2019;20 ) = 6.5 J. In the Charpy impact test, the transition temperature is higher when the steel has been subjected to ageing.

However, old steels in riveted girders generally behave satisfactorily over many yearsâ&#x20AC;&#x2122; use. In the upgraded bridge, after dismantling the corroded top plates in the upper ďŹ&#x201A;anges in the four oldest spans (types 1 and 2), the rivet holes were visually investigated for cracks, and no cracks were detected. The results of testing on the old mild steel bridges conďŹ rmed that only partial knowledge exists concerning the material properties, which complicates the assessment of old steel bridges.

Weldability Owing to contractual requirements, the weldability of the steels was assessed as some modernisation works Type of specimen Naturally aged

The mild steel in the types 1 and 2 spans was assessed to be weldable with no restrictions, and for the lowalloy steel, the weldability was a little limited, and preheating was recommended. For the welding works, a welding technology plan was also drawn up, and a trial weld test was advised. The probability of cracking in the shim ďŹ llet welds was assessed to be small as the design stress range in

KV(T) (J) â&#x2C6;&#x2019;20 C

â&#x2C6;&#x2019;10 C

0 C

10 C

20 C

1

5.2

6.4

9.9

23.7

34.9

2

5.7

7.3

11.2

24.1

36.5

3

5.9

7.8

15.1

26.3

41.8

5.6

7.2

12.0

24.7

37.7

1

5.9

10.8

18.5

29.0

55.6

2

6.6

11.6

23.3

31.8

56.1

7.0

16.8

26.3

39.4

59.3

6.5

13.0

22.7

33.4

57.0

3 Average value

content <0.18% (Ito-Basseyo) and for C content >0.18% (IIW). The values of the carbon-equivalent content CEV for the mild steel (types 1 and 2 spans) was 0.05% and for the low-alloy steel (type 3 spans) was 0.41%. The value of 0.41% is lower than the 0.45% for good weldability, but for CEV between 0.40% and 0.60%, some precautions are necessary. The factors determined for hot and cold (hydrogen) cracking resistance and the inďŹ&#x201A;uence of the heat-affected zone (HAZ) are positive. In determining all factors, the elements not assessed in the wet method were taken in the quantities assessed for sample 2Np (see Table 1).

Number

Average value Normalised

were executed by welding, that is, joining the centring steel shims onto the girder top ďŹ&#x201A;anges and new deck steel cantilevers. Centring shims with saddles are used to separate timber and steel to avoid the accumulation of moisture and dirt, along with secondary stresses. Welding the centring shims is standard procedure for retroďŹ tted bridges on Polish Railways, but it was necessary to verify the steel weldability for each span. The basic inďŹ&#x201A;uencing parameter for steel is its metallurgic weldability, determined by chemical content. High carbon results in high hardening properties and eventual loss of ductility. Excess phosphorus increases brittleness, and high sulphur results in porous welds. The greater the alloying content, the worse the weldability. The ďŹ rst proof of weldability9 is the content of phosphorus and sulphur, P + S < 0.10%. For mild steels, manganese is also lower than 1%, while there are only traces of silicon. The low levels of phosphorus and sulphur allowed carbon-equivalent formula CEV (carbon equivalent content) to be devised.

Two formulas were used10: for C

Table 5: Charpy V-notch toughness for mild steel (sample 2Np â&#x20AC;&#x201C; type 2 span)

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[3] Schaper G. Stählerne Brücken, Band I, Teil 1. Verlag von Wilhelm Ernst Sons: Berlin, 1949; 207.

60 Naturally aged Normalised

Charpy impact energy (J)

50

[4] Hołowaty J. Removing weak spots from Polish Railways. Struct. Eng. Int. 2013; 23(1): 85–88.

40

[5] Hołowaty J, Wichtowski B. Properties of steel in a railway bridge constructed in 1887.

Roads and Bridges 2015; 4: 271–283.

30

20

[6] Reid ILK, Milne DM, Craig RE. Steel Bridge Strengthening. Thomas Telford: London, 2001; 141.

10

0 –30

–20

–10

0

+10

+20

[7] Tilly GP, Matthews SJ, Deacon D, De Voy J, Jackson PA, Iron and Steel Bridges: Condition Appraisal and Remedial Treatment.

CIRA: London, 2008; 141.

+30

Temperature (°C)

F ig . 7: Steel toughness and temperature – ageing effect (2Np)

the welds is low, and cracks would not appear.

Conclusions The bridge data reveal that variable mechanical properties may be found in the spans of a single steel bridge. It was necessary to take this into account in static calculations to avoid inadequate strengthening. The analysis of the bridge steels resulted in the confirmation that the plate girders were fabricated from two steel grades.

The oldest four spans were fabricated from mild steel of yield strength fy = 240 MPa, and the newer two spans were fabricated from low-alloy steel of yield strength fy = 360 MPa. The evaluation of the steel grades made it possible to prepare accurate strengthening schemes for the plate girders. The mild steel of span types 1 and 2 is readily weldable, and the low-alloy steel of the span type 3 is conditionally weldable; in the case of welding works, a technological plan is required.

The material testing undertaken allowed clearing of some uncertainties regarding the structural steels used in the bridge and the confirmation on necessity for strengthening the two youngest spans. The degradation of the spans had different intensities depending on the span age. Severe

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[8] Czapli nski K. Old Products from Ferroalloys. DWE: Wrocław, 2009; 100 (in Polish).

corrosion of the upper parts was evident in the oldest spans.

The assessment of notch toughness was a surprise as the two oldest spans have the highest toughness and heat treatment to this steel that might be presumed. The tests on naturally ageing and normalised steel were believed to allow the steel ageing rate to be assessed. However, a lowering of ultimate and yield strengths was found, instead of an increase. It some old bridges, this phenomenon has also been measured.

Low-impact energy, measured on standard specimens, is a characteristic feature for mild steels after many years of operation, and it is not the measure for brittle fracture. Brittle resistance of old steels may be more precisely estimated using fracture mechanics if necessary,11 and fitnessfor-purpose criteria should be developed.

[9] Kühn B, Luki c M, Nussbaumer A, Günter HP, Helmerich R, Herion S, Kolstein MH, Walbridge S, Androic B, Dijkstra O, Bucak Ö. Assessment of Existing Steel Structures: Recommendations for Estimation of Remaining Fatigue Life. ECCS JRC, European Communities: Luxembourg, 2008; 92.

[10] Cremona C, Patron A, Johansson B, Larsson T, Eichler B, Höhler S, Kühn B. Improved assessment methods for static and fatigue resistance of old steel railway bridges.

Background document D4.6, Sustainable Bridges, 2007; 218.

[11] Sedlacek G, Feldman M, Kühn B, Tschickardt D, Höhler S, Müller C, Hensen W, Stranghöner N, Dahl W, Langenberg P, Münstermann S, Brozetti J, Raoul J, Pope R, Bijlaard F. Commentary and worked examples to EN 1993-1-10 “Material toughness and through thickness properties” and other toughness oriented rules in EN 1993. ECCS JRC, European Communities: Luxembourg, 2008; 250.

SEI Data Block Owner: PKP PLK SA, Unit Bydgoszcz Upgrade design: J. Hołowaty, Szczecin

References [1] Hołowaty J, Wichtowski B. Properties of structural steel in early railway bridges. Struct.

Eng. Int. 2013; 23(4): 512–518.

[2] Albrecht R. Richtlinien zum Brückenbau.

Band 1 - Stählerne Brücken einschließlich Stahlträger in Beton und Verbundkonstrutione. Bauverlag: Wiesbaden und Berlin, 1975; 131.

Bridge deck surface (m2): Steel weight (t): Service date

570 150 1872/1930/1950/ 2007

Structural Engineering International Nr. 3/2017

Retrofitting of Three Historic Stone Arch Bridges over the Cetina River in Croatia Jure Radni c, Prof., Dr.; Marija Smilovi c, Dr.; Nikola Grgi c, Dr., Department of Concrete Structures and Bridges, Faculty of Civil Engineering, Architecture and Geodesy, University of Split, Split, Croatia; Frane Borovina, Civil Eng.; Berislav Borovina, Civil Eng.; Ante Mlinar, Civil Eng., Spegra d.o.o., Split, Croatia. Contact: marija.smilovic@gradst.hr DOI: 10.2749/222137917X14881938991320

105 km and flows into the Adriatic Sea near the city of Split. Although flowing through a sparsely populated rural area, many bridges were built across the river. Many of them are historic stone bridges built in the past centuries. The Pavica Bridge, the Panj Bridge and the Balecki Bridge are the largest and the most elegant among them. The bridges are located in a beautiful natural landscape, with a deep river canyon and numerous old water mills. Along the entire course of the river, the water is clean, clear and safe for drinking.

Abstract The basic characteristics of the three historic stone arch bridges over the Cetina River in Croatia are presented first. The bridges are under the protection of the Ministry of Culture of the Republic of Croatia as valuable architectural heritage. The current state of the bridges is then presented, and the main solution of the renovation and strengthening of the bridges is briefly explained. A conclusion is reached that the renovation of the bridges was performed successfully, with excellent cooperation between the engineers and the conservators.

All three bridges were built during the Austro-Hungarian monarchy, in the late 19th and the early 20th centuries, after the occupation of Bosnia.

No design documents based on which they were built were saved, and hence are unavailable. All bridges have similar construction solutions and many details. The bridges are under the protection of the Ministry of Culture of the Republic of Croatia as valuable architectural heritage.

Keywords: historic stone bridges; architectural heritage; deterioration renovation; strengthening.

Introduction The Cetina River is a small mountain river in Croatia, in the province of Dalmatia (Fig. 1). Its length is about

Zagreb CROATIA

Rijeka

Cetina River Split

Balecki Bridge Panj Bridge Pavica Bridge

Adriatic Sea

Dubrovnik

F i g. 1: Cetina River in Croatia with the locations of the studied bridges

Over time the traffic in the area had increased and because of the higher vehicle weight in recent decades the bridges were in a very unfavourable state, with a danger of collapsing.

