February 2018 Volume XXIV, Number 1 (3)

Environmental & Engineering Geoscience FEBRUARY 2018

VOLUME XXIV, NUMBER 1

SPECIAL ISSUE ON DAMS GUEST EDITORS: BRIAN H. GREENE AND KERRY CATO

THE JOINT PUBLICATION OF THE ASSOCIATION OF ENVIRONMENTAL AND ENGINEERING GEOLOGISTS AND THE GEOLOGICAL SOCIETY OF AMERICA SERVING PROFESSIONALS IN ENGINEERING GEOLOGY, ENVIRONMENTAL GEOLOGY, AND HYDROGEOLOGY

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Single copies are $75.00 each. Requests for single copies should be sent to AEG, 1100 Brandywine Blvd, Suite H, Zanesville, OH 43701-7303. © 2018 by the Association of Environmental and Engineering Geologists All rights reserved. No part of this publication may be reproduced or transmitted in any form or by any means, electronic or mechanical, including photocopying, recording, or by any information storage and retrieval system, without permission in writing from AEG. THIS PUBLICATION IS PRINTED ON ACID-FREE PAPER Abdul Shakoor Department of Geology Kent State University Kent, OH 44242 330-672-2968 ashakoor@kent.edu

EDITORS

Brian G. Katz Florida Department of Environmental Protection 2600 Blair Stone Rd. Tallahassee, FL 32399 850-245-8233 eegeditorbkatz@gmail.com

EDITORIAL BOARD Jerome V. DeGraff CSU Fresno Chester (Skip) F. Watts Radford University Thomas Oommen Michigan Technological Univ.

Syed E. Hasan University of Missouri

Thomas J. Burbey Virginia Polytechnic Institute Abdul Shakoor Kent State University Brian G. Katz Florida Department of Environmental Protection

ASSOCIATE EDITORS John W. Bell Nevada Bureau of Mines and Geology Richard E. Jackson Geofirma Engineering, Ltd. Jeffrey R. Keaton AMEC Americas Paul G. Marinos National Technical University of Athens, Greece June E. Mirecki U.S. Army Corps of Engineers Peter Pehme Waterloo Geophysics, Inc Nicholas Pinter Southern Illinois University

Paul M. Santi Colorado School of Mines Robert L. Schuster U.S. Geological Survey Roy J. Shlemon R. J. Shlemon & Associates, Inc. Greg M. Stock National Park Service Resat Ulusay Hacettepe University, Turkey Chester F. “Skip” Watts Radford University Terry R. West Purdue University

­­­­­­SUBMISSION OF MANUSCRIPTS Environmental & Engineering Geoscience (E&EG), is a quarterly journal devoted to the publication of original papers that are of potential interest to hydrogeologists, environmental and engineering geologists, and geological engineers working in site selection, feasibility studies, investigations, design or construction of civil engineering projects or in waste management, groundwater, and related environmental fields. All papers are peer reviewed.

The editors invite contributions concerning all aspects of environmental and engineering geology and related disciplines. Recent abstracts can be viewed under “Archive” at the web site, “http://eeg. geoscienceworld.org”. Articles that report on research, case histories and new methods, and book reviews are welcome. Discussion papers, which are critiques of printed articles and are technical in nature, may be published with replies from the original author(s).

Discussion papers and replies should be concise.

To submit a manuscript go to http://eeg.allentrack.net. If you have not used the system before, follow the link at the bottom of the page that says New users should register for an account. Choose your own login and password. Further instructions will be available upon logging into the system. Please carefully read the “Instructions for Authors”.

Authors do not pay any charge for color figures that are essential to the manuscript. Manuscripts of fewer than 10 pages may be published as Technical Notes.

For further information, you may contact Dr. Abdul Shakoor at the editorial office.

Cover photo Crafton Hills Reservoir, San Bernardino County, California. The first dam (center of photo) was completed in 2002; the second dam (left-of-center) was completed in 2014 to expand reservoir capacity with removal of the adjoining ridge between the two drainages. See article on page 23. Photo credit: California Department of Water Resources.

Foreword BRIAN H. GREENE Gannett Fleming, Inc., 730 Holiday Drive, Building 8, Suite 400, Pittsburgh, PA 15222

KERRY CATO Department of Geological Sciences, California State University, San Bernardino, 5500 University Parkway, San Bernardino, CA 92407

This special issue of Environmental and Engineering Geoscience (E&EG) had its origins from the Dams Symposium held at the 2016 Annual Meeting of the Association of Environmental & Engineering Geologists (AEG) in Kona, HI. Many interesting papers on new and existing dams were presented over this 2-daylong symposium. At the Kona Annual Meeting, it was proposed that a special journal issue be published to capture many of the outstanding papers that came out of this conference. Dams represent an important part of the aging infrastructure within the United States. Federal government agencies, state agencies, universities, consulting firms, and individual consultants have been working on many important projects. It is very clear that, although few new dams are being constructed, this era of our practice is a renaissance in dam technology and understanding, compelled by the necessity of extending the life of dams beyond what they were designed for, in some cases 75 years or more.

It is with pride that we took on the role of guest editors for this journal issue of E&EG focusing on the topic of dams. Members of the AEG Dams Technical Working Group contributed to the special issue, both as co-authors of papers, and serving as peer reviewers of draft manuscripts. The efforts of these individuals are greatly appreciated. The work of volunteer peer reviewers contributed significantly to the publication of this special issue of the journal. Peer review is an arduous but important task that is vital to maintaining the quality of published journal papers.

The articles in this issue define the full range of technical aspects of dam assessment, evaluation, and remediation. Highlights of each paper are given below: Michael Nield, of the U.S. Army Corps of Engineers Dam Safety Production Center in Huntington, WV, reviewed the construction of a grout curtain emplaced at the Corps’ Bolivar Dam, located in northern Ohio.

The embankment dam is founded on glacial outwash and ties into bedrock at its left abutment. Ongoing seepage in the bedrock had been monitored, and as part of the overall dam remediation work, a doubleline grout curtain was installed in the left abutment to a depth of 65 ft (19.8 m), over a distance of 400 ft fmii

(121.9 m). The paper reviews risk-informed decisions that were made during design and construction of the grout curtain, and it presents important lessons learned from this project.

Holly Nichols, of the California Department of Water Resources, writes on the seepage investigation for remedial grouting at the Crafton Hills Reservoir Project.

Located in San Bernardino County, CA, the Crafton Hills Reservoir experienced areas of seepage upon refilling of the enlarged reservoir. After monitoring and investigations, the decision was made to seal off seepage pathways with foundation grouting. The geologic investigation is described in the paper, which reviews the use of multiple testing methods, leading to a more robust plan of remedial grouting.

Co-authors Bruce Hilton, Ronn Rose, William McCormick, and Todd Crampton review the use of highresolution light detection and ranging (LiDAR) in fault delineation at the location of the existing Martis Creek Dam, located in Truckee, CA. The embankment dam, operated by the U.S. Army Corps of Engineers, has been rated as very high risk and in need of remedial work. A history of excessive seepage during even moderate reservoir levels has prevented the project from fulfilling its full potential water storage function. Use of high-resolution LiDAR uncovered a previously unknown through-going lineament between the spillway and the main dam embankment. Paleoseismic trenching at the site was performed. The fault trenching confirmed the fault zone location. LiDAR was a key factor in discovering lineaments that subsequently revealed faulting at the site.

In their paper, Kerry Cato and J. David Rogers examine Alexander Dam, located on the Hawaiian island of Kauai. The hydraulic fill dam was constructed in 1929– 1932 to provide irrigation for sugar cane fields. The dam was designed to have a maximum height of 125 ft (38 m) and to store a reservoir volume of 800 million gallons of water. On March 29, 1930, a section of the core pool 60 ft (18.3 m) wide suddenly dropped 30 ft (9.1 m).

The volume of the slide debris mass was approximately 275,000 cubic yards (210,000 m3 ). At the time of failure, the embankment was 95 ft (29 m) high. The failure

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Foreword

occurred extremely rapidly and resulted in the deaths of six workmen, and it injured an additional two. The paper examines the cause of the failure and the rebuilding of the dam in 1932, and it shows that hydraulic fill structures can be hazardous during construction. The reconstructed hydraulic fill embankment, with its mitigation measures, has performed well for 85 years, and it is the second highest embankment dam in Hawaii.

Todd Loar’s paper provides a qualitative analysis of wedge stability performed for Green Peter Dam, located in west-central Oregon. A risk assessment identified continuous and adversely oriented low-angle rock shear zones underlying portions of the foundation of the concrete gravity dam. These shear features were of concern because they could potentially facilitate instability of one or more dam monoliths during earthquake loading. After performing a qualitative evaluation of the geometry of rock wedges, the study revealed that the wedges would likely be stable under even large probabilistic seismic loading. The paper provides very useful information on methodologies that can be employed for evaluating similar concrete gravity dams that could be subject to seismic loads.

In a paper that summarized the Keynote Lecture at the 2016 Kona Dams Symposium, J. David Rogers describes the construction and operation of Gatun Dam, a megastructure of the Panama Canal. Gatun Dam was a kingpin of the American scheme to construct a ship canal across Panama in 1907–1914 to connect the Atlantic and Pacific Oceans. Gatun Dam was an earthen structure of unprecedented scale that was designed to retain the aggregate flow of the Chagres River and all its principal tributaries. The dam was the key structure— it created a man-made lake that, rising 85 ft (25.9 m) above sea level, permitted ships to cross the continental divide in Panama. The dimensions of Gatun Dam were enormous and without precedent at the time of its construction. It had a crest length of 8,200 ft (2,500 m), a maximum width of 2,300 ft (701 m), and a height of 105 ft (32 m). The dam stored a huge operating pool and included a critical feature: a mass concrete spillway capable of passing very unpredictable flow volumes from the Chagres River. The paper describes the challenges imposed during construction of the dam related to the underlying geology.

Co-authors David Schug, Paul Salter, Christopher Goetz, and Derek Irving describe the fault investiga-

tions that were undertaken for the Borinquen Dam 1E of the new Pacific Access Channel of the Panama Canal Expansion. The 1.4-mi-long (2.3-km-long) zoned rockfill dam forms the key navigational channel providing access from the Gaillard Cut to the new Post-Panamax Pacific Locks of the Panama Canal. During construction, an important objective for project geologists was to confirm the locations and activity of faults mapped in the foundation for the dam during design. A significant feature was the Pedro Miguel Fault and its trace with respect to the dam. The paper describes the paleoseismic trenching that was performed, as well as age dating of alluvium overlying the faults. The paper describes the widening of the core and filters of Dam 1E to accommodate potential fault rupture of the Pedro Miguel Fault. Further conclusions derived from the extensive fault studies are presented.

Co-authors Vanessa Bateman and Georgette Hlepas of the U.S. Army Corps of Engineers describe in their paper the important lessons learned from construction of a seepage barrier wall at the Corps’ Wolf Creek Dam in Kentucky. The agency is maintaining a lessons-learned goal for all major projects to capture knowledge gained. This is of key importance as future projects are designed and constructed. From the first barrier wall installed in the 1970s at Wolf Creek Dam to the recent deep barrier wall emplaced at this project, as well as at other embankment dams, documentation of and sharing of important lessons learned in the areas of foundation grouting, data management, and quality control procedures have been documented. In doing so, the paper describes the increased efficiencies and effectiveness of barrier wall designs, preparation of specifications, and project monitoring, with the goal of improving current and future barrier wall projects.

Environmental and Engineering Geoscience co-editor Abdul Shakoor was invaluable in maintaining the schedule of this publication and in offering frequent assistance to authors requiring guidance on manuscript format. In addition, E&EG publications assistant Karen Smith was a constant source of support and guidance, providing us with all that was needed to assemble this issue on dams from its inception to final publication. The AEG Dams Working Group recognizes and appreciates support from the Association of Environmental and Engineering Geologists for the publication of these papers in E&EG.

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Environmental & Engineering Geoscience Volume 24, Number 1, February 2018 Table of Contents Guest Editors: Brian H. Greene and Kerry Cato 1

Gatun Dam—Megastructure of the Panama Canal J. David Rogers and Manuel H. Barrelier

23

Seepage Investigation for Remedial Grouting, Crafton Hills Reservoir, California Holly J. Nichols

39

Pedro Miguel Fault Investigations: Borinquen Dam 1E Construction and the Panama Canal Expansion David L. Schug, Paul Salter, Christopher Goetz, and Derek Irving

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Qualitative Rock Wedge Stability Evaluation Performed for Foundation of Green Peter Dam, Oregon Todd N. Loar

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Lessons Learned from USACE Seepage Barrier Wall Construction: Wolf Creek to Present Georgette Hlepas and Vanessa Bateman

89

Failure of the Alexander Dam Embankment and Reconstruction Using Drainage Mitigation on Kauai, Hawaii, 1930–1932 Kerry D. Cato and J. David Rogers

111

Discovering the Polaris Fault, Martis Creek Dam,Truckee, California Lewis E. Hunter, Ronn S. Rose, Bruce Hilton, William McCormick, and Todd Crampton

121

Grout Curtain Construction at Bolivar Dam, Ohio Michael C. Nield

fmi

Gatun Dam—Megastructure of the Panama Canal J. DAVID ROGERS

∗

Missouri University of Science and Technology, Rolla, MO 65409

MANUEL H. BARRELIER Engineering Division, Panama Canal Authority, Panama City, Panama

Key Terms: Dams, Canals, Hydraulic Sluicing, Hydraulic Fill, Landslides, Bearing Failure, Embankments ABSTRACT The kingpin structure of the American scheme to construct a ship canal across Panama in 1907–1914 was an earthen dam of unprecedented scale and scope at Gatun, to retain the aggregate flow of the Chagres River and its principal tributaries. Upon this structure alone, the entire plan rested, because it created the man-made lake rising 85 ft (25.9 m) above sea level, which allowed ships to cross the 550 ft (167.6 m) continental divide between the Atlantic and Pacific Oceans. Everything about Gatun Dam was enormous. Its dimensions were without precedent: a crest length of 8,200 ft (2,500 m) and a maximum width of 2,300 ft (701 m). With a height of 105 ft (32 m) above sea level, it stored sufficient water to maintain an operating pool covering 164 mi2 (425 km2 ). At its center was the most critical structure, a mass concrete spillway capable of passing flood flows of the unpredictable Chagres River.

The biggest problem with the site was the underlying geology, which included two deepley incised paleo-channels.

The massive embankments were placed over these paleochannels, which were up to 258 ft (78.6 m) deep. The channel infill of the upper 50 ft (15.2 m) was of relatively low permeability, mostly sandy silts and clay. There were more pervious sands and gravel lying beneath these, which allowed deep seepage cutoffs to be precluded.

DECISION TO CONSTRUCT A DAM AT GATUN The original American scheme in 1903 for construction of a dam along the Chagres River was at a natural narrows, 17 mi (27 km) upstream of the river’s mouth on the Atlantic coast, near a small village named Bohio.

The proposed earthen dam was 100 ft (30.5 m) high, where the French had planned to construct a similar dam 75 ft (22.9 m) high. This scheme was championed by George Morison, a New York bridge engineer on the Americans’ first Isthmian Canal Commission ∗ Corresponding

author email: rogersda@mst.edu.

(ICC), appointed in 1899. In 1902, Morison envisioned a summit lake 90 ft (27.4 m) above sea level, utilizing two locks on either coast. He was the only member of the commission favoring a locked canal across the Panamanian Isthmus.

After convincing President Theodore Roosevelt of the advantages of a canal across Panama in lieu of Nicaragua, the Bohio dam site was probed with deep borings and deemed unsuitable because the depth to bedrock was as much as 158 ft (48 m) below sea level. In an article for the ASCE Transactions (Morison, 1902), he compared the foundation conditions at the Bohio dam site with those of the North Dike of the Wachusett Reservoir (near Worchester, MA), which appeared to possess similar geologic conditions.

In the fall of 1904, the American engineers working for the second ICC learned that channel fill lying beneath the Bohio Dam site (8 mi [12.9 km] upstream of the village of Gatun) was filled with pervious sands and gravels to depths as great as 168 ft (51.2 m) below sea level. The presence of pervious materials was a revelation that worried the ICC (Rogers, 2014). This discovery of “changed conditions” reminded everyone of the innumerable problems that had plagued French efforts to construct a canal during the two previous decades.

In June 1905, an International Board of Consulting Engineers (IBCE) was formed to settle the debate about whether the Americans should construct a locked or a sea-level canal across Panama. The board was of unprecedented size and diversity, composed of 13 engineers: five from Europe, six American civilians, and two U.S. Army engineers. Five of the board members went onto write a “minority report,” recommending a locked canal with a massive embankment dam (IBCE, 1906).

These individuals were American engineers Alfred Noble, Isham Randolph, Frederic P. Stearns, Joseph Ripley, and Henry L. Abbot (Figure 1).

Influence of Wachusett Dam Project In September 1905, IBCE member Frederic T. Stearns was the incoming president of the American Society of Civil Engineers. Just before the IBCE

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Figure 1. The International Board of Consulting Engineers at their inaugural meeting in Washington, D.C., on September 1, 1905. From left, standing: Secretary John C. Oakes, General Henry L. Abbot, Eugen Tincauzer, Edouard Quellennec, Isham Randolph, Frederic P. Stearns, and Professor William H. Burr. From left, sitting: Joseph Ripley, William Henry Hunter, Adolphe Guerard, J. W. Welcker, Alfred Noble, General George W. Davis, and William Barclay Parsons (source: Bentley Library, University of Michigan).

departed New York for their initial reconnaissance of Panama, Stearns invited its members to take a field trip to Worchester, MA, to tour the Wachusett Reservoir along the Nashua River, which was then nearing completion.

The North Dike of the Wachusett Dam (Figure 2) was 65 ft (19.8 m) high and 2 mi (3 km) long, and it contained over 5,500,000 yd3 (4,202,000 m3 ) of earth fill, at that time the largest embankment in the world.

A massive cutoff trench 5,245 ft (1,599 m) long and 50 ft (15.2 m) deep, with 45 degree side slopes, had been excavated beneath the dike, and wooden sheet piles had been driven to depths of 46–48 ft (14–14.6 m) from the base of this excavation (Figure 2), flooring in glacial tills at a depth of 150 ft (45.7 m) beneath existing grade (Stearns, 1902). The similarity to the geologic conditions at the Bohio and Gatun sites in Panama was striking; both sites were underlain by pervious unconsolidated materials at depths >50 ft (15.2 m), overlain by increasingly finer grained materials grading upward.

The Unpredictable Flow of the Chagres River On December 19, 1905, the Panama Canal’s chief engineer, John Frank Stevens, sent a detailed report to the ICC and each member of the IBCE (Stevens, 1928).

It contained extensive data on the fluctuating flow of the Chagres River and the other 37 water courses traversed by the canal, 18 of which were intersected by the Culebra Cut through the continental divide, where the deepest cuts were proposed (>500 ft [>152 m]).

Figure 2. (Upper) Massive cutoff trench for the North Dike of the Wachusett Dam, during driving of the sheet-pile cutoff in 1899. (Lower) Geologic profile along the North Dike, showing the depth of the sheet piles (Stearns, 1902).

2

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Gatun Dam—Megastructure of the Panama Canal

Figure 3. Original design concept for the Gatun Dam, prepared in February 1906, which assumed a 1:25 (vertical to horizontal) slope for the downstream face (IBCE, 1906).

Stevens (1928) recounted how he had been predisposed to constructing a sea-level canal but, upon observing the highly irregular flows of the Chagres River, had changed his mind. He noted that the base flow of the Chagres was about 750 cubic feet per second (cfs) (25.25 m3 /s), but that within just 12 hours, the flow could increase to as much as 150,000 cfs (5,250 m3 /s) because of intense tropical storms.

Stevens asserted that the only plausible manner to control the Chagres would be to entrap it in an enormous reservoir with massive spillways. An expansive reservoir could absorb the extreme fluctuations in flow and accommodate large volumes of sediment that would surely play havoc with any alterative scheme, seeking to separate the Chagres from the canal. Stevens’ foresight in this matter reveals that he was much more than just a “railroad engineer.”

In 1904, Charles D. Ward penned an article titled “The Gatun Dam,” which was published in the prestigious ASCE Transactions (Ward, 1904). At the time, he served as chief of maintenance and extension for the Brooklyn Waterworks. In 1880, Ward and ASCE President Ashbel Welch had visited Panama to view the French attempts to construct a canal. While touring the isthmus, Ward suggested that a dam approximately 1.25 mi (2 km) long could be constructed across the lower Chagres River at Gatun, to capture that stream’s discharge for a locked canal, and the seed of this concept was planted.

The Minority Report Recommends a Dam at Gatun The IBCE minority members Noble, Randolph, Stearns, Ripley, and Abbot were impressed with the testimonies of Chief Engineer John Frank Stevens and Army Engineer Major Cassius E. Gillette, who pointed to a large volume of data that favored the construction of a locked canal in order to control the tempestuous Chagres River. The reduced volume of excavation would mean a locked canal could be completed within a 10 year time frame, while the sea-level canal was estimated to need an additional 4 years to construct.

After viewing the North Dike of the Wachusett Dam and the Gatun Dam site, the Minority Report of the IBCE concluded that the thick cover of low permeabil-

ity materials would make an excellent foundation for an earthen embankment dam (IBCE, 1906). They also recommended that additional borings be undertaken at the dam site before finalizing its design, to ascertain the footprint of the filled paleo-channels.

The Minority Report was released in February 1906, and it presented a practical scheme for a locked canal across Panama, with the most prominent structure being a massive earthen dam at Gatun (Figure 3), just 3.5 mi (5.63 km) from Limon Bay on the Atlantic Coast.

It would be the largest earthen dam ever constructed up until that time and would retain a year-round reservoir with a surface area of 110 mi2 (177 km2 ), which later swelled to 164 mi2 (264 km2 ). This lake would be able to absorb flashy tropical floods from the Chagres watershed.

The most challenging aspect of the Gatun site was how to handle subsurface seepage through the deeply incised paleo-channels, filled with unconsolidated sediments to as much as 258 ft (78.6 m) below sea level (Figure 4). The low bearing capacity of these surficial deposits, and the likelihood for long-term differential settlement were without precedent at the time decsions were made (late 1905). The unusually wide footprint of the two massive embankments was intended to spread their weight over the paleo-channels, to lessen the severity of differential settlement. The extremely broad footprint scheme was based on 27 borings advanced before September 1905, shown in Figure 4. The principal difference between the Bohio and Gatun sites was that the paleo-channel fill was considerably less pervious at Gatun than the materials encountered at similar depths at the Bohio site, which could make under-seepage and uplift more troublesome. The 1906 Minority Report presumed that a deep keyway with sheet-pile cutoff wall, similar to that employed for the North Dike of the Wachusett Dam, could be employed across the two incised paleo-channels.

The ICC’s chief engineer, John Frank Stevens, agreed with the “minority plan” to construct a locked canal at an elevation of 85 ft (25.9 m), using water from the Chagres River to create a vast inland lake. This reduced the required depth of excavations by 70 vertical feet (21.3 m). The Minority Report plan was favored by President Roosevelt and eventually approved

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Figure 4. Initial geologic section along the centerline of the proposed Gatun Dam, locks, and spillway, from the February 1906 Minority Report of the International Board of Consulting Engineers. The dam would traverse two deeply incised paleo-channels of the Chagres River (IBCE, 1906).

by Congress on June 29, 1906, after several months of Congressional inquiry. Though initially skeptical of a locked canal in Panama, members of Congress were persuaded by the confident testimony of Chief Engineer Stevens. Although Stevens’ premature departure in April 1907 precluded his being credited as the father of the Panama Canal, history has shown that he played a pivotal role in most of the critical aspects of the project, which was completed 7.5 years later.

In November 1906, Theodore Roosevelt became the first American president to leave the United States while in office when he traveled to Panama to visit the canal project. The president’s visit was filled with surprises, dispensing of protocol and schedule. He wanted to see first-hand what the average worker thought of the project’s management. He arrived at the height of the rainy season, when things could be deplorable, but it was abnormally dry, except for one day when it rained 3 in. (7.6 cm) in 2 hours.