Therefore the bridges were renovated between 2009 and 2014 by the Road Administration of the Split-Dalmatia

(a)

0.45

10.2

2.5

7.7

2

7.7

2.1

7.7

3.5

7.7

2.2

7.7

2.2

7.7

3.2

5

4.1

0.45

13.2

0.4 4.15-5.10 0.4

(b)

2

5

32

7

2

2

7

7 120

2

7

2

(c)

7

2

7

2

7

20

0.5

7.3

8.8

2.3

8.8

2.3

8.8

4.9

0.5

7.2

Fi g. 2: Side view and cross-section of the bridges. (a) The Pavica Bridge; (b) the Panj Bridge; and (c) the Balecki Bridge (Unit: m)

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County based on previous engineering and conservation projects.1â&#x20AC;&#x201C;6 The original appearance of the bridges is however retained. The internal structure above the stone arches is greatly strengthened, and the bridges now have far greater carrying capacity, safety and durability. This is one of the rare examples of successful restoration of less-known historic stone bridges in Croatia.

Basic Data about the Bridges A side view of the Pavica Bridge and its cross-section is presented in Fig. 2a. The length of the bridge is ~92 m. It has six arched openings of 7.7 m width each and one opening of 5 m width. The top of the arch intrados is about 5.3 m above the river level. The arrow of a larger arc is about 2 m, that is, about one fourth of the arc opening width. The bridge width between the fences is 4.1 m.

The width of the fences is 0.45 m. The arch curves are part of a semicircle.

The arches are formed by regular stone blocks with a width of about 0.3 m. The thickness of the arch in the head and the vertex is 0.8 and 0.6 m, respectively.

Stone blocks are arranged radially, with a narrow mortar joint between the blocks. The bridge was built between 1893 and 1900.

Basic drawings of the Panj Bridge are presented in Fig. 2b. The length of the bridge is ~120 m. The main portion of the bridge over the river has seven arched openings of 7 m width each.

The arch curves are also part of a semicircle. These arches are also formed by regular radially arranged stone blocks with a width of ~0.2 m.

The bridge width between the fences is 4.15 to 5.1 m. The fence width is 0.4 m. The bridge was built between 1880 and 1885.

A side view of the Balecki Bridge and its cross-section is presented in Fig. 2c. The total length of the bridge is ~45 m. It has three arched openings of 8.8 m width each. The top of the arch intrados is ~9.1 m about the river level. The arrow of the arches is about 2.3 m, that is, about one fourth of the arch opening width. The height and width of the radially arranged regular stone blocks are about 0.6 and 0.4 m, respectively. The keystone in the arch vertex is accentuated. The bridge width is 4.9 m. The fence width is 450

Technical Report

0.5 m. The bridge was built between 1907 and 1911.

All stone blocks of the bridges are made of solid limestone, excavated from the nearby quarries. The stones on the sides of the arches and coverings on the fence are ďŹ nely processed.

The stone faces of spandrel walls, intrados of arches and surfaces of columns are processed more roughly.

The most roughly processed stone surfaces are the ones on the abutment. All bridges have side openings in the fence for rainwater drainage from the pavement. The Pavica Bridge and the Panj Bridge have similar solutions and similar quality of construction. The Balecki Bridge has the best quality of construction.

The State of the Bridges Just Before Renovation Images of the bridges just before renovation are shown in Fig. 3. Undermining of the foundations and foundation settlement were not registered. This indicates the high quality of bridge foundation construction.

All the stone arches of the three bridges had very wide longitudinal cracks on the intrados near the sides, which stretched along the entire length of the arch (Fig. 4). The cracks are a consequence of the lateral horizontal forces transmitted by spandrel walls transverse to the arch (Fig. 5). As the arch has almost no compressive stresses in the transverse direction, the

(a)

(b)

(c)

F i g. 3 : Images of the bridges just before renovation. (a) The Pavica Bridge; (b) the Panj Bridge; and (c) the Balecki Bridge

Structural Engineering International Nr. 3/2017

There were damaged stone blocks, cracks and gaps between the blocks on the side walls, the intrados of the arches, piers and abutments. This was probably because of the effects of aging, increased traffic, water leakage, settlements and temperature changes.

Pedestrian traffic over the bridges was difficult at night because there were no lights.

The Pavica Bridge suffered heavy damages during the Second World War. The four-arch assembly on the north side of the bridge and a large part of the fence were destroyed.

After the war, the destroyed parts of the arches, side walls, cornices and fences were rebuilt using concrete.

The reconstruction was performed unprofessionally.

The Panj Bridge was damaged by enemy grenades in 1991 during the war in Croatia. Parts of the arches and pillar in the west and a part of the fence on the east side of the bridge were destroyed. The parts were quickly repaired after the war in order to re-establish traffic. Unfortunately, it was performed unprofessionally.

Destroyed parts of one arch were not repaired. The bridge fence damaged in the Second World War was repaired using concrete.

Spandrel wall

Fi g. 4: Typical damages in the bridges

Embankment

Renovation and Strengthening of the Bridges Conservation Expertise and Repair Project

Displacement Crack

Arch

Displacement Crack

Fi g. 5: Schematic presentation of the causes of longitudinal cracks in the arch (Unit: −)

occurrence of these cracks is logical. In structural terms, separation of the central part of the arch from its sides occurred. Increased transverse horizontal forces in the arch probably occurred because of an increase in traffic over the bridge, poor quality of the embankment above the arch, deteriorated drainage and the effects of ageing.

Cross cracks on the arch intrados and cracks on spandrel walls had also formed. The boundary walls had partly visible transverse displacements.

Because of the soft base and the increased traffic load, pavement surface on the bridges had settled in several places, and free lateral water

drainage was difficult or partially impossible. This resulted in increased seepage of water through the bridge that was visible on the intrados of the arches. Because of the impact of vehicles, fences were partly deformed, damaged and destroyed. The bridges were completely overgrown, which was visually evident by vegetation and trees growing from the bridges (Fig. 4). The trees on the Panj Bridge had diameter up to 0.15 m. Extracted dry stumps testify that the trees were even thicker before. Growth of the trees on the bridges confirmed that the embankment on the bridge had a lot of earth content that had water penetrating into it.

Structural Engineering International Nr. 3/2017

The studies1–3 done by the Ministry of Culture of the Republic of Croatia– Conservation Department in Split, documented the conservation expertise and the current state of the bridges, followed by the defining of the main bridge renovation principles regarding the conservation aspect of valuable architectural heritage. The bridges were required to be restored to their original state in terms of visual appearance, functionality, solutions related to the stone blocks and brickwork, as well as many other details. It was also required that the properties of the new stones; block size, shape and binding; their processing; joints and other solutions regarding the stone masonry be as close as possible to the original.

The bridge reconstruction designer was consulted during the conservation studies. The conservators accepted Technical Report

451

(a)

existing RC

new RC

new RC

(b)

Stone arch Epoxsy glue

Epoxy glue Stone wall

Stone arch

Vertical section

Horizontal section

F ig . 6: Basic principle of the structural strengthening of the bridges and reconstruction of external concrete surfaces. (a) Structural strengthening of the bridges by new reinforced concrete infill (Panj Bridge) and (b) reconstruction of external concrete surfaces of the bridges (a)

the need for strengthening the bridges following an increased traffic load, resistance to strong earthquakes and need for longer durability. The conservators agreed for replacing the embankment above the arches of the bridges with concrete, which would be invisible from the outside and would significantly increase the bridges’ bearing capacity. It was also agreed not to remove the concrete arches and spandrel walls and instead to line them with stone as much as possible.

The need to implement other solutions that were not encountered in the original bridges was also agreed (such as lighting, installation keeping).

(b)

Engineering Design After completion of the conservation studies, an engineering expertise and design project for renovation of the bridges was prepared.4–6

(c)

F ig . 7: The view of the renovated bridges. (a) The Pavica Bridge; (b) the Panj Bridge; and (c) the Balecki Bridge 452

Technical Report

The basic principle of the structural strengthening of the bridges, which complied with the conservation principles, was the removal of the existing embankment above the stone arches.

Instead, a rigid reinforced concrete infill—that will be invisible from the outside—was constructed, thus forming a composite stone–concrete structure with stone arches and stone sidewalls. Structural strengthening of the Panj Bridge is presented in Fig. 6a. The solution enables significant increase in the bearing capacity of the original stone arches, particularly the composite superstructure (about two to four times). Stone masonry of all the arches, spandrel walls, columns and abutments were grouted and all joints between the stone blocks cleaned and re-processed.

Structural Engineering International Nr. 3/2017

Existing concrete on the external surfaces of arches and side walls was removed. It was replaced by stone blocks like those in the original bridges (Fig. 6b). The stone blocks were joined to the existing concrete by epoxy glue and by mortar to each other.

Waterproofing was provided atop the concrete infill. Linear lamps were designed in the vertices of the arch intrados to provide a beautiful appearance to the bridges at night. The stone railing of the bridges was also indirectly illuminated. Bridge renovation designs comprised all implemented solutions and details related to the new reinforced concrete infill, reconstructed concrete arches and spandrel walls and all replaced stone blocks as well as a fence with a cover, lighting, grouting, reconstruction of joints and so on.

All design solutions of the bridge renovation as well as structural calculations were carried out in compliance with the respective Croatian regulations7–10 and professional rules.

Construction First, vegetation was removed from the bridges and the embankment above the arches. Then, external surfaces of the stone blocks were cleaned by water under high pressure. After that, the bridges were strengthened by reinforced concrete infill, followed by the renovation of all damaged stone parts of the bridge, crack and joints grouting, joints preparation and so on.

Finally, the finishing work was performed (waterproofing, lighting, protection of stone surfaces, etc.).

Besides permanent conservation, design and technical supervision were conducted during the renovation of the bridges. It helped the mutually agreed implementation of designdefined solutions as well as the resolution of all unexpected and unforeseen circumstances on the spot. The views of the renovated bridges by day are presented in Fig. 7.

Conclusion Renovation of three historic stone arch bridges across the Cetina River in Croatia was performed successfully, with excellent cooperation between the engineers and the conservators.

The renovated bridges were restored to their original state in terms of their functionality and visual appearance, with increased mechanical strength and safety. This is one of the rare examples of the successful renewal of less-known historic stone bridges in Croatia.

[4] Expertise and design project of renovation of the bridge over Cetina River on Panj, Radnic d.o.o., Split, 2006 (in Croatian) [5] Expertise and design project of renovation of Pavica Bridge over Cetina River, Radnic d.o. o., Split, 2008 (in Croatian) [6] Expertise and design project of renovation of Balecki Bridge over Cetina River, Radnic d.o.o., Split, 2008 (in Croatian) [7] Eurocode 2: Design of Concrete Structures.

Part 2: Concrete Bridges—Design and Detailing Rules (EN 1992-2-2:2005+AC:2008), Croatian Standards Institute, Zagreb, 2015 [8] Eurocode 6: Design of Masonry Structures.