During the first day of touring, it took the president almost 10 hours to cross the isthmus from Colon to Panama City, because he insisted on stopping his train and questioning the construction workers (the normal rail journey across the isthmus took about 1 hour and 20 minutes. At the Gatun Dam site, Dredging & Harbors Division Engineer Frank Maltby (Figure 5) briefed Roosevelt on the great earthen barrier that would stretch 1.5 mi (2.4 km) across the valley of the Chagres River (Maltby, 1945). Several prominent civil engineers voiced opinions that the vast lake would never hold its reservoir pool due to bank infiltration and excessive evaporation (Bates, 1906; Sorzano, 1910).

early work on Gatun Dam, which focused on clearing and grubbing of the dam site, and he began diverting muck from the lock excavations to be placed as fill for containment dikes at the upstream and downstream margins of the dam’s clay core pool (Figure 6). Gerig departed Panama after the reorganization of mid-1908, instituted by Lieutenant Colonel George W. Goethals, who assumed responsibilities as chairman and chief engineer of the ICC on April 1, 1907, after John Frank Stevens departed.

In March 1907, Major William L. Sibert (Figure 7) of the U.S. Army Corps of Engineers was appointed to the ICC by the newly appointed ICC chief enginer, Lieutenant Colonel George W. Goethals. Both men graduated near the top of their respective classes at West Point (Goethals in 1881 and Sibert in 1884). Sibert had the most experience with construction of locks and dams of any officer in the Army because he had supervised projects at the Soo Locks in upper Michigan, as well as numerous projects along the Ohio River and its principal tributaries while commanding the Army Corps of Engineers district offices in Louisville, KY, and Pittsburgh, PA. Sibert was recommended to

DESIGN METHODOLOGY In 1905, William Lee Gerig, a graduate of the University of Missouri whose father had worked on the Suez Canal, arrived in Panama to serve as chief engineer for construction. In the spring of 1907, he directed the 4

Figure 5. Division Engineer Frank Maltby points to the Gatun Dam site as President Roosevelt looks on (in white suit), from the promontory forming the dam’s right abutment (courtesy of National Archives and Records Administration).

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Gatun Dam—Megastructure of the Panama Canal

Figure 6. Side-dumping railcars being loaded with muck from excavation for the Gatun locks, which began being placed along the dam’s upstream and downstream margins in 1907 (courtesy of National Archives and Records Administration).

Goethals by Major David Gaillard, who had been Sibert’s roomate at West Point. Prior to being appointed, Gaillard and Goethels were members of the Army’s General Board in Washington, DC, composed of 43 officers from various branches of the service.

Given Sibert’s experience and background, from March 1907 through June 1908, he supervised the structural design of all lock and dam construction on the canal. These responsibilities included the design of the Gatun locks, dam, and spillway. In the reorganization of June 30, 1908, Goethals split the canal project into three geographic sectors, appointing three division engineers with over-arching authority over every aspect

of construction in the Atlantic, Central, and Pacific Divisions.

Sibert was given command of the Atlantic Division, which included the approach channels, Gatun Locks, Gatun Dam, Gatun spillway and outfall channel, and Powerhouse (Figure 8). Over the succeeding months, the design of locks and dams was also reorganized and placed under the command of Lieutenant Colonel Harry F. Hodges.

Sibert was promoted to lieutenant colonel in September 1909. His Atlantic Division engineers (Figure 7, right) had charge of designing and constructing the Gatun Dam and spillway, the approach channel excavation, and the Gatun triple locks along the dam’s right abutment. In carrying out these responsibilities, Sibert was assisted by some of the brightest military and civilian engineers, who went onto considerable fame. These included: Captain (later Major) George M. Hoffman, assistant division engineer in charge of the construction of Gatun Dam; Caleb Saville, a graduate of Harvard in 1889, who had previously worked on the North Dike of the Wachusett Dam; and Edward C. Sherman, a graduate of Massachusetts Institute of Technology (MIT) in 1898, who supervised the final design and construction of the Gatun spillway and outfall channel. These men are shown in Figure 9.

Saville’s Design of August 1908 The detailed site investigation and design work on the dam began in December 1907 under the direction of Caleb Saville, summarized in his 70-page report of August 30, 1908, which included 112 plates

Figure 7. (Left) William L. Sibert commanded the Atlantic Division of the Panama Canal, which included Gatun Dam (courtesy of the Library of Congress). (Right) Sibert (seated seventh from right) and the office staff of the Atlantic Division at Gatun in 1911 (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

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Figure 8. Map showing the Atlantic Division of the Panama Canal, which extended 8 mi (12.9 km) from the open Atlantic Ocean to Gatun and included construction of the triple locks, embankment dam, and spillway structures (IBCE, 1906).

Figure 9. Sibert’s principal assistants in the Atlantic Division included, from left: Major George H. Hoffman; Edward C. Sherman, and Caleb M. Saville (Jackson & Son, 1911).

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Figure 10. Schematic section through proposed Gatun Dam, from Saville’s report of August 30, 1908. Note three sheet-pile cuts associated with the rockfill containment dikes (Egineering and Contracting, 1914).

(Saville, 1908). Saville’s biggest challenge was to find sufficient quantities of suitable materials for the twin embankments, which focused on two borrow areas: one in the Gatuncillo Valley above the dam site, and the other in the Chagres Valley just north of the dam site, which appeared to have 20 million yd3 (15.28 million m3 ) of suitable material that could be excavated with suction dredges. If these sites proved unable to supply sufficient fill, then Limon Bay would be dredged, but this was 6 to 8 mi (9.6 to 12.9 km) distant, so it would cost more.

The Minority Report had assumed most of the puddled hydraulic fill would come from the Chagres Valley immediately upstream of the dam. Both of Saville’s borrow sites were within 1.5 mi (2.4 km) of the embankments, which was considered feasible, until the planned embankment reached an elevation of 45 ft (13.7 m) above sea level. At that time, the expectation was that centrifugal pumps would be employed on both embankments to provide additional lift for sluicing dredge tailings in the core pools.

The most important portion of Saville’s studies was the question of hydraulic conductivity of the sediments filling the two deeply incised channels of the Chagres, beneath the embankments (Figure 10). Diamond drill and wash borings were employed to ascertain the stratigraphy of the valley fills, and test pits were employed to gain some idea of the material permeability. Saville also supervised a series of 1:12 scale model tests to examine the likely position of the zone of saturation through the embankments (Saville, 1908, 1924).

The most significant revelation was that the upper 50 ft (15.2 m) section of channel fill contained an average 21 percent clay, and that between −50 and −100 ft (−15.2 to −30.5 m) depth, there was “blue clay containing little sand and numerous shells.” Below the blue clay, there were interstratified horizons of sand, gravel, and boulders that were of a “consolidated nature” and “cemented together with clay and silt.”

Most importantly, the deep borings and downhole permeability tests confirmed that no continuous layer of

loose sand or gravel was disclosed. Saville’s opinion was that the discovery of relatively low permeability materials and the absence of semi-continuous permeable strata made Gatun a suitable site for an embankment dam.

In his report of August 29, 1908 (Saville, 1908), he dispensed with the Minority Report’s recommendation for deep keyways under the twin embankments, opting instead for three lines of sheet-pile seepage cutoffs, shown in Figure 10. The first two sheet piles were 40 ft (12.2 m) deep and spaced 250 ft (76.2 m) apart, from a position beneath the upstream rockfill containment dike (select rockfill in Figure 10), and the third set was about 30 ft (9.1 m) deep, just beyond the lower, downstream containment dike. Seville recommended that shallow keyways be employed in the weathered bedrock (Gatun Formation) supporting the spillway structure, where test pits had revealed the presence of open joints.

Gatun Reservoir The kingpin structure of the American canal scheme was Gatun Dam and Lake Gatun, which retained the aggregate flow of the Chagres River and all its principal tributaries. Upon this structure rested the entire scheme of the locked canal, because it created a year-round reservoir pool 85 ft (25.9 m) above sea level, extending through the continental divide to the Pedro Miguel Lock. This reservoir that would allow ships to pass from ocean to ocean using a three-lock lift situated next to the dam (Figure 11).

Everything about the Gatun Dam was enormous.

Its proposed dimensions were without precedent at the time: 8,200 ft (2,500 m) long and a maximum width that was subsequently enlarged to 2,300 ft (701 m). With a reservoir pool elevation of 85 ft (25.9 m) above sea level, it could store sufficient water to impound a reservoir covering up to 164 mi2 (425 km2 ). At its center would sit the most critical structure, a mass concrete gravity dam gated spillway, able to discharge up to 222,000 cfs (6,216 m3 /s) from the largest man-made lake in the world (Figure 11).

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Figure 11. Map showing the areal extent of Gatun Lake, which inundated 164 mi2 (425 km2 ) of land area. In 1933–1935, Madden Dam was constructed on the upper Chagres River at Alhajuela, at the right margin of this map. The American-administered Canal Zone was 10 mi (16 km) wide, except along Gatun Lake’s shoreline. The Canal Zone was turned over to Panama between 1979 and 1999 (National Academy of Sciences, 1924).

Concerns about Settlement over the Incised Channels The biggest design challenge with the site was the underlying geology, which included deeply incised paleochannels, shown in Figure 12. The massive embankment would be placed upon these channels, which were as much as 258 ft (78.6 m) below sea level. There were some pervious horizons composed of silty and clayey sands and gravel, but they were at considerable depth (Thompson, 1943).

The most pondered design considerations revolved around how best to handle subsurface seepage, the low bearing capacity of the surficial deposits, and the amount of differential settlement that might occur across the filled channels. The concrete spillway was to be founded on the bedrock rise between the two 8

channels, along with the hydroelectric power plant and outfall channel (Figure 12).

The engineers who prepared the Minority Report of the IBCE in February 1906 were of the opinion that a sheet-pile seepage cutoff, like that used on the North Dike of the Wachusett Dam, would serve as an effective seepage cutoff structure at Gatun because the upper portion of the valley fill contained low-permability silt and clay. The differential settlement and bearing capacity issues went hand-inhand with one another. Though several decades before soil mechanics began to evolve as an analytical tool, the minority board recommended that gradual side slopes be utilized, so as to spread the dam’s bearing load over as large of an area as possible (Figure 13).

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Figure 12. Detail of subsurface conditions exposed along the centerline of Gatun Dam, where the Chagres River excavated incised channels into the mid-Miocene Gatun Formation (modified from Saville, 1916).

CONSTRUCTION CHALLENGES Work began on constructing the new townsite of Gatun 2 days after the U.S. Senate voted to approve the locked canal scheme, in late June 1906. Laborers began clearing and grubbing the dam site, which ran 1.5 mi (2.4 km) between two low hills. This clearing and grubbing took the better part of an entire year because of the thick tropical foliage.

Construction of two massive embankments on ether side of a bedrock knob began in early 1908, a few months before the reorganization of the project into three geographic divisions (described previously). This filling began with the construction of upstream and downstream “containment dikes,” common in that era for hydraulic fill embankments. The rockfill dikes were placed by dumping riprap from wooden trestles using side-dumping railcars. The dikes were originally spaced about 1,600 ft (488 m) apart, so the dam would be 15 times as wide as it was high (and adding 100 ft [30.5 m] for the crest width).

Four dredges, 10 steam shovels, and 10 muck trains were assigned to the placement of fill for the dam (Figures 14 and 15). The fill material came from a number of sources: the excavations for the Gatun Locks, dredge spoils from the 3-mi-long (4.8-km-long) channel between Gatun and Limon Bay, and from the Culebra Cut, up to 23 mi (37 km) to the south. It was a long trip, but about a dozen trainloads per day of fill were diverted from the waste dumps at Tabernilla and taken onto the dam (Figure 16).

The dam’s cross section was modified several times as the project evolved, to provide greater flexibility in the quarrying of rockfill and muddy spoils excavated by

suction dredges. It turned out that throughout 1907– 1908, much more fill material was being made available every month because of increased efficiency in excavation and rail and dredge transport (railborne spoils came out of the Culebra Cut, while dredged tailings came from the Gatun Locks approach channel).

Bearing Failures and Landslides During November 1908, the upstream containment dikes were raised rapidly in preparation for hydraulic filling. These operations progressed smoothly until November 20th, when the rock dike along the upstream core-to-shell transition of the eastern embankment suddenly settled, dropping as much as 60 ft (18.3 m) over a zone about 300 ft (91.4 m) wide, destroying the rail trestle (Figures 17–20).

Hoffman’s crews had been dumping rock in that area over the previous 10 days, to build up the containment dike. This work had coincided with a period of intense rainfall, when 28 in. (71 cm) of rain fell in just 24 days, which was believed to have played a role in the failure. It is not likely that the rainfall triggered the bearing capacity failure: The rockfill surcharge had been placed on soft unconsolidated clay muck infilling the old French Canal (Figure 18), and the rockfill was placed much too quickly to allow for adequate dissipation of the pore-water pressure developed in the clay/silt.

SPECIAL BOARD OF CONSULTING ENGINEERS On November 25, 1908, a few American newspapers ran a front page story stating that the “Gatun

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Figure 13. Proposed layout of Gatun Dam, spillway, and triple locks, from the International Board of Consulting Engineers Minority Report in February 1906 (IBCE, 1906).

Dam fails,” with ominous-appearing photographs of water rapidly running across the wrecked trestle. Seeking to assuage fears about the project encountering unforeseen dangers, President Roosevelt asked Presidentelect William Howard Taft to go down to Panama and personally investigate the situation. A “Special Board of Consulting Engineers” was appointed to accompany Taft, which included: Frederick P. Stearns and Isham Randolph, who had previously served on the IBCE in 1905–1906; James Dix Schuyler, an expert on hydraulic fill embankment dams in California; Allen Hazen, a consulting engineer in hydraulics and hydrol10

ogy from New York City; John R. Freeman, a consulting waterworks engineer in Boston; Arthur Powell Davis, former hydrographer with the U.S. Geological Survey; and Henry A. Allen, a consulting engineer in Chicago who had designed the pumping stations for the Chicago Sanitary District. It was a well-rounded and experienced group of engineers, with considerable technical expertise and construction experience with embankment dams.

In January 1909, Taft and his board of consultants (Figure 21) took a steamer down to Panama to assess the situation at Gatun Dam. By the time of their

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Figure 14. One of the 20 in. (50 cm) suction dredges working the Chagres River bottom at Gatun in 1908 (courtesy of National Archives and Records Administration).

Figure 17. View looking southwest, along the elevated rockfill containment dike 1 day after the bearing failure. Note presence of old French canal at extreme right, which passed beneath the dike (modified from the A. B. Nichols Collection of the Linda Hall Library).

arrival, the breach had been repaired, but they were shown photographs and technical data documenting the “landslide.” They correctly assessed that the soft clays comprising the foundation were likely of low permeability and susceptible to “squeezing” and “plastic flow” when surcharged with significant loads, which they believed to have been the cause of the localized settlement/landslide problems.

The group was impressed by the conservative nature of Gatun Dam’s main embankment cross sections, which sloped from 4:1 (near the dam’s crest) to 25:1

Figure 15. Suction dredge spoils being discharged into one of the core pools of Gatun Dam from a bounding dike, as seen in July 1911. Note train in distance and a hand-written annotation about “balls of clay” (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

Figure 16. Muck train using side-dumping cars to place training dike for hydraulic filling of the upstream side of Gatun Dam in 1911 (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

Figure 18. Plan view showing location of the November 1908 bearing failure, along the old French canal excavation, which had been allowed to fill with muck over several decades (modified from Wegmann, 1917).

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Figure 19. View of differential settlement of the rock containment dike at Gatun Dam on November 21, 1908, looking southerly, from the old French canal (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

(horizontal to vertical), along most of the dam’s downstream mass, a flat slope without precedent at that time.

In their report of March 3, 1909, the Special Board of Consulting Engineers (SBCE, 1909) recommended that the dam’s crest elevation be lowered from 135 to 115 ft (35 m), or 30 ft (9.1 m) above the normal operating pool, and that government forces could dispense with the driving of a sheet-pile cutoff wall beneath the dam’s central core because the underlying materials were of sufficiently low permeability. The board also recommended that the narrow keyway beneath the dam should be infilled with puddled clay (Figure 22).

They also recommended increasing the downstream slope from 25:1 to 16:1 (horizontal to vertical). The special board’s recommendations were summarized in the April 1, 1909, issue of Engineering News, with a comparison of the sections initially proposed in 1906; that prepared by ICC Assistant Engineer Caleb Saville in August 1908 (Figure 10), which was guiding con-

Figure 21. The Special Board of Consulting Engineers of the Isthmian Canal Commission pictured in early 1909. From left: William H. Taft, George W. Goethals, Frederick P. Stearns, James D. Schuyler, Allen Hazen, Isham Randolph, Henry A. Allen, John R. Freeman, and Arthur Powell Davis (courtesy of Autoridad del Canal de Panama).

struction when the slump occurred; and the section recommended by the Special Board of Consulting Engineers (Figure 23).

The board’s recommendations included broadening the rock dike to spread its load over a larger area to reduce the bearing pressure (Figure 23). Although a sound recommendation, the actual trigger for the bearing failure was the build up of pore-water pressure within the clay, because the rock surcharge was placed too quickly (it might have taken many weeks or months to equilibrate elevated pore-water pressures in a material composed of >20 percent clay).

Figure 20. Schematic sections through the bearing failure of November 1908, which occurred where the rockfill containment dike was founded on 20 ft (6.1 m) of muck infilling the old French canal. Note the bounding sheet-pile cutoff walls, driven across the channels beneath the eastern embankment.

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Figure 22. Puddled ditch keyway beneath the western embankment of Gatun Dam, as seen in August 1910. The Special Board of Consulting Engineers recommended that this narrow cutoff trench be continued beneath the two main embankments (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

Hydraulic Filling Hydraulic filling of the two embankment core pools began in March 1909 on the eastern embankment (Figures 23 and 24) and continued more or less without major interruptions until September 1912 (Figures 23 and 25). During this period, approximately 10,728,965 yd3 (8,196,929 m3 ) of hydraulic fill were placed by suction dredges. Dry filling of the dam began in 1907 and was completed on May 20, 1914, with a total placement volume of 12,195,017 yd3 (9,316,993 m3 ).

More Bearing Failures and Slides In late April–early May 1910, several slides occurred near the toes of the west embankment, soon after the west diversion channels were plugged with rockfill on April 25th. The trestle and rockfill containment dike (“northern diversion dam”) had been accomplished using timber piles driven 30 to 40 ft (9.1 to 12.2 m) into the channel bed (Figure 26).

The trestle and dike began moving laterally when the core pool rose to 15 ft (4.6 m) above sea level.

This sudden movement allowed the core pool to drain through the opening, which quickly widened, eroding

Figure 24. July 1911 view of the rising core pool of Gatun Dam’s east embankment, as seen from the Atlantic Division office at Gatun, looking across the excavations for the Gatun Locks (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

between 10,000 yd3 (7,640 m3 ) and 15,000 yd3 (11,460 m3 ) of material in just 15 minutes (Canal Record, 1910).

In October 1911, a roughly 1,000-ft-long (305-mlong) section of the western embankment “began heaving upward and outward as much as 14 ft (4.3 m) in places,” as described in the annual report of the ICC released on November 1912 (Engineering News Record, 1912). These movements were generally ascribed to poor drainage of the fine-grained soils within the core pools. Remedial measures included buttressing the containment dikes, draining the core pool, and allowing time for additional drainage of the sediment. A similar, but slightly smaller series of slumps occurred in April 1912.

On August 29, 1912, yet another series of slides occurred in the main embankment, between the locks and the spillway. By this time the east embankment had risen to an elevation of 101 ft (30.8 m), and the west end of the embankment crest gave way, slumping 10 to 15 vertical feet (3 to 4.6 m) across 750 ft (229 m) of the dam’s crest. Several images of these bearing failures are presented in Figures 27 and 28.

This failure occurred on a slope of 4:1 (horizontal to vertical). The upstream face inclination was then lowered to 7.67:1 (horizontal to vertical), as shown in Figure 29. The “random fill” sections often included large blocks of rock blasted out of the Culebra Cut or carried downslope by landslides. These often forced

Figure 23. Safe cross section for Gatun Dam with a crest elevation of 115 ft (35 m), suggested by the Special Board of Consulting Engineers in their report of March 1909 (ENR, 1909).

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Figure 25. Similar view of Gatun Dam’s east embankment, as seen on March 9, 1913, after hydraulic filling had ceased, when the embankments reached an elevation of ∼95 ft (29 m). By this time, the reservoir had filled to 45 ft (13.7 m) (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

“mud waves” like those shown in Figure 28, which were localized bearing capacity failures.

Design Modifications After April 1, 1907, the ICC was dominated by military engineers, who were concerned about the possibility of sabbotage of the dam. By increasing the height 50 ft and broadening the crest to a width of 100 ft, sabbatours would have to excavate an outlet trench more than 500 ft (152.4 m) long and at least 60 ft (18.3 m) deep, through a 16-ft-thick (4.9-m-thick) layer of riprap across the crest and down the upstream face (Figure 29). Such activity would require a considerable volume of excavation, likely requiring several weeks (Sibert and Stevens, 1915). In March 1909, the Special Board of Consulting Engineers recommended that the crest ele-

Figure 26. Remains of the trestle used to construct the rockfill containment dike (“north diversion dam”) across the west diversion channel beneath the western embankment. Timber piles driven 40 ft (12.2 m) into the channel bed were laterally displaced, allowing the core pool to drain and eroding considerable material (Sibert and Stevens, 1915).

vation be lowered from 135 ft (41.2 m) to 115 ft (35 m) above sea level.

In February 1912, Colonel Sibert asked the ICC to allow the crest to be taken down to 105 ft (32 m), and to reduce the volume of riprap cover to just 3 ft (0.91 m) along the upstream face of the two embankments, which is how it was completed (Figure 30). This reduced the embankment weight, because the soft fill within the paleo-channels was more compressible than originally assumed (Sibert and Stevens, 1915). When the embankment was topped off with rolled fill in March 1914, it was taken to an elevation of 108 ft (32.9 m) above sea level, as an allowance for up to 3 ft (0.9 m) of future settlement (Canal Record, March 25, 1914).

Figure 27. A series of bearing capacity failures occurred in 1912 along ∼750 lineal feet (229 m) of the embankment crest (from Sibert and Stevens, 1915).

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Water slowly began to build up behind the new dam.

By the time the embankment was topped off (Figure 31), it contained 22,923,982 yd3 (17,513,922 m3 ) of fill, a world record until the completion of Fort Peck Dam in 1940.

GATUN SPILLWAY AND OUTFALL CHANNEL

Figure 28. “Mud waves” were local bearing-capacity failures that occurred when excessive volumes of fill were dumped from fill trestles to construct containment dikes (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

In the end, Sibert’s engineers decided to construct an even more conservative cross section than that recommended by the Special Board of Consulting Engineers, because they had excess fill to waste from all of the excavations for the canal, plus the landslides plaguing the Culebra Cut. What began as a 3:1 (horizontal to vertical) slope of 18.1 degrees was winowed down to a 11:1 slope of just 5.2 degrees (Figure 30), which was later dropped to 16:1, or just 3.6 degrees. This flattening of the side slopes had the added benefit of lengthening “seepage paths” through and around the dam and its foundation, lowering the potential for destablizing hydraulic uplift or high seepage pressures. These sorts of field observations and the resulting modifications to the design became commonplace many decades later, when Karl Terzaghi and Ralph Peck perfected what came to be known as the “observational method” (Peck, 1969).

The west diversion channel was closed in late April 1912, before the main embankment had been completed, employing fill dumped from a trestle. There had been some concern about whether the deeply incised channel would be able to sustain the surcharge of the diversion embankment, given how quickly it was placed.

The Gatun spillway is an arched concrete gravity structure up to 100 ft (30.5 m) high with a crest length of 808 ft (246.3 m). It was conceived by Isham Randolph and Frederic P. Stearns of the IBCE in February 1906.

The design of the 30 × 30 ft (9.1 × 9.1 m) Stoney Gates assumed the spillway sills would normally be submerged 16 ft (4.9 m) below the operating pool. The size of the gates were within 12 in. (30.5 cm) of those employed on the Chicago Sanitary and Ship Canal, overseen by Isham Randolph.