Part 1-1: General Rules for Reinforced and Unreinforced Masonry Structures (EN 1996-11:2005+A1:2012), Croatian Standards Institute, Zagreb, 2012 [9] Eurocode 8: Design of Structures for Earthquake Resistance. Part 2: Bridges (EN 1998-2:2005+AC:2010+A1:2009+A2:2011), Croatian Standards Institute, Zagreb, 2011 [10] Act of the Protection and Preservation of Cultural Heritage Official Gazette 69/99, 151/ 03,157/03 Correction, 87/09, 88/10, 61/11, 25/12, 136/12, 157/13, 152/14, 98/15, Zagreb, 1999

SEI Data Block

References [1] Conservation expertise and repair project of the bridge over Cetina River on Panj, Ministry of Culture of the Republic of Croatia, Conservation Department in Split, 2005 (in Croatian) [2] Conservation expertise and repair project of Pavica Bridge over Cetina River, Ministry of Culture of the Republic of Croatia, Conservation Department in Split, 2007 (in Croatian) [3] Conservation expertise and repair project of Balecki Bridge over Cetina River, Ministry of Culture of the Republic of Croatia, Conservation Department in Split, 2007 (in Croatian)

Structural Engineering International Nr. 3/2017

Owner: Structural design: Contractor: Supervision: Total cost (EUR millions): Service date:

Road Administration Split-Dalmatia County Professor Jure Radni c PhD, Radni c d.o.o., Split Spegra d.o.o., Split, Croatia Ministry of Culture of the Republic of Croatia 2.1

2008–2014

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Superstructure of the National Museum of African American History and Culture Rebecca Laberenne, P.E., Assoc. Dir., 100 Resilient Cities, New York, NY, USA; Guy J. P. Nordenson, P.E., S.E., Partner; Erich Oswald, P.E., Associate Partner, Guy Nordenson and Associates, New York, NY, USA; Ninoslav Krgovic, Architect, Rogers Partners, New York, NY, USA. Contact: rv@nordenson.com DOI: 10.2749/222137917X14881938991366

Abstract The five-story above-grade structure of the new National Museum of African American History and Culture in Washington, DC, USA, features a number of unusual elements, including composite steel and concrete shear walls, a system of cantilevered and perimeter belt trusses, a glass and perforated cast metal façade hung from the top of the building and a longspan free-standing steel entrance canopy, all of which presented both technical analysis and construction challenges for the designers and contractors. The design intent and analysis of these elements is described in this paper.

Keywords: cantilever; composite shear wall; steel truss; plate girder; blast loading; vibration.

Introduction In September 2016, President Barack Obama inaugurated the new Smithsonian National Museum of African American History and Culture (NMAAHC) located on The National Mall in Washington, DC. The design of the new museum, which displays artifacts and documents the history of African Americans in the USA, was selected through an international design competition in 2009. The building is over 35 000 m2, consisting of four below-grade levels and five above-grade levels. This article describes the design of the primary above-grade structural components, including the composite steel and concrete structural cores, the long-span and cantilever steel framing, the hung façade structure (“the Corona”) and the entrance canopy structure (“the Porch”) (Fig. 1).

Above-Grade Building Structural Design The five-story above-grade structure of the museum, including its iconic hung façade system, is supported 454

Technical Report

F i g. 1: NMAAHC under construction (Photo by Richard Barnes)

entirely from four structural cores in order to create a column-free groundfloor lobby. The four structural cores are composite systems comprised of steel floor beams and corner columns, made composite through shear studs with reinforced cast-in-place concrete infill walls. In addition to supporting the entire gravity load of the building, the cores serve as the primary lateral force-resisting system. This uncommon hybrid structural system—which has mainly been used in high-seismic regions—was selected for the cores because compared to steel-braced frames, it provided greater lateral stiffness, which was a requirement for the suspended façade system; allowed for greater flexibility and delay in the coordination of large penetrations through the walls for building services; and provided greater resistance to blast loading and progressive collapse, and compared to reinforced concrete shear walls, it allowed for simpler connections between the cores and the long-span and cantilevered steel framing in the building.

Deep composite plate girders bridge the structural cores at each level to support the main gallery levels. Additionally, each gallery floor plate projects outward of the cores and is supported by cantilevered steel beams on the north side and perimeter vertical trusses and cantilevered brace “outriggers” framing into the composite cores on the east and south sides (Fig. 2).

The above-grade building enclosure, which consists of slender, vertical and horizontal exposed steel trusses supporting an inner layer of glazing and an outer layer of perforated bronzecolored panels, is suspended from the fifth floor of the building by built-up, architecturally exposed steel plate box cantilevers extending from a stiff perimeter comprised of perimeter trusses supported from the outriggers (Fig. 3).

In addition to gravity loading, the above-grade structure was designed for wind, seismic, blast loading and progressive collapse.1 The design of the composite cores was governed by seismic loading rather than wind due

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Fi g. 2: Composite cores, perimeter trusses and floor framing

Fi g. 3: L5 cantilevers for Corona support and Corona with overall superstructure

to the short, stiff character of the building and the imposition of significant seismic loading from the hung façade structure concentrated at the fifth floor of the building.2 The cores also experience significant shear and overturning demands from asymmetric gravity loads resulting from the vertical outrigger trusses on the south and east sides of the building. Additionally, the building was designed to meet stringent antiprogressive collapse requirements set forth by the client.

Design and Analysis of Composite Core Walls Steel columns are located at each of the corners of the above-grade composite structural cores and are supported on reinforced concrete walls below grade. Due to highly concentrated loads where the columns are supported, the corners of the concrete walls below grade were designed with enlarged, high-strength concrete capitals.

According to the US code provisions governing the design of reinforced concrete walls with steel elements, the steel columns in the frame were designed for the full-gravity loads of the building in addition to overturning forces due to lateral loads. The concrete infill walls were designed for

full shear force demands on the walls due to the lateral and gravity loads.3,4 The reinforced concrete infill walls vary in thickness from 254 to 406 mm, depending on the shear force transfer required, and are perforated in numerous locations for architectural openings, mechanical ducts and other utilities. The steel beams in the core were designed for construction loads imposed on them prior to the placement of the concrete infill walls. Once the walls were cast, the beams were assumed to share the loads with the concrete infill walls; therefore, superimposed dead loads and live loads in addition to other loads occurring after the placement of the walls were shared by the walls as bearing walls (Fig. 4).

The concrete infill walls are connected to the steel columns and beams with headed shear studs, which allow shear and axial force transfer between the steel frame and walls. Additionally, in some locations, reinforcing bars are coupled to the steel frame. The boundaries between the steel and concrete where the shear studs are located are more heavily reinforced, with closed ties to provide confinement around the studs.5

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A blast loading analysis was conducted on the core systems to determine out-of-plane pressures on the infill walls and to identify upgrade requirements to steel framing sizes.

The four separate structural models built in ETABS6 and used for the analysis and design of the composite cores in order to capture the complex interaction between the steel and concrete elements are as follows: • Model 1 was a steel-only model with “dummy” diagonal braces used to design the steel core beams and columns for gravity loads and overturning forces due to lateral loads. This model was also used to determine connection design forces for the steel framing.

• Model 2 was a concrete-only model with large openings explicitly used to determine the required thickness and reinforcement for each infill wall and the total number of shear studs required between the concrete and steel elements.

• Model 3 was a composite model with steel frame elements and concrete shell elements used to calculate the sharing of load between these components and to calculate drift.

• Model 4 was similar to Model 3 except that the stud connector elements were modeled explicitly Technical Report

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See note 2 Steel column see framing elevations

25

230

S/2

See wall elevations for concrete infill thickness, reinforcing size and spacing (“S”)

20 clear cover TYP

CL Wall/ steel column S

See note 1 25

See note 1

S/2

#13 crosstie @ 115 OC (alternate hook) #19 U bar @ 115 OC (except where wall reinforcing is #13 use #13)

S

See note 2

230

See note 1 See note 1

19 diameter 150 long headed shear stud typical

Infill wall connection to steel column JI

Plan section

S567

F i g. 5: L2 platform with perimeter vertical steel trusses (Photo by Richard Barnes)

Notes 1. Vertical bars in vertical confining reinforcing shall match bar size used in corresponding wall 2. Provide Class B tension lap splice between horizontal bars in wall and confining U-bars. See typical details

Design and Analysis of Long-Span and Cantilevered Floors U bars @ 115 see section JI Develop horizontal confining reinforcing above and below beam to steel column with dowels coupled to column

Construction joint at top of floor slab (clean and roughen) 190

Top/slab elevation varies Top/steel elevation varies

# 19 dowels @ 115 couple to top of steel beam

Provide extra pair horizontal dowels coupled to beam between horizontal confining reinforcing

Steel beam and column see framing elevations Develop vertical confining reinforcing to steel beam above and below with dowels coupled to beam

U bars @ 115 below beam

Provide extra pair horizontal dowels coupled to beam between horizontal confining reinforcing Infill wall corner reinforcing FI S567

Detail Notes 1. Wall field reinforcing not shown for clarity 2. Detail is similar where wall frames into steel column web

F ig . 4: Composite core wall sections [Units: mm]

as nonlinear elements, an approach that was developed for this structural system to better represent the behavior of and interaction between the steel and concrete wall components.

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The engineer responsible for the erection of the structure also developed a separate model representing the sequential stages of construction of the cores to ensure that design loads were not exceeded during construction.

Composite steel plate girders that are 1120 mm deep span 20 200 mm between the four cores of the building and on most levels support a 190 mm deep composite floor slab made of 115 mm normal weight concrete on a 75 mm thick deck. The relatively thick floor slabs provide better vibration control and were more easily designed to resist uplift due to blast loading as well as in-plane shear loads due to the lateral loads imposed at the fifth floor from the hung façade mass.

On the north side of the building on Floors 2, 3 and 4, the floor plate extends beyond the northern edge of the supporting cores, requiring deep cantilevered plate girders to support the escalator system. On the east and south sides of the building, at these same floor levels, the projection of the floor plate is accommodated by a system of vertical steel trusses that wrap around the perimeter of the galleries, concealed in the gallery walls, which are supported by vertical steel brace “outriggers” framing into the southern and eastern faces of the cores. The vertical trusses are most pronounced between Floors 2 and 3 in order to allow the lower portion of the structure to function as a stiff platform for the erection of the upper levels. The erection of the steel structure, however, required shoring in order to limit floor deflections and weight from

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upper slabs to level the floors in the galleries (Fig. 5).