Caleb Saville was assigned to solve this problem because he had worked on the Wachusett Dam. Saville surveyed the literature and performed the calculations.

He designed reinforced concrete impact blocks 9 ft (2.7 m) high at the base of the spillway skirt, to hasten energy dissipation. Theorems were not yet in existence to deal with dynamic forces, fatigue failure, or resonant frequency of vibration. Saville decided to place steel impact jackets around the upstream faces of the energy dissipator blocks, or “baffles,” and these have served their intended purpose fairly well, considering their design was without precedence (Isthmian Canal Commission, 1908; Saville, 1924).

Upon completion of his report on Gatun Dam and spillway in July 1908, Saville was named special assistant engineer in the office of the Chief Engineer George W. Goethals in Culebra, where he supervised investigations of meteorology, hydrology, and the water supply systems. He was replaced by Edward C. Sherman, from Kingston, MA, who had been working under Lieutenant Colonel Hodges. In August 1908, Sherman was placed in charge of designing spillways for Gatun and Miraflores Dams, as well as other technical aspects of these mass concrete structures. He supervised the hydraulic model tests of the straight and curved

Figure 29. The upstream face and crest of Gatun Dam were paved with a 10-ft-thick (3-m-thick) layer of basalt riprap, shown as the dark stippled pattern. This required 175,000 yd3 (133,700 m3 ) of material placed between March and December 1913 (Engineering & Contracting, 1914).

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Figure 30. As-built section through Gatun Dam, with a crest height of 105 ft (32 m), taken from Siebert and Stevens (1915). In total, 10,728,965 yd3 (8,196,929 m3 ) of hydraulic fill and 12,195,017 yd3 (9,316,993 m3 ) of “dry fill” were used to construct the massive embankments.

spillway structures, shown in Figure 32. He was one of the youngest engineers in responsible charge of a design team, signing the final plans (shown in Figures 33–37).

At normal pool elevation of 85 ft (25.9 m), the spillway was designed to release 140,000 cfs (3,920 m3 /s), close to the maximum observed flow of the Chagres River when the canal was under construction (John Frank Stevens had estimated the peak flow of the Chagres to be about 150,000 cfs [4,200 m3 /s] in 1907). This design allowed the gates to be lifted and the water to spill from a crest elevation 16 ft (4.9 m) lower than the normal operating pool, which would allow greater flood storage, should the need ever arise.

The system was hydraulically efficient. The only problem was impact force on the baffle blocks, and the 43 ft (13.1 m) drop was much higher than in Chicago.

With an additional 2 ft (0.6 m) of head (elev. 87 ft [26.5 m]), the spillway could pass about 182,000 cfs (5,096 m3 /s), and 222,000 cfs (6,216 m3 /s) with 9.3 ft (2.83 m) of excess head (elev. 94.3 ft [28.7 m]).

Construction of the Spillway

Figure 31. Aerial oblique view of Gatun Dam, looking northeast. The spillway structure lies between two massive embankments (courtesy of Autoridad del Canal de Panama).

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In early 1909, the first concrete was poured for the spillway, and this work was completed 19 months later, after which the Stoney Gates were set into place, which took another year and a half. During the next year, the discharge of the Chagres was allowed to build up behind the new dam and spillway. The rising waters of Gatun Reservoir eventually allowed testing of the 14 Stoney Gates along the spillway, during the second week of June 1913 (Figure 38).

Much to everyone’s relief, they performed flawlessly.

On June 27th, when the reservoir elevation had risen to

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Figure 32. 1:32 scale hydraulic models of the Gatun spillway, supervised by George Hoffman in 1909–1910. The straight spillway shown at right was recommended by the International Board of Consulting Engineers and Colonel Sibert. The arched spillway shown at left was built with only a single row of baffles, but it proved to be more hydraulically efficient (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

48 ft (14.6 m), the last of the Stoney Gates was closed, allowing the lake to rise to its operating pool elevation of 85 ft (25.9 m).

The flood-of-record occurred in December 2010 (Figure 39), raising the reservoir to 88.36 ft (26.9 m), allowing spillage of 198,000 cfs (5,544 m3 /s), and hastened the opening of the culverts in the Gatun Locks, which conveyed an additional 27,000 cfs (756 m3 /s).

The short- and long-term impacts of the 2010 La Purisima flood were described by Alfaro et al. (2014).

SITE GEOLOGY The earliest geologic reconnaissnace of the Isthmus of Panama was in 1897–1898 by Robert T. Hill of the U.S. Geological Survey (USGS), working for the

Nicaraguan Canal Commission chaired by Admiral James G. Walker, which examined the possible routes across Central America (Hill, 1898). At the time of Panama Canal’s construction (1904–1914), there had been little work undertaken to critically examine the geologic characteristics of the units encoutered, until landslides began to plague the project (Rogers, 2014).

The stratigraphic sequence of the Gatun Reservoir area was initially described by Donald F. MacDonald of the USGS, working for the ICC (MacDonald, 1913, 1915; Rogers, 2014). The formations at the Gatun Dam site range in age from Upper Eocene to Holocene, lying upon a basment complex of pre-Tertiary age. The precise stratigraphy of these formations was examined in much more detail for the Third Locks Project in 1939– 1942 (Thompson, 1943).

Figure 33. Elevation view of the Gatun spillway and outfall channel, which is lined with concrete (Engineering and Contracting, 1914).

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Figure 34. Sections through Gatun spillway sills, which rose to an elevation of 69 ft (21 m) above sea level (16 ft [4.9 m] above the spillway sills). The fixed piers were spaced 45 ft (13.7 m) apart, supporting the Stoney Gates, which controlled discharge from 14 bays (courtesy of the A. B. Nichols Collection of the Linda Hall Library).

Unconsolidated Surficial Materials Thompson (1943) divided the unconsolidated channel fill into four phases filling the incised channels passing beneath Gatun Dam, with the lower phase being gray to blue-gray silty clay. The next facies is a clay deposited in brackish marine water, with abundant mollusk shells in a black organic silt. This is covered by a littoral swamp unit containing abundant wood and organc matter mixed with silt, and it is usually of a brown to black color. The uppermost facies was observd to be a light gray plastic clay of very low shear strength.

Thompson felt that the four phases intergraded, with abundant sand lenses locally, but was seldom continuous. These four facies are found throughout the floodplains of the lower Chagres River and its tributaries in vicinity of Gatun Lake. The top and bottom phases are the thinnest, but they are also the most widepread. The brackish marine phase thickens towards the ocean, while the littoral unit progressively thins in the same direction. These stratigraphic relationships sug-

Figure 36. Plan view of the Gatun Dam as constructed between 1907 and 1913. The gravity arch concrete spillway was constructed on the middle on a bedrock knob, while Gatun locks rested on the dam’s right abutment (Wegmann, 1917).

gest that the four facies were likely deposited during the Holocene, as sea level rose to its present level.

Gatun Formation

Figure 35. Final design for the Gatun Dam spillway, dated January 15, 1910. With a crest length of 808 ft (246.3 m), it was founded on a resistant knob of brown conglomerate between the deeply incised paleo-channels (Engineering and Contracting, 1914).

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Stewart McReddie Jones (1947) was a graduate student at Oregon State College (before and after World War II) when he mapped the geology of Gatun Reservoir and vicinity, summarized in Jones (1950). While visiting Panama he worked with the Canal Zone’s

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Gatun Dam—Megastructure of the Panama Canal

Figure 37. Panorama of the concrete spillway structure at Gatun Dam, situated between the massive earthen embankments. Two rows of baffle blocks that were installed on the spillway skirt dissipate the energy of the flowing water and hasten the transition to less turbid flow in the strait outfall channel (courtesy of National Archives and Records Administration).

Engineering Geologist H. G. Martin and Professors Donald F. MacDonald and W.P. Woodring (between 1941 and 1947). A portion of Jones’ expansive geologic map is reproduced in Figure 40.

Jones’ map shows the dam site to be underlain by unconsolidated deposits laid down by the Chagres River, with the predominant bedrock beneath the spillway belonging to the Gatun Formation of Middle Miocene age. Jones mapping suggested that the Gatun Formation was composed of sandstone, siltstone, conglomer-

Figure 39. View from Gatun spillway during the La Purisima flooding of December 7–10, 2010, which raised Gatun reservoir 3.36 ft (1.02 m), allowing a record overflow of 198,000 cfs (5,544 m3 /s) (courtesy of Autoridad del Canal de Panama).

ate, and volcanic tuff, all massively bedded. The members were observed to be “variably marly and tuffacious, highly fossiliferous,” and perturbed by regular systematic “joints, spaced every 20 feet.” Jones mapped extensive exposures in the walls and floor of the Third Locks excavations at Gatun, which exposed a section 546 ft (166.4 m) thick. Correlations with other outcrops in the vicinity suggested a minimum thickness of 1,400 ft (426.7 m), and “probably much more” (Jones assumed a thickness of 3,000 ft [914.4 m]). The beds dipped north to northwesterly, with inclinations as little as 2 degrees near Limon Bay, and upwards of 20 degrees southeast of Limon Bay. The depth of weathering within the formation was reported to average about 30 ft (9.1 m) (Jones, 1950, p. 902).

CONCLUSIONS

Figure 38. Aerial oblique view of the Gatun spillway in operation some years after the project’s completion. The concrete training walls of the outfall channel serve to protect the dam’s embankments.

The large structure at left center is the hydroelectric power station, which provides electricity to run the Gatun locks and Stoney Gates (courtesy of the U.S. Army Corps of Engineers).

Gatun was home to the Panama Canal’s key structural elements: the Gatun locks, Gatun Dam, spillway, outfall channel, and powerhouse. If any of these structures failed to perform as intended, the canal project would not have functioned properly. Gatun Dam was the mega-earth structure of its era (1906–1914), which had a significant impact on the subsequent evolution of embankment dam technology (Wegmann, 1917; Justin, 1932).

Gatun Dam and spillway included multiple layers of detailed professional peer review; the use of scaled physical models to evaluate through seepage and under seepage and spillway hydraulics; multiple slope stability and bearing failures during construction; and design changes during construction, which were carried out in light of new information and increased appreciation of the geotechnical conditions, half a century before

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Rogers and Barrelier

Figure 40. Excerpt of the Geologic Map of Gatun Lake and Vicinity by Jones (1950), focusing on the area of the dam, locks, and spillway structure (near the circled “4” on the map). Note the criss-crossing faults, several of which appear to be potentially active.

the observational method became routine practice in geotechnical engineering.

The ICC drew upon the combined expertise of the best and brightest engineering minds of the era, so the design of Gatun Dam was essentially a “team effort.”

That team included some of the most experienced waterworks engineers of that era, such as: Frederick T. Stearns, John R. Freeman, James Dix Schuyler, Allen Hazen, and Arthur Powell Davis. Many of the structural elements, such as the Stoney Gates on the spillway, had previously been employed by Isham Randolph and Henry A. Allen on the Chicago Sanitary & Ship Canal.

Other facets, like the energy dissipation blocks in the spillway skirt, were new innovations, evaluated through scale model tests employing dynamic similitude.

The most important decisions were of a geotechnical nature, dealt with almost three decades before the introduction of modern soil mechanics in the late 1920s. The engineers realized that the surcharge load of the embankments would result in some differential settlement over the buried paleo-channels, but they 20

judged that this strain was acceptable if spread over a sufficiently broad area (the 16:1 slope of the dam’s downstream face would not be exceeded until Waco Dam was retrofitted with 20:1 slopes in 1964–1965).

To date (100 years later), the embankments have settled up to 6.5 ft (2 m) where they cross the two deeply incised paleo-channels, without any overt transverse seepage complications.

The twin embankments were about half hydraulic fill and half rolled fill (Figure 30). Gatun’s embankment volume of 23 million yd3 (17.6 million m3 ) was a world record until the construction of Fort Peck Dam by the U.S. Army Corps of Engineers in the late 1930s (126 million yd3 [96.3 million m3 ]).

The Gatun triple locks structure was socketed into the underlying bedrock of the Gatun Formation and formed the dam’s right abutment. Most of the rocky fill for the embankment toe fills came from the lock excavations in the Gatun Formation. Large volumes of “surplus fill” came from landslides in the Culebra Cut, which began plaguing the project in 1912 (Rogers,

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2014). The spillway structure was founded on a bedrock knob between the incised paleo-channels of the Chagres River. This allowed a stable foundation and a secure outfall channel, equipped with a 3-mi-long (4.8km-long) training dike to convey runoff away from the locks.

All of these elements were constructed with considerable efficiency and cost savings. The final cost of Gatun Dam was US$ (1910) 8 million, US$ (1910) 2 million below budget. These savings were in stark contrast to their counterparts on the Pacific side. The Miraflores and Pedro Miguel Locks cost more to construct and to operate and maintain than the triple locks at Gatun (Sibert and Stevens, 1915).

Having been constructed under the direction of U.S. Army Corps of Engineers officers and civil service engineers, the canal project catapulted the Corps into preeminence as America’s premier designer and constructor of earthen dams for the next seven decades, which witnessed the construction of more embankment dams than any other type.

ACKNOWLEDGMENTS AND PICTURE CREDITS The senior author was stationed at Rodman Naval Station in the Canal Zone as a naval intelligence officer, where he was graciously hosted by the geotechnical engineers and geologists of the Panama Canal Commission and Panama Canal Authority.

Both authors are indebted to the staff of the Panama Canal Commission Library and Technical Resources Center (prior to 1999), and the Autoridad del Canal de Panama (or ACP, after 1999), who supplied countless photos and records. Photograph sources include: the A. B. Nichols Collection of the Linda Hall Library (NC-LHL), the National Archives and Records Administration (NARA), International Board of Consulting Engineers (IBCE), Library of Congress (LOC), the serial publication Engineering & Contracting (E & C), and the U.S. Army Corps of Engineers (USACE). The Canal Record cited in the text was a weekly newspaper published by the Isthmian Canal Commission (ICC) under the direction of ICC Secretary Joseph Bucklin Bishop, beginning in September 1907. It continued until the end of the project, in the late summer of 1914. These are scanned and archived in the University of Florida Digital Collections (http://ufdc.ufl.edu/UF00097368/00002).

REFERENCES ALFARO, L. D.; BARRELIER, M. H.; AND DE PUY, M., 2014, Gatun Dam history and developments. In Dennis, B. G., Jr. (Editor), ENGINEERING THE PANAMA CANAL: A CENTENNIAL RETROSPECTIVE: ASCE Press, Washington, DC. pp. 367–383.

BATES, L. W., 1906, The Crisis at Panama: L. W. Bates, New York.

ENGINEERING AND CONTRACTING, 1914, Lakes Gatun and Miraflores Dams and spillways: Engineering & Contracting, Vol. 41, No. 1, pp. 33–40.

ENGINEERING NEWS, 1909, Concerning the Gatun Dam and earth dams in general: Engineering News, Vol. 61, No. 13, p. 354–358.

ENGINEERING NEWS RECORD, 1912, Misbehavior of the Gatun Dam: Engineering News Record, Vol. 66, No. 21, pp. 562–563.

HILL, R. T., 1898, The Geological History of the Isthmus of Panama and Portions of Costa Rica: Bulletin 28, Harvard College Museum Comparative Zoology, Cambridge, MA, 285 p., 19 plates.

INTERNATIONAL BOARD OF CONSULTING ENGINEERS, 1906, Report of Board of Consulting Engineers for the Panama Canal: U.S. Government Printing Office, Washington, D.C., 426 p.

ISTHMIAN CANAL COMMISSION, 1908, Minutes of Meetings of the Isthmian Canal Commission; March 1904 to September 1905 Inclusive: U.S. Government Printing Office, Washington, D.C., 324 p.

ISTHMIAN CANAL COMMISSION, 1914, Minutes of Meetings of the Isthmian Canal Commission and of its Executive and Engineering Committees; April 1905–March 1914: U.S. Government Printing Office, Washington, D.C., 349 p.

JACKSON, F. E. & SON, 1911, The Makers of the Panama Canal and Representative Men of the Panama Republic: ChasmarWinchell Press, New York, 261 p.

JONES, S. M., 1947, Geology of Gatun Lake and Vicinity: unpublished thesis, Oregon State College, Corvallis, OR, 47 p., 2 plates.

JONES, S. M., 1950, Geology of Gatun Lake and vicinity, Panama: Geological Society of America Bulletin, Vol. 61, pp. 893–922.

JUSTIN, J. D., 1932, Earth Dam Projects: John Wiley & Sons, New York, 345 p.

MACDONALD, D. F., 1913, Geologic section of the Panama Canal Zone: Geological Society of America Abstract with Program, Vol. 24, pp. 707–711.

MACDONALD, D. F., 1915, Some Engineering Problems of the Panama Canal in the Relation to Geology and Topography: Bulletin 86, U.S. Bureau of Mines.

MALTBY, F. B., 1945, In at the start at Panama: Civil Engineering, Vol. 15 (June-September 1945), pp. 6–9.

MORISON, G. S., 1902, The Bohio Dam: ASCE Transactions, Vol. 47, pp. 235.

NATIONAL ACADEMY OF SCIENCES, 1924, Report of the Committee of the National Academy of Sciences on Panama Canal Slides: Vol. XVIII, National Academy of Sciences, Washington, D.C. PECK, R. B., 1969, Advantages and limitations of the observational method in applied soil mechanics: Geotechnique, Vol. 19, pp.

171–187.

ROGERS, J. D., 2014, The American engineers who built the Panama Canal. In Dennis, B. G., Jr. (Editor), Engineering the Panama Canal: A Centennial Retrospective: ASCE Press, Washington, DC. pp. 112–349.

SAVILLE, C. M., 1908, Gatun Dam investigations, Appendix E. In Annual Report of the Isthmian Canal Commission: U.S. Government Printing Office, Washington, D.C. (August 29, 1908), pp. 127–196, plates 62–173.

SAVILLE, C. M., 1916, Dam and reservoir foundations: Engineering News Record, Vol. 75, pp. 1229.

SAVILLE, C. M., 1924, Discussion on the design of earth dams: ASCE Transactions, Vol. 87, pp. 94–103.

SIBERT, W. L. AND STEVENS, J. F., 1915, The Construction of the Panama Canal: D. Appleton & Co., New York.

SORZANO, J. F., 1910, Water supply for the lock canal in Panama: ASCE Transactions, Vol. 67, pp. 61–77.

STEARNS, F. P., 1902, Discussion on the Bohio Dam: ASCE Transactions, Vol. 47, pp. 259–277.

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Rogers and Barrelier STEVENS, J. F., 1928, The Panama Canal: Address at the Annual Convention at Denver, Colorado: ASCE Transactions, Vol. 91, pp. 946.

THE CANAL RECORD, 1911, Engineers on Canal: Full Text of the Report by the Special Board of Engineers Submitted to Congress by the President, Vol. 11, No. 27, pp. 212–13.

THE CANAL RECORD, 1914, Grading East Wing of Gatun Dam, Vol. 7, No. 31 (March 25, 1914), pp. 289–290.

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THOMPSON, T. F., 1943, Final Report on Modified Third Locks Project, Part II Design, Chapter 3 Geology: Special Engineering Division, Panama Canal, Balboa, CZ. 33 p.

WARD, C. D., 1904, The Gatun Dam: ASCE Transactions, Vol. 53, pp. 36–44.

WEGMANN, E., 1917, The Design and Construction of Dams: John Wiley & Sons, New York.

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Seepage Investigation for Remedial Grouting, Crafton Hills Reservoir, California HOLLY J. NICHOLS∗ California Department of Water Resources, 3500 Industrial Boulevard, West Sacramento, CA 95691

Key Terms: Dams, Engineering Geology, Foundations, Site Investigations ABSTRACT Engineering geologists are often tasked with a dam or levee seepage problem that needs to be repaired, but it is not clear where the seepage pathway is, how fast the seepage is flowing, or how best to repair the problem area.

This article discusses a dam seepage area at the Crafton Hills Reservoir in San Bernardino County, CA. Upon refilling the enlarged reservoir, new seepage was observed in unexpected locations. Although the seepage amount appeared to be small, project managers decided to seal off the new seepage pathway utilizing subsurface grouting so the seepage would not become a long-term problem. In order to develop an effective remedial grouting program, additional subsurface information was needed. It was already known that the foundation materials are variable, likely have a variety of hydraulic conductivity values, and may respond differently to grouting depending on the pressures applied. Several methods were employed to determine the primary seepage pathway(s) and the range of pressures to use that would provide the most effective grout penetration. Although the geologic investigation led to some contradictory and unexpected results, the use of multiple testing methods provided a much better understanding of the foundation rock, which led to a more robust remedial grouting plan.

INTRODUCTION The Crafton Hills Reservoir is part of the existing East Branch Extension (EBX) of the State Water Project (SWP), which includes the California Aqueduct. The EBX is a cooperative water conveyance project between the Department of Water Resources (DWR), San Bernardino Valley Municipal Water District, and San Gorgonio Pass Water Agency to deliver State Water Project water to the eastern portion of the San Bernardino Valley and San Gorgonio Pass service areas. The EBX portion of the State Water Project con∗ Corresponding

author email: holly.nichols@water.ca.gov.

veys water from the DWR Devil Canyon facilities to Yucaipa in San Bernardino County and Cherry Valley in Riverside County through a series of pump stations, pipelines, and reservoirs.

The Crafton Hills Reservoir is located on the eastern margin of the Crafton Hills, approximately 2 mi north of the City of Yucaipa in San Bernardino County (Figure 1). The Crafton Hills Reservoir is impounded by two dams: the original dam completed in 2002 and the enlargement dam completed in 2014. The total capacity of the reservoir is about 225 acre-ft (277,533 m3 ). Although small, the Crafton Hills Reservoir is an important component of the EBX water conveyance system that helps regulate flows through the system.

GEOLOGIC SETTING The Crafton Hills are located in the northern portion of the Peninsular Ranges geomorphic province and are situated between the San Andreas and San Jacinto Fault zones in a right-stepping trans-tensional basin. The Crafton Hills represent a horst with grabens to the northwest (Mill Creek Drainage) and southeast (Yucaipa Valley).

The Crafton Hills Reservoir is located in the eastern part of the Crafton Hills (Figure 2). The site is bound by the Holocene-active Crafton Hills Fault to the southeast and the inactive, ancient Vincent Thrust to the north (Matti et al., 2003). The late Cretaceous to early Tertiary Vincent Thrust juxtaposes meta-granitic bedrock (hanging wall) with the Pelona Schist (footwall). The bedrock encountered at the dam sites is a Mesozoic sheared, mylonitic meta-granitic rock that is cut by Oligocene granodiorite dikes and Miocene mafic dikes (Morton and Miller, 2006).

Detailed geologic mapping of the foundation materials was completed during construction of the two dams.

The materials encountered in the foundation consisted of highly sheared, intensely fractured meta-granitic rock, blocky meta-granitic rock, sheared granodioritic dikes (referred to as “felsic dikes”), and fine-grained, massive mafic dikes, which is consistent with the geologic descriptions presented by Morton and Miller (2006).

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Figure 1. The Crafton Hills Reservoir is located in Southern California and is part of the East Branch Extension of the California State Water Project.

Rock joints were typically tight with planar to undulating surfaces and were closely to moderately spaced.

Bedrock shears were typically brittle (fractured), contained clay gouge (pulverized), or represented a combination of both types. Brittle shears typically ranged from hairline to 1/8 in. (0.32 cm) thick containing clay gouge or a slight clay film along the shear surface.

Pulverized shears typically ranged from 1/16 to 4 in.

(0.063–10.2 cm) thick and contained clay gouge greater than or equal to 1/16 in. (0.063 cm) thick.

SEEPAGE HISTORY The original Crafton Hills Dam is a 95-ft (30-m) tall earthen embankment with a single row, 50-foot (15-m)–deep grout curtain composed of neat cement grout (Nichols et al., 2014). Within about 11 days of first filling, a wet spot appeared at the left groin of the dam (Figure 3). DWR engineering geologists inspected the wet spot, conducted additional geologic mapping, and concluded that water from the reservoir was travelling along a through-going dike. The distance between the upstream and downstream toes of the dam along the dike is about 320 ft (98 m), resulting in an initial, 24

rough seepage velocity through the rock of about 29 feet (8.8 m) per day. Seepage observed at the left groin of the dam peaked within about 2 months and slowly decreased without the need for remediation. However, a wet spot at the left groin area has persisted since first filling, allowing the sustained growth of hydrophilic plants.