• Floor vibration analysis7 to ensure that comfort levels would be met in the galleries despite the long spans • Detailed finite element analysis of the architecturally exposed fabricated steel plate box cantilevers supporting the hung façade trusses at Floor 5, including consideration of fatigue stresses • Consideration of construction sequencing of structural cores versus trusses to understand forces imposed in bracing elements and resulting deflections

Floor 5 has similar floor plate extensions to the north and west, which are also accommodated by floor-deep vertical trusses cantilevered from the structural cores between Floor 5 and the roof.

The design and analysis of the steel framing and trusses bridging between, and cantilevering from, the four structural cores included the following considerations: • Accommodation of significant, oversized web penetrations through the deep composite plate girders • Lateral-torsional buckling analysis of the deep composite plate girders due to net uplift in some locations from blast loading analysis

3780 EQ K5 S588

K1 S588

6.1

See note 4

EQ

W1000 beam see plan Flange splice plate (HS-SC bolts) see note 5 Stiffener plate below TYP

C B L ox rd er gi

CL Beam/ column below EQ

K10 S588

Fabricated plate girder see plan

WP

WP

3 sides

Fit box girder side plates to plate girder web Stiffener plate TYP

CJP

CJP

A

(1554 east side)

1539

19 mm thick stiffener fitted to plate girder both sides

PG9 and PG9a only

CL x Bo der r gi

B.1 2.1

TYP @ east UON

CL SYM ABT

CL Beam/ column below

See note 4 TYP @ north and west

The iconic design feature of the museum is its delicate cladding system

Early on in the design of the project, it was determined that suspending the structure supporting the glazing and cladding systems from the top of the museum rather than from the base would be advantageous in limiting the member sizes of the structural elements as gravity loading would effectively pre-tension these elements against compression loads induced by wind and other transient loads. This

A

F10 S587

12.7 mm thick stiffener plate fitted to box 4 sides

F10 S587

Tapered fabricated plate box girder

1278

East adjacent to gridline C.O only

F5 S588

Structural Design of the “Corona” Building Enclosure

with a sawtooth profile, resembling an African carved wood crown, or Corona. Unlike conventional building enclosures, which are typically attached to each floor level of a building, the Corona frames into the building only at the base of the museum and at Floor 5, thereby creating a continuous atrium on all four sides of the museum.

R = 50

C L Monorail support CJP F10 S587

WP 633

12.7 mm thick welded stiffener plate fitted to box 4 sides

932

76.2 mm thick end plate (91.2 mm east side) see note 3

C L Corona attachment H1 S587

Corona truss attachment work point see note 2

W360 monorail support beam see plan Provide slotted shear connection on one end of monorail support beam to release axial forces

Plan detail Notes: 1. Columns below on gridline 6.1 only not shown for clarity 2. Corona truss attachment workpoint corresponds to centroid of applied vertical and lateral loads from corona truss.

Corona structure not in contract 3. See detail by Northstar/Enclos for bolted connection detail. Tack weld nuts to match bolt type and configuration on/inside of end plate.

Access hole may be provided at top of beam if approved by engineer and closed prior to full loading 4. See plan for extent of fabricated plate girder and end shear splice 5. HS-SC bolts shall be designed for slip as a serviceability limit state per AISC to control deflections of Corona structure

Fi g. 6: Typical Corona truss attachment plan details [Units: mm]

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850

Corona truss attachment work point see note 1

76.2 mm thick end plate TYP see note 3

CL x Bo der r gi

12.7 mm thick stiffener 19 mm thick stiffener

W360 monorail support beam see plan

450

925

CJP R=200

CL Beam

WP

185

WP

850

D5 S588

185

1420

Corona truss attachment work point see note 1

152 MIN TYP

0

1278

F10 S587

40

Flange splice plate (HS–SC bolts) see note 2

B

76.2 mm thick end plate TYP (91.2 mm east side) see note 3

300

CJP WP

389

CJP

CL Monorail support CL Splice

932

(404 east side)

CL Corona attachment

Connection forces for design of the splices are to be coordinated with the requirements of the monorail manufacturer and submitted to engineer for review

932

389

300

CL Corona attachment

R = 200

925

R = 100 Top and bottom connection plate 50mm

1420

Plate girder see plan

Web and stiffener plates 19 mm thick UON

CL Splice 450 Flange splice plate (HS–SC bolts) see note 2

F10 S588

Plate girder see plan CL Beam A1 S587

1278 (1293 east side)

CL Monorail support

Plan detail Notes: 1. Corona truss attachment workpoint corresponds to centroid of applied vertical and lateral loads from Corona truss.

Corona structure not in contract 2. HS–SC bolts shall be designed for slip as a serviceability limit state per AISC to control deflections of Corona structure 3. See detail by Northstar/Enclos for bolted connection detail. Tack weld nuts to match bolt type and configuration on/inside of end plate. Access hole may be provided at top of beam if approved by engineer and closed prior to full loading

F ig . 6 Continued

decision led to numerous complexities and considerations in the structural design of the building itself, including strict drift requirements for the lateral system of the building, complex base connections requiring fixity and freedom of movement along different axes and consideration of unconventional load paths for wind and other transient loads, such as blast and seismic loading, acting on the façade system.

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The schematic design of the Corona structure included a series of hung horizontal trusses supported by vertical cables and in-plane X-bracing cables. All gravity loads were resisted in tension by the vertical cables, and lateral loads such as wind loading on each face were transferred laterally by the horizontal trusses to the orthogonal faces of the structure, which, in turn, functioned as inverted braced frames to transfer these lateral forces

in-plane to the fifth floor of the building. As a result of value-engineering and construction sequencing considerations, the three-dimensional structural system was replaced by hung vertical trusses comprised of architecturally exposed hollow-structural steel sections (HSS), spanning 28 140 mm from the base of the building to Floor 5 in the final design. The glazing provided in the building enclosure is located on the inner face of the

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vertical trusses, and perforated bronze-colored cast aluminum panels clad the outer faces. Service catwalks are located in the cavities between these surfaces.

Wind tunnel testing was conducted for the Corona system to determine wind loading requirements on both the glazing and the cladding. Additionally, special analysis was conducted to determine ice loading requirements, which were significant, on the exposed steel and cladding panels (Fig. 6).

Structural Design of the “Porch” Canopy Visitors who enter the museum lobby through the south entrance pass below a massive canopy structure referred to as the “Porch”, which is independent from the rest of the above-grade building structure and supports an unoccupied green roof. The structure consists of a 2800 by 1872 mm steel box-girder free-spanning 52 490 mm (52.5 m) between two built-up steel plate box columns supported by concrete columns in the loading dock below grade. Tapered fabricated steel plate beams cantilever to the south from the box-girder, supporting the green roof surface and creating a triangular or wing-shaped cross-section to the canopy. As a result, gravity loading on the Porch roof as well as wind and

Fi g . 7: Porch canopy structure under construction

blast loading induce significant torsional forces on the primary box-girder.

The entire steel structure is clad in ultra-high performance precast concrete panels (Fig. 7).

Conclusion The design of the above-grade structure of the new NMAAHC required a number of unconventional approaches to accommodate certain architectural features. To provide a column-free lobby, the entire building is supported by four vertical cores with bridging and cantilevering steel trusses supporting the upper gallery levels of the building. To allow for the sawtooth profile and free-spanning nature of the

Corona façade, the support structure for the building enclosure is hung from the upper floors of the building.

To accommodate this hung façade as well as blast and other building service coordination requirements, the structural core walls of the building are a composite system of steel framing with reinforced concrete infill walls (Fig. 8).

References [1] GSA Progressive Collapse Analysis and Design Guidelines for New Federal Office Buildings and Major Modernization Projects, US General Services Administration, Washington DC: June 2003.

Fi g. 8: NMAAHC integrated Revit model section: perspective view

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[2] American Society of Civil Engineers.

ASCE/SEI 7-05, Minimum Design Loads for Buildings and Other Structures American Society of Civil Engineers: Reston, 2005.

[6] ETABS structural analysis software by Computers and Structures International Version v9.7.2, Computers and Structures Inc, Berkeley

[3] American Concrete Institute. ACI 318-08, Building Code Requirements for Structural Concrete American Concrete Institute: Farmington Hills, 2008.

[7] AISC. AISC Design Guide 11: Vibrations of Steel-Framed Structural Systems Due to Human Activity, 1st edn. AISC: Chicago, 2003.

[4] American Institute of Steel Construction. ANSI/AISC 341-10, Seismic Provisions for Structural Steel Buildings American Institute of Steel Construction: Chicago, 2010.

[5] American Institute of Steel Construction.

AISC 360-05, SpeciďŹ cation for Structural Steel Buildings American Institute of Steel Construction: Chicago, 2005.

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SEI Data Block Owner

Smithsonian Institution, National Museum of African American History and Culture

Architect Structural engineers

Contractor Steel (t) Concrete (m3) Estimated cost (USD million) Service date

Freelon Adjaye Bond/SmithGroup Guy Nordenson and Associates (Superstructure) and Silman (Substructure) Clark/Smoot/ Russell 3800 55 050 500 September 2016

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engineering connections®

Not the first reader? Card taken?

@ September 21–23

Come and meet our engineers at the IABSE Symposium, at booth 38 or at one of our presentations.

Structural bearings Expansion joints

Seismic protection Structural monitoring

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IABSE News IABSE Awards 2017 The International Association for Bridge and Structural Engineering (IABSE) is pleased to announce the IABSE Awards 2017. Prof. Fernando Branco, President of IABSE, will present the Awards on September 21, 2017, on occasion of the 39th IABSE Symposium in Vancouver (Sept. 21–23, 2017).

Honorary Membership

Prof. David A. Nethercot, UK

Honorary Membership is presented to Professor David A. Nethercot “in high appreciation of his outstanding and dedicated services to the Association”.

David Nethercot was for 12 years prior to his retirement in 2011, Head of the Department of Civil and Environmental Engineering at Imperial College London; previously he was on the staff at Cardiff, Sheffield and Nottingham Universities, including 5 years as Department Head at Nottingham.

He was awarded a DSc degree in 1993 and elected to the Royal Academy of Engineering in the same year.

David is a Past President of the Institution of Structural Engineers and a former Council Member of the Royal Academy of Engineering. In 2006 he was awarded an OBE for services to structural engineering. He received the IStructE’s Gold Medal in 2009 and was elected a Foreign Fellow of the Academy of Technical Sciences and Engineering of Australia in the following year and a Foreign Member of the National Academy of Engineering of the USA, 5 years later.