The Crafton Hills Enlargement Dam is a 75-ft (23-m)–tall earthen embankment with a dual-row, opposing-angle, 60-ft-deep grout curtain composed of stable, high-mobility grout (Nichols et al., 2014). The construction of the enlargement dam and reservoir required the excavation of about 430,000 cubic yards (∼330,000 m3 ) of meta-granitic rock, including removal of a ridge that would connect the two pools, referred to as the “connector channel” (Figure 4). The excavation required to create the connector channel exposed a large surface area of unsaturated fractured rock.

During first filling of the enlarged reservoir, no unexpected seepage was observed around the new enlargement dam. However, field engineers observed that bubbles were emanating from the rock in the connector channel for several days after it was submerged by

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Figure 2. The Crafton Hills are situated between the San Andreas and San Jacinto Fault zones and are expressed as a pop-up (horst) block bound on the north by the inactive Vincent Thrust and on the south by the Holocene-active Crafton Hills Fault Zone (Matti et al., 2003).

water. Within 2 days of beginning to fill the enlarged reservoir, the pre-existing seepage area at the left groin of the original dam, which had dried out while the reservoir was empty during construction, reappeared, suggesting a seepage velocity of about 160 ft (49 m) per day—nearly five times as fast as at first filling in 2002. The reappearance of the left groin seepage area was expected, and the apparent increase in seepage velocity was not unexpected. It is likely the rock under the dam remained wet during construction, and the increase in reservoir elevation simply applied pressure to remobilize the seepage water.

Within about 30 days of filling the enlarged reservoir, which maintained the same maximum pool elevation as the original reservoir, new, unanticipated wet spots and seepage areas appeared downstream of the origi-

nal dam toe (Figure 5). Some of the new wet spots (i.e., areas of saturated ground, but no flowing water) appeared to be related to the left groin seepage (Figure 5, Areas A, C, and G). One wet spot (Area F) was located near the seepage discharge conduit. Several small wet spots and seepage areas (Areas D, E, I, and H) were located in the axis of the drainage away from the dam and had uncertain origins. However, the first and most significant new seepage area, Area B, appeared after 30 days and is coincident with a different through-going dike than the one responsible for the left groin seepage (Area A on Figure 5). This other dike extends from the connector channel, through the right abutment of the original dam, and past the toe of the dam. The distance between the dike in the connector channel and the first new seepage area (Area B) is about 640 ft (195 m),

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Figure 3. Dam foundation mapping of the original Crafton Hills Dam (modified from Bruce and Glick, 2003). Blue areas are mapped dikes.

The left groin seepage appeared at the margins of a dike at the downstream toe within about 11 days after first filling of the reservoir in 2002, resulting in a rough seepage velocity of 29 ft (8.8 m) per day.

Figure 4. Oblique aerial view of the two Crafton Hills dams. The original dam is on the right; the enlargement dam is on the left. The ridge between the two pools was removed to form the connector channel.

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providing an initial, rough seepage velocity through the rock of about 21 ft (6.4 m) per day. This is similar to the initial seepage velocity observed in 2002 at the left groin wet spot (29 ft/8.8 m per day), suggesting similar subsurface conditions.

Additional new wet spots/seepage locations continued to appear downstream of the original dam for nearly 2 months. Many seepage locations appeared to be coincident with the connector channel dike, but other wet spots had more ambiguous origins. This led to the possibility that perhaps several factors were influencing the new seepage of water from the enlarged reservoir.

The total seepage downstream of the original dam appeared to be relatively small, collectively less than about 10 gallons (38 L) per minute. While this amount of seepage does not likely pose an imminent dam safety threat, there is the potential for long-term internal

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Figure 5. The seepage along the left groin reappeared after about 2 days of filling the enlarged reservoir (modified from Dean et al., 2016).

New seepage areas appeared downstream from the original Crafton Hills Dam after about 30 days.

erosion that could lead to sinkholes, collapse, or even catastrophic dam failure if left unattended. Because of this potential future risk, DWR project managers committed to stopping, or at least reducing, the new seepage observed downstream from the original dam.

Conceptually, the remediation would involve grouting the bedrock discontinuities. To ensure the remedial grouting would target the most appropriate location, the problem area(s) needed to be better defined, which required additional subsurface geologic investigations and testing.

3. Monitor response times to fluctuations in the reservoir; 4. Conduct hydraulic conductivity testing (packer testing); 5. Conduct falling head tests; and 6. Conduct a dye tracer test.

GEOLOGIC INVESTIGATION

Potential Seepage Pathways

The purposes of the geologic investigation were to (1) identify the seepage pathway(s) contributing to the new seepage areas observed at the toe of the original dam and (2) determine the hydraulic properties of the rock types observed under various pressure conditions.

In order to accomplish the objectives of the geologic investigation, the following phases of work were completed:

Prior to filling the enlarged reservoir, seepage downstream from the original dam was limited to the wet area at the left groin. After filling the enlarged reservoir, four potential seepage pathways were identified based on the observed wet spots and seepage areas.

The construction of the enlarged reservoir created at least two new potential seepage pathways downstream from the original dam:

1. Examine foundation geology maps and design features for potential seepage pathways; 2. Drill and install piezometers screened in the three main rock types: meta-granitic rock, mafic dike, and felsic dike;

1. Through or along the through-going dikes from the connector channel to the area downstream from the original dam. The connector channel dike is predominantly composed of the older, felsic material, with thin, cross-cutting stringers of the younger, more mafic rock; and

The following sections discuss the subsurface geologic investigation that was performed to help identify the seepage pathway(s) and ultimately incorporate that information into the development of a plan for remedial grouting.

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Figure 6. New piezometers were installed to evaluate the groundwater conditions in the foundation materials. Their locations were chosen to create groups of comparison piezometers: one to the left (north) of the dike, one in the dike, and one to the right (south) of the dike.

2. Through fractures in the meta-granitic rock in the connector ridge between the original and enlargement dams.

Seepage through fractures in the meta-granitic rock under the dam and at the right abutment was not considered to be significantly contributing to the new seepage areas because these areas had been exposed to water for over 10 years with no signs of seepage, aside from the left groin area.

Two pre-existing conditions were also identified that could potentially be contributing to the new seepage areas downstream from the original dam: 1. Along potentially dried/cracked lake sediments or foundation rock along the outlet pipeline due to a prolonged dry reservoir during construction of the enlargement dam; and 2. Through ungrouted fractures in the foundation rock of the original dam. Though an unlikely candidate, the use of neat cement in the grout curtain left some question as to its effectiveness, because it has a tendency to have high bleed, potentially leaving voids behind.

The first two new potential seepage pathways were considered to be the most likely, but limited investigations were included to confirm that the pre-existing conditions were not significantly contributing to the new seepage areas.

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Drilling and Piezometer Installation A total of 14 rock core borings were completed through the original dam crest, along the connector channel access road, and downstream from the original dam (Figure 6). The borings were intended to be completed as groups of piezometers, one to be screened within the felsic dike material and one on either side of the dike screened within the fractured meta-granitic host rock. The piezometer locations were selected based on cross-section analysis of the geologic mapping of the foundation rock. A total of 14 open tube piezometers were installed, and five vibrating wire piezometers (VWP) were used in three of the holes. Table 1 summarizes the hole completions.

The purpose in positioning the piezometers in groups was to be able to observe dynamic changes in groundwater levels. Typical reservoir operation brings the reservoir levels up at night, with slow discharge during the day. This fluctuation in water levels provides a window into how water may be moving through the different rock types.

The four borings through the dam crest were positioned to be about 10 ft (3 m) downstream from the grout curtain, in part to observe the presence or absence of grout in the rock core and to test the effectiveness of the grout curtain by water pressure testing. This position along the dam crest also avoided drilling through the chimney drain and drainage

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Seepage Investigation for Remedial Grouting Table 1. Summary of well completions.

Total Depth (ft bgs)

Well Screen Interval (ft bgs)

CHE-20

76.6

56–66

North gr

CHE-21 CHE-21A

77 76

46–51 55–65

North gr Mafic dike

CHE-22 CHE-23 CHE-24 CHE-25 CHE-26 CHE-27 CHE-28 CHE-29 CHE-30 CHE-30A CHE-31

76.5 51 51.5 25 26 25 13.4 51 51 46 51

56–66 28–48 28–48 18–23 17–23 8–23 8–13 28–48 28–48 19–39 28–48

South gr Felsic dike South gr Felsic dike South gr Mafic dike Pipe backfill North gr Dike (both) Mafic dike South gr

Well ID

Rock Type through Screen Interval

VWP (ft bgs)

Rock Type at VWP

Dye Observed

40 75 — 50 75 75 — — — — — — — — — —

Dam core North gr — Mafic dike South gr South gr — — — — — — — — — —

Yes Yes No Yes No No Yes No No No No No Yes Yes

North gr = meta-granitic rock left (north) of the dikes; South gr = meta-granitic rock right (south) of the dikes; bgs = below ground surface; VWP = vibrating wire piezometer.

blanket. Drilling through the dam required special drilling techniques. The hole was first drilled dry using a 12-in. (30-cm) outside diameter hollow-stem auger.

Once the augers encountered foundation rock beneath the dam, the hole was continued using coring methods with a double-tube wire-line core barrel using HQ-sized diamond core bits.

Upon completion of rock coring to total depth, each of the holes was reamed to about 5.5 in. (14 cm) in diameter to accommodate 2-in. (5-cm) well casing. Piezometers were constructed using Schedule 80 polyvinyl chloride and were screened at similar elevations, but in the different materials encountered: fractured metagranitic rock, felsic dike, and mafic dike.

The felsic dike was difficult to identify in the rock core because it tended to be intensely fractured and weathered, which led to very poor core recovery (Figure 7). The felsic dike also contained thin (typically less than 1-ft/0.3-m-wide) cross-cutting mafic dikes.

Attempts were made to use acoustic televiewer methods to confirm rock types in the holes through the dam prior to reaming the holes, but the boreholes were too unstable to provide useful images. The unstable holes also proved to be difficult to water pressure test.

In the four borings through the dam, evidence of the grout curtain was observed in one boring, CHE20, which had a roughly 0.25-in. (0.64-cm) seam of hard cement grout (Figure 7). As is evident in the core photo, the meta-granitic rock is intensely weathered and broken into many pieces (Rock Quality Designation of 0) with increasing distance from the grout seam.

Ultimately, only one piezometer out of the total 14 was installed with its screen exclusively in the felsic connector channel dike (CHE-23). The other piezometers intended to be screened in the felsic dike were either installed mostly in the mafic dike (CHE-30A) or in some combination of the two (CHE-30, CHE-21A, and CHE-25). The difficulty in finding the felsic dike where it was anticipated to be encountered suggests that its geometry was likely oversimplified. The other piezometers hit their targets of meta-granitic rock easily.

Figure 7. Rock core from CHE-20, drilled 10 ft (3 m) downstream from the grout curtain. A seam of cement grout from the grout curtain was recovered in this core run of predominantly meta-granitic rock, which is evidence that some grout went into the formation.

However, the rock on either side is very weak and intensely fractured, suggesting there may be more open fractures than filled fractures.

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Hydraulic Conductivity Testing (Packer Testing) As is traditional of many pre-grouting investigations, hydraulic conductivity testing (packer testing) was performed as the holes were being drilled. The intent was to determine the hydraulic conductivity of the different rock types under various pressures by performing a step test. One step test was performed per hole using a single packer. Pressure steps were 20 psi, 40 psi, 60 psi, 40 psi, and 20 psi (138 kPa, 276 kPa, and 414 kPa) at 5-minute durations for each step. The average Lugeon value (Lu) was calculated for each pressure step. Each Lugeon value was calculated as shown below: Q × P × 182.28 × 10, Lu = L where Q = flow rate during test (gallons per minute); L = length of hole being tested (ft.); P = test pressure (psi); 182.28 = conversion factor to convert feet to meters, gallons to liters, and psi to bar; and 10 = correction for the standard test pressure of 10 bar.