David has 50 years’ experience of teaching, research and specialist advisory 462

IABSE News

work on major projects, principally in the field of steel construction. He is the author of more than 400 research papers as well as textbooks and design guides and has addressed conferences in some 50 countries worldwide.

David Nethercot has been one of the pillars of the worldwide structural engineering community for years, including within IABSE where his leadership and example has been an inspiration to many.

David joined IABSE in 1986. He has been a member and Chair of WC 2 (1989–1999), a member of the Outstanding Paper Committee (1991–1992), a member of the Chair of the Technical Committee (1995–2005). He was the Chair of the British Group for many years (1997–2013) and has been a Vice President and member of Admin Committee and Executive Committee (2005–2013). He has served as a member on several Scientific Committees of IABSE conferences and was the Chair of the Scientific Committee for the IABSE-IASS London Symposium in 2011. During his term as President (2013–2016) he was actively involved in the implementation of the new IABSE Strategy for the rejuvenisation of the Association and saw the successful organisation of four IABSE Conferences.

He is currently a member of the Strategy Group, a member of the SEI Advisory Board and a Delegate to the Permanent Committee.

International Award of Merit IABSE presents the International Award of Merit in Structural Engineering, Professor Juan José Arenas, Spain, “in recognition of his outstanding contributions for more than 50 years to structural engineering including several span world records and outstanding structures, and his sensitivity to bring structural engineering design close to art and poetry”.

Juan José Arenas is a renowned Spanish bridge engineer, born in Huesca in 1940. He graduated from the Civil Engineering Faculty (ETSI Caminos,

Prof. Juan José Arenas, Spain

Canales y Puertos) in Madrid in 1963, and obtained his Doctorate in 1970 from the same Polytechnic University (UPM). Between 1970 and 1976 he was appointed Professor of Prestressed Concrete at ETS Ingenieros de Caminos de Madrid. In 1976, he became Chair Professor of Bridge Design in Santander’s Engineering Faculty (University of Cantabria).

After his graduation, Arenas started his professional career as an independent bridge designer in Madrid from 1969 to 1978. He then moved to Santander and co-founded the engineering firm APIA XXI in 1988.

During that period, Arenas’ most relevant projects included La Barqueta steel arch Bridge in Seville, La Arena cable-stayed Viaduct, Hispanoamérica cable-stayed Bridge in Valladolid, La Regenta concrete arch Bridge and Barcelona’s Harbor Movable Bridge.

In 1999, Arenas founded the engineering firm Arenas & Asociados, which continues his bridge design work until today. Some of the noted bridges that Mr. Arenas has designed as President of Arenas & Asociados include the Third Millennium Bridge in Zaragoza, Almonte HSR Viaduct, Las Rozas cable-stayed bridge in Madrid, Chilina Bridge in Peru, and the future Ponte dei Congressi in Rome.

During his career, Arenas won many prestigious awards in recognition of his design work. These include the 1993 ECCS European Award for Steel Structures; 1999 VI Alcántara Bridge Award; 2009 Gustave Magnel Gold Medal, 2010 fib Award for

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IABSE News 2008. He is a Fellow of both the Institution of Civil Engineers (FICE) and the Institution of Structural Engineers (FIStructE), and is highly sought after as an expert consultant, and chairs and serves on many UK, European and International technical committees. In particular, he is the UK representative for the development of Eurocode 3, and he chairs the UK Mirror Group.

Outstanding Paper Award

Barqueta bridge, Spain

Outstanding Concrete Structure and 2011 1st ACHE Engineering Prize.

He is member of IABSE, fib, ACHE, IASS, APTA, AITIM, etc., and has published many technical papers and delivered numerous lectures at international conferences.

IABSE Prize

Prof. Leroy Gardner, UK

The IABSE Prize honours an IABSE Member early in his or her career for outstanding achievements in structural engineering. IABSE has awarded the IABSE Prize 2017 to Professor Leroy Gardner “in recognition of his outstanding contributions to structural engineering in the areas of structural steel research, code development and teaching”.

Following graduation from the University of Southampton in 1998 with a first class degree in Civil Engineering and as the top ranked student, Leroy went on to study at Imperial College London for an MSc (Distinction) and PhD. Since then, he has followed an

academic career from Lecturer to Professor at Imperial, while retaining close links with practice through consultancy work, training courses and the Institutions. He is the youngest ever Professor in the Department of Civil and Environmental Engineering and having spent 5 years as Director of Undergraduate Studies, now serves as Director of Research.

His principal research interests lie in the areas of structural testing numerical modelling and the development of design guidance for steel structures. In particular, through his involvement in European and other research projects, he has devised methods for designing with non-linear materials and has contributed to the formulation of design rules for Eurocode 3. His research now features in UK, US, European and Chinese codes of practice.

Leroy leads an active research group in structural engineering, and has supervised over 30 PhD students. He is an inspiration to his students, 15 of whom have gone on to pursue academic careers themselves, and in 2015, he was awarded the President’s Prize for Excellence in Research Supervision at Imperial. Leroy has coauthored 4 textbooks, 7 book chapters and over 250 technical papers. He also has won a number of research prizes, has given invited and keynote lectures around the world, and is Editor-in-Chief of two international structural engineering journals.

Leroy was awarded an Imperial College Excellence in Teaching Prize in

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IABSE presents the Outstanding Paper Award to the author(s) of a paper published in the preceding year’s issues of the IABSE quarterly Journal Structural Engineering International (SEI), encouraging and rewarding contributions of the highest quality. IABSE presents the award in the categories ‘Scientific Paper’ and ‘Technical Report’.

The Outstanding Paper Award in the category Scientific Paper is presented to: “Post-Tensioned Glass Beams for a 9 m Spannglass Bridge” by Michael Engelmann, Bernhard Weller, Germany, SEI 2/2016 (May), pp. 103–113.

Glass beams may be reinforced to allow for a fail-safe behaviour and to avoid brittle and total collapse. This approach from reinforced concrete design is applied to glass and extended by post-tensioning the reinforcement by using Spannglass beams.

A high-grade steel cable relieves the brittle material from unbeneficial tensile stress and accommodates permanent dead-load deflection. A structural concept was evolved that led to the development of Spannglass beams, which were tested for a span of 2 m. This concept was extended to larger spans in order to consider size effects and to provide this option to a broader field of application. To verify this, a 9 m pedestrian glass bridge with post-tensioned reinforcement was designed and tested experimentally. Three full-size glass specimens were utilised with adhesive connectors and two reinforcement designs. A post-tensioning concept was finalised and put into practice. Finally, fourpoint bending tests were performed to yield the maximum loads. This paper explains the experimental design, presents the results and provides conclusions of how to design workable IABSE News

463

IABSE News structural glass beams with posttensioned reinforcement. The paper also presents the design process and the construction of a 9 m Spannglass Bridge used by pedestrians at Glasstec 2014 in Düsseldorf, Germany.

The Outstanding Paper Award in the category Technical Report is presented to: “Construction of a Large Composite Gridshell Structure: A Lightweight Structure Made with Pultruded Glass Fibre Reinforced Polymer Tubes” by Lionel du Peloux; Frédéric Tayeb; Olivier Baverel, and Jean-François Caron, France, SEI 2/2016, (May) pp. 160–167.

The Ephemeral Cathedral of Créteil at Paris, France, is a gridshell structure made of composite materials. Built in 2013, this 350 m2 religious edifice is a temporary church meant to gather the parishioners during the 2 year renovation of their permanent cathedral. This large-scale prototype represents a first in the building industry which still shows excessive apprehension for the use of non-traditional materials such as composites, especially when it comes to structural applications.

After a brief description of the structure, this paper gives an insight into the field of elastic gridshells and discusses the benefits of composite materials in such structures. The overall design process employed, from the three dimensional (3D) shape designed by the architect to a full

Phoenix Centre, Beijing, China 464

IABSE News

structural shell, is then explained.

Finally, complexities due to the connections are highlighted and solutions are presented.

Outstanding Structure Award The Winner IABSE awards the Phoenix Centre, Beijing, China, with the Outstanding Structure Award 2017. This Award recognises the most remarkable, innovative, creative or otherwise stimulating structure completed within the last few years.

“Phoenix Centre represents a complex spatial structure originated from 3D parametric design techniques resulting in an extraordinary interaction of architectural and structural aspects”.

The Phoenix Centre is located next to the southwest corner of Chaoyang Park in Beijing. The site is 18,000 m2, with a construction floor area of about 72,000 m2, and a building height of 54 m. The Phoenix TV station proposed to build a media centre in Beijing in 2007. The building was to represent not only the corporation spirit of Phoenix TV but also the profound cultural tradition of China. The image of perpetual motion of “the Mobius strip” matches the spiritual pursuit of Phoenix for wisdom and ideal. It also resembles the Phoenix bird in the structural geometry, and rhythm balances the Yin and the

Yang. The theory of Yin-Yang is a perspective and methodology originated in ancient China. The shaded (Yin) part as production studio and the sunny (Yang) part as office building are wrapped up by the ecology Mobius shell. Several spatial public spaces are created to demonstrate the cultural value of the Phoenix Media, which makes the entire building vibrant and energetic. The building shape is in harmony with the surrounding infrastructure including the irregular shaped site, the varied intersection angles between the urban roads and the scenic environment.

Using accurate 3D digital modeling techniques, the structures, especially the curtain wall supporting system, create enough spaces for different architectural functions, and also contribute to the aesthetic effects of the whole building. The overall design not only makes the entire building more vibrant and energetic, but also allows the public to have a closer experience with the Phoenix culture.

The building successfully seeks the interaction and integration of structure, architecture, landscape and art.

It achieves the extraordinary combination of architecture and structure.

The Finalists In addition to the Winner, the Outstanding Structure Award Committee, chaired by Paul Lüchinger, Switzerland, also selected three Finalists: New Street Station, Birmingham, UK. The transformation of Birmingham New Street Railway Station fulfilled the most ambitious of architectural visions for a tired 1960’s RC building.

The existing building could not be proven to satisfy any current codes of practice. Therefore a profound, firstprinciples understanding of the building’s structural behaviour had to be developed, which was central to modelling the significant changes in the forces that would be applied during construction and when complete.

Engineering design and constructability remained entirely in synergy to safely accomplish the works, with paramount consideration given to public safety as this principal UK transportation hub remained operational throughout the works.

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IABSE News from its base to its top. The steps rings form a cylindrical space which is filled with reinforced concrete, providing a stiffening core to the structure.

New Street Station, Birmingham, UK (photo: Network Rail)

The project aspirations were delivered through a complete remodelling of the old station layout, 6000 t demolition of two floors from the old shopping centre, 7000 t of demolition from the mezzanine floor of an old car park and a similar amount from the demolition of the 61 m tall Stephenson Tower to the south of the site. A new atrium, formed within the old structure has allowed natural light to enter the station for the first time since the 1960s, and central shopping facility in the city centre of Birmingham. The works included extensions to the building amounting to about 27,900 m2 of floor space, accommodating a new stainless steel façade on the old building.

Port of Valencia Lighthouse, Spain, is the first structure ever built in the world using all composite materials. It is an innovative structure using new technologies especially in regards of the highly aggressive maritime environments and the difficult accessibility.

This structure demonstrates how composites can be used intelligently and effectively.

Port of Valencia Lighthouse, Spain

Dandeung Bridge, Gunsan-si, South Korea

The new lighthouse is a 32 m tall, five storey structure. The floors, placed at every 6 m, are 200 mm thick glass fibre and polyurethane sandwich panels with an octagonal shape. Passing through holes near the vertices of these octagonal panels, and blocked by specially designed glass fibre cones, eight carbon fibre columns with a 250 mm diameter support the structure. A spiral staircase with 130 steps made of composite materials is placed in the centre of the structure, going

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Dandeung Bridge, Gunsan-si, South Korea, is a suspension bridge supported by a sail shape “D” design pylon, connecting Sinsi Island and Munyeo Island in South Korea and has been constructed in the second section of connecting road works of the Gogunsan islands in Jeollabukdo (2009–2015).

The bridge consists of the only one steel box-girder for two traffic lanes and two pedestrian lanes with a 400 m span and a single concrete pylon with the height of 105 m. As this bridge has been designed as an earth anchored suspension bridge, two main cables with a diameter of 0.38 m are anchored within the concrete anchorages at both ends of the bridge. Innovative structural systems were introduced to realise the single pylon suspension. To overcome the difficulties, e.g., the atypical shape of the pylon, the spatially arranged main cables, and new type of lifting gantry could come into the world. It is expected that the techniques developed for this new bridge will contribute to the creation of more advanced construction techniques in the future.

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IABSE News IABSE Conference Kuala Lumpur 2018 ‘Engineering the Developing World’ Technical Sessions: these sessions will provide the platform to present developments from an academic perspective within the context of the global theme. The sessions will start with a number of short presentations on the specified topic followed by an open floor discussion for the room to participate in.

Themes and Topics Large-Scale Infrastructure Project: Regional and international bridge, tunneling and structural projects are becoming ever more ambitious, especially in the developing world, where they become showcases for nascent economies.

Night view of Kuala Lumpur with the Petronas Towers

Introduction Preparation for the IABSE Conference Kuala Lumpur 2018 ‘Engineering the Developing World’ is well under way.

The Scientific Committee, chaired by David Collings is now reviewing the abstracts received and preparing the programme with the support of the Organising Committee, chaired by Paul Mullins. The Conference will be held from April 25–27, 2018, and submission of abstracts is still open.

The challenges facing engineering practices in the developing world is the core theme of the 2018 Spring Conference. The huge growth in urban development fuelled by dramatic economic development and fundamental changes to socio-economic structures and norms across the developing world has left society and the required infrastructures with enormous challenges. As structural and civil engineers, we are at the forefront of overcoming these obstacles and shaping the world that will evolve over the next century.

Once a muddy confluence thriving on tin mining activities, Kuala Lumpur has risen into a global city and 466

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beyond. Consistently ranked among the top developing cities in the world, this is a modern city built up on a cross-cultural assimilation of national historical heritage values, coupled with the best engineering endeavours.

Driven by the will to transform into a developed nation, Kuala Lumpur continues to stamp its mark as a model city amongst the developing nations and the rest of the world, and is a perfect venue for what promises to be a stimulating, informative conference.

Practicing engineers and researchers, architects, consultants, design software engineers, contractors, designers, students, educators, owners, suppliers and government agencies are invited to join us in Kuala Lumpur to share their visions for the future!

Technical Programme Scientific Sessions: these sessions will provide the platform to present new research results, technical developments, specific project presentations, lessons learnt and future expectations.

Each session will be based on accepted papers of 6–8 pages, which will then be presented orally in 10 minute slots.

Engineering Innovations for Rapid Urbanisation: As the world becomes increasingly urbanised, technical innovation is paramount where urban constraints continually challenge designers.

Towards Achieving a Low Carbon Footprint: The importance of achieving a low carbon footprint is already recognised in developed economies, however, are the same considerations being given in the developing world?

Harmonising New and Old: The integration of new with old is a challenge that requires sensitive consideration, especially in developing nations where often the gulf between the two can almost seem eternal.

The Future of Aesthetic Design: Is there a future for aesthetics in design where cost increasingly becomes the only driving factor? This is an important question in rapidly developing economies, where it is often left behind.

Submission of Abstracts is still open: interested authors willing to present their paper are invited to submit their abstract to above themes and topics under following link: https://mc. manuscriptcentral.com/kl2018.

Authors who do not have the opportunity to present a paper at this year’s IABSE Symposium in Vancouver, are encouraged to submit their abstract too, an expansion on themes and

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IABSE News topics will be considered by the Organiser if necessary.

fellow delegates, speakers and the Organising Committee.

Young Engineers Programme

The Cocktail Reception – Poolside Double Tree at Hilton

Any author of a paper published at the Conference who is born after 1984 and who personally presents his/her paper at the Conference, will be considered for the Young Engineers Outstanding Contribution Award. An exclusive Young Engineers networking reception will be held on Tuesday, April 24, 2018 immediately prior to the Ice Breaker Party.

On Wednesday April 25, 2018 19:00, all delegates will be officially welcomed to Kuala Lumpur and Malaysia.

Venue The five-star Double Tree at Hilton will host the event. It is located in the heart of the city with shopping malls, dining, entertainment and nightlife, and the world famous Petronas Twin Towers close by. It is conveniently accessible to and from the international airport and adjacent to the Ampang Park Light Rail Transit.

Accommodation Kuala Lumpur offers a wide variety of very affordable accommodation to meet any budget. Five-star luxury to one star can all be conveniently booked in close proximity to the Conference venue. On-site the fivestar Double Tree at Hilton housing the conference facilities, offers superior rooms with breakfast inclusive of all local and government tax at Ringgit Malaysia 400 (approx. €80) a night.

Technical Tours Technical tours to the following sites will be offered on a first-come firstserved basis.

FRIM canopy walk

Accompanying Persons Merdeka PNB 118 to be completed in 2019

KVLRT Line 3 Casting Yard Thursday April 26, 2018, 13:30–17:00 The proposed alignment is 37.7 km, which comprises 26 stations, 35 km U-shape viaduct and 2 km underground section, will run between the district of Bandar Utama and Johan Setia to the west of Kuala-Lumpur.

The full span precast U-shape viaduct system was first used here in Malaysia with standard span length of 30 m.

Merdeka PNB 118 Friday April 27, 2018, 08:30–12:00 Merdeka PNB 118, formerly known as KL118 or Warisan Merdeka is a 118 storey, 630 m mega tall skyscraper currently under construction in Kuala Lumpur, Malaysia. When completed in 2019, the tower will be the tallest building in Malaysia.

KVMRT SSP Line Tunnelling Works Wednesday April 25, 2018, 13:30–17:00

The Social Events

The KVMRT SSP line will serve a corridor with a population of around 2 million people stretching from Sungai Buloh to Putrajaya. The proposed alignment is 52.2 km of which 13.5 km is underground and to be constructed by tunnel boring machine (TBM). A total of 37 stations, 11 of them underground, will be built.

Young Engineers Reception and Ice Breaker Party – Cellar Door All Young Engineers (aged below 35) will be invited for an exclusive social event the evening preceding the first day of the IABSE Conference Kuala Lumpur 2018. At 20:00, the event will be opened to all delegates for an informal ice breaker event to meet

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Programmes free of charge will be offered to accompanying persons, such as FRIM canopy walk; Batik painting and KL city open air bus tour; a jungle trek; tribal experiences and Masak Masak cooking course; old city walking tour and a Royal Selangor pewter tour. More details and information can be found on: www.iabse.org/KualaLumpur2018.

Sponsorship/Exhibition A commercial exhibition will be held at the conference venue and there are sponsor packages to ensure all individual partner aspirations. Potential sponsors and exhibitors are invited to contact: paul@integer-asia.com or, niggeler@iabse.org

Important Dates Aug. 31, 2017: Acceptance Abstracts Nov. 30, 2017: Submission Full Papers Dec 31, 2017: Acceptance Full Papers Nov. 30, 2017: Final Invitation Nov. 30, 2017: Registration opens Jan. 15, 2018: Deadline Early Bird registration Apr. 25-27, 2018: Conference

More Information www.iabse.org/Kualalumpur2017

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IABSE News IABSE Denmark’s Structural Award

Cirkelbroen by night (photo: Anders Sune Berg for Nordea-fonden)

On March 31, 2017, the Danish Group of IABSE presented its Structural Award to Cirkelbroen (The Circle Bridge) in Copenhagen. The prize is awarded in recognition of a remarkable, innovative, interesting and challenging structure where the chosen alignment and spatial arrangement creates a unique public space inviting for reflection.

Cirkelbroen is a 35 m long bridge that spans Copenhagen’s Christianshavns Canal, designed by the DanishIcelandic artist Olafur Eliasson. The bridge is part of the pedestrian and cycle route along the quays of Copenhagen, a route that is used by thousands of people in Copenhagen every day.

The bridge consists of five steel platforms that vary in diameter from 10 to 14 m and are arranged to create a staggered, zigzag path across the water. The platforms are topped by tall masts from which thin steel cables extend to the platform’s circular railings, outlining barely perceptible conical screens around the masts—the

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From right: Lord Mayor Frank Jensen, artist Olafur Eliasson, chair of Nordea-fonden Mogens Hugo, engineer Lars Tørrild Thorbek, chair of IABSE-Denmark Tina Vejrum (photo: Niels Jørgensen for Nordea-fonden)

highest of which rises 25 m above the bridge.

One section of the bridge functions like a swing bridge, rotating on submerged pontoons to allow large boats to pass into and out of the canal. Two of the platforms are fixed to the foundations whereas the remaining three form the moveable part of the bridge.

These three platforms are supported by one column in the centre of the largest platform. From there the bridge is cantilevered with no support under the other two platforms except for a 110 m3 buoyancy tank that carries most of the self-weight of the moveable part.

Cirkelbroen has been selected as the official Danish logo of the UEFA soccer championship in 2020 hosted by 13 European cities to celebrate the 60th anniversary of UEFA.

The prize was awarded to: • The owner: Municipality of Copenhagen represented by Lord Mayor Frank Jensen • The artist: Studio Olafur Eliasson represented by Olafur Eliasson • The engineer: Ramboll Denmark represented by Lars Tørrild Thorbek • The client: Nordea-fonden represented by chairman Mogens Hugo At the event, the award winners explained how the concept was developed, their role, the importance of the project from their perspective and the special challenges that the unique design imposed. The engineer demonstrated how the bridge opens to let ships pass while the participants enjoyed a refreshment.

Tina Vejrum Chair, Danish Group of IABSE

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IABSE News 4th Young Engineers Colloquium Bochum 2017 - Report The jury awarded contributions:

the

following

YEC 2017 Oral Presentation Award 1st Prize Participation at the 39th IABSE Symposium in Vancouver 2017.

Question and answers session

Presenters of the oral presentations

From March 31 to April 1, 2017, the Young Engineers Colloquium took place in Germany for the fourth time.

This year, the colloquium was coorganised by the German and the Dutch Groups of IABSE, as well as the Institute for Structural Engineering of Ruhr-Universität in Bochum (RUB KIB).

presented their projects from research and construction practice, which is only 20% of the submitted abstracts and makes the colloquium extremely selective. In addition, 36 young engineers had been chosen for poster presentation in combination with an impulse lecture. The three minute impulse lectures were intended to give the young engineers the opportunity to present the essential aspects of their work to the audience and to stimulate the discussions for the following poster presentations. The impulse lectures were introduced as part of the Young Engineers Colloquium for the first time and were highly appreciated by the delegates. All conference contributions were also published in full extent in the proceedings.

The Young Engineers Colloquium series intends to provide an international platform for young engineers to meet and exchange ideas as well as knowledge between students, young engineers and experienced professionals across the borders. As part of the colloquium, young engineers under the age of 35 present their projects from engineering practice and academia to a broad audience.

Based on the number of submitted abstracts and the more than 100 conference delegates, participation was a great success. Considering the 60 contributions submitted by young engineers from six countries, the number of previous events was once more exceeded. All submitted abstracts were reviewed by a Scientific Committee consisting of international experts from industry and academia in advance of the colloquium.

The colloquium started with an introductory lecture by Dr. Roland von Wölfel on the arch bridges of the highspeed railway across the Thuringian Forest and their erection methods.

Afterwards, 12 young engineers selected by the Scientific Committee

Coffee breaks and lunch breaks offered excellent opportunities to socialise with good friends and colleagues, as well as to develop new relationships. The evening event at the ‘Alte Lohnhalle’ in Bochum was a fantastic moment to further intensify the discussions and relationships.

During the colloquium dinner the award ceremony for the oral and poster presentations took place, as well as further lectures like “Engineering for Extreme Winds” by Ahsam Kareem. Carsten Machentanz talked about the ‘Emscher Conversion Project’. His presentation gave a wonderful outlook for the technical excursion on the following day.

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Etienne Bouleau for “The ‘Jet d’Eau’ Movable Footbridge of Geneva” 2nd Prize Participation at the Future of Design Conference 2018 in London: Emilie van Wijnbergen for “Exploration and analysis of low-cost Seismic Retrofit Measures to Improve Boxaction for Traditional Brick Masonry Houses in Nepal” 3rd Prize Britte van Kortenhof for “Control of Vortex Shedding Diagonals Railway Bridge Muiderberg” and Frederik Bomholt for “Fatigue Resistance of Stay Cable Tendons on the Free Length and in the Anchorage Zone”

YEC 2017 Poster Award Ibrahim Selek for “Reliability of Nondestructive Testing Methods for Detecting Steel Rebar in Existing Concrete Structures”

RUB KIB Award 2017 Kars Haarhuis for “Ultra Slim Hybrid Staircase” and Jörn Remitz for “Fatigue of PostTensioned Steel” The technical excursion to the construction sites ‘Lennetalbrücke’ and ‘Emscher Conversion Project’ on the following day was the final highlight of the Young Engineers Colloquium 2017.

Prof. Marion Rauch, FHNW Prof. Markus Knobloch, RUB Members of the Organising Committee

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IABSE News IABSE Conference Bath 2017 - Report ‘Instilling Imagination and Innovation in Structural Design’ The IABSE Spring Conference 2017 was successfully held on April 19–20 in Bath, UK, and focused on creativity and collaboration in the structural engineering profession. The conference, which was proposed during the Stockholm Congress in September 2016, was excellently prepared by the Organising and Scientific Committees chaired by David Knight and Ian Firth, respectively, in just seven months.

The call for papers had clearly attracted people with a variety of backgrounds and interests. Presenters were required only to submit 2 page abstracts, and full 8 page papers were only submitted by authors who wished to do so. This opened the conference for speakers who could describe ideas and research in-the-making, but the contents were still vividly discussed during the presentations.

The conference was held in the University of Bath during two days, and included a balanced mix of keynote presentations, traditional individual presentations, discussion tables and workshops: a real opportunity to discuss trending topics and concerns and be actively involved throughout the whole conference.

A total of around 90 individual presentations were accepted that were divided into 17 sessions. Six seminars were held, in which a panel of industrial professionals and academics made presentations as a basis for a general discussion among both the panel and the delegates. The topics covered included: conceptual design, teaching, digital workflows, collaboration and integration, innovative methods for strengthening and various themed sessions (e.g. structural glass, balancing innovation and risk, codes versus innovation).

Technical Programme From the very beginning it was clear that the conference would have a somewhat different approach to interacting with and including the delegates, following the concept set in the Geneva Conference in 2015. Compared to traditional conferences that mainly make use of traditional oneway speaker–listener presentations, the Bath conference had also planned a number of workshops and seminars which were intended to include delegates actively in discussions and enable exchange of viewpoints.

Keynotes Seven inspiring and altogether very different keynote presentations were held, which were perfectly in line with the themes of the conference. Roger Ridsdill Smith (Foster + Partners), Wolf Mangelsdorf (Buro Happold), Vincent Servais (Bureau Greisch), Tim Ibell (University of Bath),

Kristina Shea (ETH Zurich), ClaudiaAntonia Merkle (Tate Modern, Tate Britain) and Ed McCann (Expedition Engineering) covered topics such as creativity, collaboration, design, education and computational design. The two finishing presentations “Dare to Break Rules!” by Claudia-Antonia Merkle and “The role of humour in high performing design teams” by Ed McCann made a nice ending. Both rounded off perfectly the creative and collaborative spirit of the conference and definitely gave some thoughts for the mind—as well as a good portion of laughter.

Workshops As a new feature, some hands-on design workshops were organised to actively involve delegates. Three workshops focused on the design of structures, such as a footbridge, a building and a shelter. The fourth workshop was about “How to have ideas” and presented innovative methods to be more creative. Workshops were attended by both younger and more senior delegates and the feedback was highly positive.

Following previous conferences, a Pecha Kucha session was also held at the end of the first day, in which

Keynote Speaker Claudia-Antonia Merkle (left) and Pecha Kucha speakers lined up (right) 470

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Gala Dinner in the Pump Room Restaurant and Roman Baths (left) and the vivid conversations during the breaks (right)

various themes were covered in a friendly and cheerful manner.

Venue The University of Bath set a perfect scene for the conference. With a tradition of educating students with emphasis on both architecture and structural engineering, the site was perfectly in the spirit of the conference. This was also demonstrated by a significant number of students from the university who participated and showed great interest in the conference and the presentations. In particular, the great design studios served as an excellent frame for the workshops.

A reception was held in the evening before the beginning of the conference at the Buro Happold offices to

welcome all delegates. The gala dinner at The Pump Room Restaurant followed the drinks reception at the Roman Baths: a spectacular scene for the evening. Some of the delegates even dared to sing and to play the piano and a harmonic!

Conclusions Altogether the conference was a refreshing and new approach to exchange of know-how and viewpoints and a great catalyst to engage in a more creative approach to future structural design. The round tables in the seminars served as a great format to engage the audience in vivid discussions, and the workshops helped to promote creativity among delegates.

In the words of one of the many young delegates: “The IABSE

conference was a unique event focused on creativity and collaboration, with the event itself being an excellent example of great collaboration between industry and academia.

The variety of talks was immensely interesting, with a personal highlight being the thought-provoking seminar sessions and debates” (Fiona Walport, Imperial College London).

The current IABSE President, Fernando Branco, closed the conference with the conclusion that the format of the conference represented a “new language”. We hope to see more conferences like this in the future!

Nikolaj Rask Pedersen, COWI IABSE Member, Denmark Fernando Madrazo-Aguirre, COWI IABSE Member, UK

Workshops on creativity and design held at the design studios

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IABSE News 2017 IABSE Future of Design Conference NYC - Reflections uncommon design technologies, from the combination of virtual reality and 4D BIM, to biomimetic concepts adapted for structures and facades.

Style and Substance Throughout the conference, innovative thinking was not only discussed, but also demonstrated in several ways.

Structural engineer Guy Nordenson (Princeton and Guy Nordenson and Associates) and writer Julian Rose (Artforum) gave a joint presentation in the format of an informal question and answer session, which featured debates about the nature of authorship and critiqued current tropes in design and construction.

“Future of Practice” Panelists, left to right: Josh Lobel (CW Keller + Associates), Ben Tait (UAP. Aaron Forrest (Ultramoderne), Christopher O’Hara (Studio NYL), Elizabeth Mahlow (Nous Engineering), Paul Kassabian (Simpson Gumpertz & Heger) (photo: Paul Laroque)

On April 29, 2017, the first IABSE Future of Design conference in New York City was a refreshing day of optimism and energy. Lee Franck (Guy Nordenson and Associates) launched the morning session by describing the organizing committee’s motivations to create a forum to celebrate “why we do what we do, what drives us as designers, and [what values we strive for].” To fulfil that mission, the conference brought together a diverse group of established and emerging talent in the fields of engineering, architecture, computation, fabricatio and art.

The role of new technologies in the design process was a unifying theme as presenters evaluated how significantly it participates in their own work and how it might be used to conceive new projects. The panel discussion touched on several

Fittingly for a conference that underscored the value of communicating the rationale and process of design to the general public, the events of panel discussions were recorded by Virginia Montgomery, a graphic artist, who illustrated key words and concepts of the presentations and panels. Her work presented a thought-provoking non-technical perspective that sparked conversation during the cocktail hour.

A Peek into the Future The morning session’s speakers illustrated how past projects and personal experiences shaped their philosophies and approaches to innovation. Their narratives demonstrated a spectrum of inspiration and collectively challenged the conventional definition of “design” by presenting projects that were developed in close partnership with experts not usually considered part of the architecture and engineering design world: biologists, programmers and historians.

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Graphic recording of 10 + 10 “Future of Design” presentation (Illustration by Virginia Montgomery)

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IABSE News Even the conference’s venue (the Manny Cantor Center in Manhattan) was a timely example of innovation and dialogue. Over dessert, architect Leonard Leung (PBDW Architects) and structural engineer Rebecca Buntrock (Silman) discussed their collaborative process of transforming the conference venue’s building. The design team used thoughtful planning and engineering to restore the historic 19th century facade and expand interior spaces, thus modernizing the interior for several key organizations that benefit the local community with various amenities and programs. Their conversation provided a tangible example of how roles of the design team frequently extend beyond technical aspects.

Collaboration and Beyond The multidisciplinary dialogue continued in the “Future of Practice” panel, when speakers discussed the value of long-term collaboration between architects, engineers and fabricators.

With consistent transfer of knowledge

Conference organizers gathering on the terrace during cocktail hour, left to right: Powell Draper (schlaich, bergermann and partner), Julia Chapman (Ennead), Alex Alaimo (PBDW), Lee Franck (Guy Nordenson and Associates), Rebecca Buntrock (Silman), Ryan Miller (Silman) (photo: Paul Laroque)

and interdisciplinary partnerships that are preserved across several projects, design solutions are less likely to be relegated to one-time use and perhaps can see second or third lives in new applications.

In the final panel session, “Future Contexts,” speakers presented design tools for emergency refugee camp planning, communication methods for relief workers that face language barriers, and disaster mitigation concepts for protecting coastlines from rising sea levels. Because distant populations or victims of natural disasters might not have the available resources or access to international consultants, the speakers spoke about their motivation to advocate for improved building standards around the world and promote the sharing of design tools developed for rapid response in times of crisis.

Looking Ahead Dessert talk by Leonard Leung (PBDW Architects) and Rebecca Buntrock (Silman) (photo: Paul Laroque)

The 2017 Future of Design conference highlighted the importance of creating relationships beyond architecture and

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engineering to meet future needs in design and construction. Innovation does not necessarily have to come in enormous, deep-budgeted leaps, applied uniquely to iconic structures— it can be incremental, and can come from any source of inspiration, and it can start with any engineer on day one of a project. According to Michael Stein (schlaich, bergermann and partner), we should think about this every day, and even “start Monday at 9 am.”

The optimism for engineering innovation, and successes and challenges of these breakthroughs in design will be celebrated in future IABSE Future of Design conferences in New York City.

For more images, and a complete list of conference sponsors and speakers please see the event program online at: https://fodnyc.org/ Hubert Chang, Knippers Helbig Jasmin Sadegh, Guy Nordenson and Associates

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IABSE News Trip to China

Chep Uytiepo introducing IABSE to the conference participants

IABSE Executive Director, Chep Uytiepo, was invited by Zhigang Yang, Chief Editor of BRIDGE Magazine (China) and member of the Chinese IABSE Group, to attend their event “Symposium on Bridge Cable System Development and Structural Innovations” held in Wuhu, Nanjing on 18–19 May, 2017. The Symposium was expected to attract construction businesses, structural engineers and government officials in China, especially around the region.

This is not the first time that we are collaborating with each other for an event, Chep said. In fact, they were the PCO behind IABSE Spring Conference 2016 in Guangzhou which was very successful and attended by over 1100 local and international delegates. Seen as an irresistible opportunity to boost promotion of IABSE in China, the Executive

Director decided to join the event with Communications and Marketing Manager, Sissel Niggeler, to set up closer collaboration between our organisations.

The event started with an exclusive VIP technical tour to the 2nd Wuhu Bridge which has a new innovative U-shape anchorage system. The day was concluded by a dinner offered by the Vice Mayor and the Chief Engineer of the Ministry of Public Transport of Anhui Province. Also invited were local contractors, keynote speakers Myint Lwin, Phillip Yen, Man Chung Tang, Naeem Hussein and IABSE Vice President and Chair of the Chinese Group, Yaojun Ge. The next day, Chep introduced IABSE to over 300 conference participants with a short presentation about IABSE during the Opening Ceremony. A big thank you goes to the organisers for

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Visiting the Wuhu Bridge

their generosity in providing a complimentary booth to IABSE that gave IABSE a prominent exposure throughout the whole event. In addition, Sissel appreciated the personal help of Emma Yang and Jack Zhao of Bridge Magazine (China) for their personal effort in providing translation assistance when she was having conversations with the locals, especially to potential advertising partners. The successful event ended with an interview to be featured in the next issue of the Bridge Magazine (China) including promotion of IABSE membership in Chinese language.

Following this trip to China, Chep Uytiepo is looking forward to strengthening IABSE’s collaboration with the BRIDGE Magazine (China), together with its Chinese Group in boosting the growth of membership in China in the near future.

Group photo with the BRIDGE Magazine team

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IABSE News IABSE Book Vernissage: SED 5 – 3rd Edition ‘Introduction to Safety and Reliability of Structures’ informative and more educational type of text and figures on safety and reliability analysis in structural engineering. It is intended both for students and practising engineers and aims to keep things understandable and to explain concepts and procedures by simple examples rather than by digging deep into the theory. A second edition was printed in 2006, and was supplemented with a free educational type of Variables Processor software, FreeVaP. The book was sold out in 2016 and IABSE decided to print this third edition. Considering the progress made in the past 20 years, a number of new chapters have been introduced, and additions and corrections have been made resulting in some 40% increase of volume.

The authors signing the books

The Book Vernissage SED 5 – 3rd reviewed and extended Edition “Introduction to Safety and Reliability of Structures” by Jörg Schneider and Ton Vrouwenvelder was launched on June 15, 2017, at a vernissage organised by IABSE, on the premises of ETH Zurich. Professors from ETH, representatives from the profession and PhD students came to meet the authors and to listen about the history and development of the book with printouts, booklets and personal stories, going all the way back to a common friend of the authors, Prof. F.K Ligtenberg, TNO Delft University (1922–1990), who at that time had influenced Jörg Schneider with his publication “Concepts of Safety of Structures and Methods of Design” (IABSE reports of the Working Commissions, 1969), on this topic in 1969. After the interesting book presentation, the participants enjoyed

an apéro and had their books signed by the authors.

It is hoped that this book serves its purpose in furthering this topic which is gaining more and more attention and finding increasing application in practice. SED 5 – 3rd Edition is sold online under the new publication pricing scheme and is available at its retail price excluding shipment. The publication comes with a free Variables Processor software, FreeVaP and is available at: www.iabse.org/onlineshop IABSE Members can download the electronic version in the Members Area at: www.iabse.org

About this Edition SED 5 third edition is based on IABSE’s Structural Engineering Document SED 5, published in 1997 upon request of Working Commission 1 of IABSE and supported by the Joint Committee on Structural Safety (JCSS). The book offers a short,

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Ton Vrouwenvelder and Jörg Schneider

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IABSE News Calendar of Events *2017 September 6–8 Berlin, Germany ‘Footbridge 2017’ 2017 September 19–23 Vancouver, Canada 39th IABSE Symposium ‘Engineering the Future’ Annual Meetings precede Symposium *2017 September 30-October 1 Zilina, Slovakia ‘9th international Conference on Bridges in Danube Basin’ *2017 October 16–21, Santiago, Chile, 2nd International Bridges Congress – Chile 2017 ‘Design, Construction and Maintenance’

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*2017 October 30-31 Porto, Portugal ‘ISDAC 2017 - International Symposium on the Dynamics and Aerodynamics of Cables’

IALCCE 2018 ‘The Sixth International Symposium on Life-Cycle Civil Engineering’

2018 April 25–27, Kuala Lumpur, Malaysia, IABSE Conference ‘Engineering the Developing World’

2019 March 27-29 Guimaraes, Portugal IABSE Conference ‘Towards a Resilient Built Environment Risk and Asset Management’

2018 September 19–21 Nantes, France 40th IABSE Symposium ‘Tomorrow’s Megastructures’ Annual Meetings precede Symposium

2019 July 24–26 Lisbon, Portugal ICSA 2019 ‘The 4th International Conference on Structures and Architecture’

*2018 October 28–31 Ghent, Belgium

*co-sponsored or supported by IABSE

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A matter of confidence.

Stressing of stay cables PAUL supplies s Prestressing installations incl. planning work s Anchor grips s Prestressing machinery (single-/multi-stressing jacks) s Strand pushing and cutting equipment s Automatic prestressing machines for railway sleepers s Prestressing equipment for bridge construction (prestressing cables and stay cables)

The experts in Prestressed Concrete Technology.

www.paul.eu

Paul at YouTube stressing-channel.paul.eu

Max-Paul-Str. 1 88525 DĂźrmentingen Germany +49 (0) 73 71 / 5 00 - 0 +49 (0) 73 71 / 5 00 - 111 spannbeton@paul.eu

SEI Volume 27 | Number 3 | August 2017 Structural Engineering International

STRUCTURAL ENGINEERING INTERNATIONAL International Association for Bridge and Structural Engineering (IABSE)

Volume 27 | Number 3 | August 2017

Building for the future.

TWIN RIVER BRIDGE, CHONGQING, CHINA T.Y. Lin International (TYLI) designed the Dongshuimen Bridge across the Yangtze River and the Qianximen Bridge across the Jialing River – two iconic structures known collectively as the Twin River Bridge in Chongqing, China. Interconnected by a tunnel, the bridges allow seamless transport between the city’s southern district, downtown Chongqing, and the northern district. Both structures carry vehicular lanes and pedestrian paths on the upper decks and light rail on the lower decks.

TYLI’s innovative design resulted in two partially cable-supported girder bridges, an economical, structurally efficient concept that provides a highly aesthetic, transparent configuration. The simple, modern towers also dramatically express Chongqing’s dynamic combination of Photo: Yaping Lai

www.tylin.com

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both historical and contemporary features.

Forensics in Structural Engineering 10/07/17 9:39 PM