Lugeon values in the meta-granitic rock to the left (north) of the dike (CHE-29 and CHE-32) ranged from 3.2 to 5.9 Lu, with an average of 4.6 Lu. Lugeon values in the meta-granitic rock to the south of the dike (CHE-22 and CHE-31) ranged from 1.1 to 13.7 Lu, with an average of 4.7 Lu. Lugeon values in the dike (CHE-21A, CHE-30, CHE-30A, and CHE-23) ranged from 0 to 2.3 Lu, with an average of 0.6 Lu. Essentially, the packer testing indicated that the hydraulic conductivity of the fractured meta-granitic rock is about the same on either side of the dike and has low to moderate permeability, while the hydraulic conductivity of the dike materials has very low to low permeability (Quinones-Rozo, 2010).

Step tests can also provide some information about rock behavior, as described in Houlsby (1976), where the pressure steps are compared to the Lu value calculated for each pressure step. The following describes the five Lu patterns:

• Laminar—All Lu values are about equal regardless of the water pressure.

• Turbulent—Lu values decrease as the water pressures increase. The minimum Lu value is observed at the state with the maximum water pressure.

• Dilation—Lu values vary proportionally to the water pressures. The maximum Lu value is observed at the stage with the maximum water pressure.

• Wash-out—Lu values increase as the test proceeds.

Discontinuities’ infillings are progressively washed out by the water.

• Void filling—Lu values decrease as the test proceeds. Either non-persistent discontinuities are progressively being filled or swelling is taking place.

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Examination of the Lu patterns derived from the current investigation revealed that the dike materials exhibit wash-out, dilation, and laminar flows; the metagranitic rock on the left (north) side of the dike exhibits wash-out and turbulent flows; and the meta-granitic rock on the right (south) side of the dike exhibits laminar and wash-out flows. While all three of the rock types had at least one hydraulic conductivity test exhibiting wash-out, they also exhibited other behavior, such as laminar and turbulent flows, suggesting the rock masses do not behave consistently according to any one Lu pattern.

Falling Head Testing During the curtain grouting program for the construction of the enlargement dam, field geologists observed that some grout holes with low hydraulic conductivity values (as determined by packer testing) had very high grout takes, while other grout holes with higher hydraulic conductivity values had very low grout takes. Some grout holes through dikes had no grout takes under pressure, yet had high grout takes to “top off” the hole. This evidence suggested that at least some of the rock materials behaved differently under different pressure conditions that could not be adequately characterized by packer testing. Therefore, falling head tests were completed as part of the subject geologic investigation to try and quantify the hydraulic conductivity of these materials under low pressures.

The falling head tests consisted of rapidly introducing up to about 15 gallons of water into the well casings; the actual volume depended on how much room there was in the well casing. Four piezometers were selected for falling head tests: CHE-23 (screened in the felsic dike); CHE-30A (screened mostly in the mafic dike, but also somewhat in the felsic dike); CHE-29 (screened in the meta-granitic rock to the north of the dike); and CHE-31 (screened in the meta-granitic rock to the south of the dike). Pressure transducers in these piezometers were set to measure water levels every 500 milliseconds. The resultant recovery curve was imported into AQTESOLV and analyzed for hydraulic conductivity using the Kansas Geological Survey Model, which is useful for determining hydraulic conductivity of unconfined aquifers (Hyder et al., 1994).

Hydraulic conductivity values calculated from the falling head tests yielded the following results:

• CHE-23 (felsic dike) = 5.3 ft/d (0.00187 cm/s) • CHE-30A (mostly mafic dike) = 2.7 ft/d (0.000953 cm/s) • CHE-29 (meta-granitic rock north of dike) = 0.41 ft/d (0.000145 cm/s)

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Figure 8. Detailed groundwater monitoring of reservoir pool elevations (right axis) and groundwater elevations in piezometers installed below the dam crest (left axis). Peaks denoted with downward-pointing arrow; troughs denoted with upward-pointing arrow. Circled numbers indicate the order in which the groundwater levels hit their maximum or minimum levels. Dashed vertical line is a projection of the maximum and minimum reservoir pool elevations.

• CHE-31 (meta-granitic rock south of dike) = 0.64 ft/d (0.000226 cm/s) The results of the falling head tests show that the dike materials are more permeable by an order of magnitude than is the fractured meta-granitic host rock. This result is contrary to the packer test results, which indicated that the meta-granitic rock was more permeable than the dike materials.

Groundwater Monitoring Pressure transducers were set in all the piezometers installed during this geologic investigation and were programmed to collect depth to water readings once per hour. Groundwater levels measured in the piezometers were plotted with the reservoir elevation data, which were also measured in 1-hour increments.

In order to assess the rock’s response to fluctuations in the reservoir elevation, a prominent peak and trough in reservoir elevation were chosen to compare with the same peaks and troughs in the piezometer groundwater readings.

The piezometers completed through the dam, as shown in Figure 8 and Table 2, all showed strong peaks

and troughs. Piezometer CHE-20 screened in metagranitic rock to the left (north) of the dike had the fastest response times to reservoir highs and lows; the piezometer CHE-21A screened in the dike material (in this case both mafic and felsic dike) the second fastest, and the piezometer CHE-22 screened in meta-granitic rock to the right (south) of the dike was slowest. All the piezometers showed a faster response to lowering of the reservoir elevation than to raising the reservoir elevation. The amplitude, or difference, between the maximum and minimum reservoir elevations was 8.35 ft (2.55 m). The amplitude of CHE-22 was about 8.3 ft (2.5 m), while CHE-21A was 11.2 ft (3.4 m) and CHE-20 was 2.2 ft (0.67 m). These results suggest that although the meta-granitic rock to the left (north) of the dike has a lower amplitude change in piezometric head, it is more transmissive (more connected and open) than the dike materials or the meta-granitic rock to the right (south) of the dike based on its response time to changes in the reservoir elevation.

For the piezometers along the dam access road, as shown in Figure 9 and Table 2, all showed strong peaks and troughs. Piezometer CHE-29 screened in meta-granitic rock to the left (north) of the dike had the fastest response time to reservoir highs and lows.

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Nichols Table 2. Summary of groundwater monitoring comparisons of reservoir peaks and troughs, as shown in Figures 8, 9, and 10.

Dam crest Access road Downstream

Feature

Peak (date, time)

Elevation (ft)

Time Difference (days)

Trough (date, time)

Elevation (ft)

Time Difference (days)

Amplitude of Change (feet)

Reservoir CHE-20 CHE-21A CHE-22 CHE-29 CHE-30 CHE-31 CHE-23 CHE-24 CHE-26 CHE-27

5/18/16 6:00 5/19/16 8:17 5/20/16 6:06 5/20/16 12:57 5/19/16 10:00 5/22/16 18:00 5/20/16 16:15 5/23/16 18:00 5/23/16 17:59 5/27/16 19:10 5/23/16 18:35

2,925.6 2,889.8 2,907.9 2,899.0 2,910.5 2,903.8 2,903.8 2,860.9 2,852.5 2,833.8 2,838.5

0.00 1.10 2.00 2.29 1.17 4.50 2.43 5.50 5.50 9.55 5.52

5/9/16 13:00 5/9/16 14:17 5/10/16 23:06 5/10/16 2:57 5/9/16 15:43 5/10/16 0:38 5/11/16 3:49 5/11/16 3:51 5/11/16 14:59 5/12/16 13:02 5/11/16 11:35

2,917.2 2,887.7 2,896.6 2,890.7 2,903.3 2,896.2 2,895.4 2,860.1 2,849.6 2,832.1 2,837.0

0.00 0.05 1.42 0.58 0.11 0.48 1.62 1.62 2.08 3.00 1.94

8.35 2.15 11.23 8.27 7.12 7.63 8.38 0.74 2.84 1.76 1.55

The piezometer CHE-30/A screened in the dike material (in this case both mafic and felsic dikes) and the piezometer CHE-31 screened in meta-granitic rock to the right (south) of the dike exhibit similar response times to both reservoir high and low. All the piezometers showed a faster response to lowering of the reservoir elevation than to raising the reservoir elevation.

The amplitude, or difference, between the maximum

and minimum reservoir elevations was 8.35 ft (2.55 m).

The amplitude of CHE-29 was about 7.1 ft (2.2 m), while CHE-30/A was 7.6 ft (2.3 m) and CH-31 was 8.4 ft (2.6 m). Similar to the piezometers under the dam crest, these results suggest that the meta-granitic rock to the left (north) of the dike is more transmissive than the dike or the meta-granitic rock to the right (south) of the dike.

Figure 9. Detailed groundwater monitoring of reservoir pool elevations (right axis) and groundwater elevations in piezometers installed along the dam axis road (left axis). Note that the flat line shown for CHE-30 is for a time period during which the water levels dropped below the transducer. In addition, around May 19, 2016, the transducer in CHE-30 was removed and placed in CHE-30A, which is about 5 ft (1.5 m) away, and screened in a similar material at a similar elevation.

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Figure 10. Detailed groundwater monitoring of reservoir pool elevations (right axis) and groundwater elevations in piezometers installed downstream from the original dam (left axis). Note: there was no discernable trend in groundwater levels at CHE-25 for this time period; the water level in CHE-28 dropped below the transducer.

The piezometers downstream of the dam do not correlate to changes in the reservoir pool elevation as well as the piezometers located under or upstream from the dam (Figure 10 and Table 2). Most piezometers do not exhibit a strong peak and trough; rather, they reflect a gentle rise and fall of groundwater levels. Although only CHE-24 exhibits a significant peak, an attempt was made to determine differences in the gentle rise and fall of the other downstream piezometers. Of the more discernable peaks and troughs, piezometer CHE24 screened in meta-granitic rock right (south) of the dike has the fastest peak, and CHE-23 screened in the felsic dike has the fastest trough. CHE-26, screened in meta-granitic rock to the right (south) of the dikes, has the slowest response times. The amplitude of CHE-23 was about 0.7 ft (0.2 m), CHE-24 was 2.8 ft (0.85 m), CHE-26 was 1.8 ft (0.55 m), and CHE-27 was 1.6 ft (0.49). These results are presented but are not considered to be conclusive.

To summarize, the results of the detailed groundwater monitoring show that the piezometers under the dam crest and along the access road show strong peaks in groundwater levels that can be correlated with the rise and fall of the reservoir level. These changes suggest that the meta-granitic rock left (north) of the dike responds the fastest to changes in the reservoir elevation.

The dikes and meta-granitic rock to the right (south) of the dikes have similar response times to fluctuations in reservoir elevation. However, the changes in groundwater levels downstream from the dam were inconclusive.

Overall, what originally seemed like a straightforward method of analyzing water flow through fractured rock turned out to be a bit more complicated.

Dye Tracer Test Because the repair alternatives involved pressure grouting near the existing dam, there were concerns that the high-mobility grout, with an initial set time of about 6 hours, might travel all the way to and contaminate the drainage blanket of the original dam. Packer testing suggested that this travel distance was extremely unlikely, but in order to allay concerns over this possibility, a dye tracer test was conducted in the access road near the right abutment of the original dam. The intent was to inject red fluorescent dye into the formation and monitor its movement through the foundation rock by manually hand-bailing and screening the water sample for red color or fluorescence.

Red fluorescent dye was injected into an uncased hole (CHE-32) using a single packer set at 40 ft (12 m) deep at 40 psi (276 kPa) for 2 hours—essentially a

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Figure 11. A 55-gallon (208-L) drum of the red fluorescent dye used in the tracer test. A single packer was set at 40 ft deep with an applied pressure of 40 psi (276 kPa) for 2 hours.

2-hour-long packer test. Over the course of 2 hours, 53.7 gallons (203 L) of dyed water was injected into the hole (Figure 11). The nearest piezometers were monitored every 15 minutes for the 2 hours during injection and the first 2 hours post-injection, then every hour for another 4 hours, then again at 18 hours, and then after 24 hours. The weir at the base of the dam was sampled every half hour, until dye was observed beneath the dam (in CHE-20). Then the weir was sampled every 15 minutes until 8 hours post-injection.

The dye first appeared in CHE-20 after 2 hours. It was observed in the well for approximately 2.75 hours.

This was the only location where the red dye was visible in the groundwater sample. CHE-20 is located approximately 165 ft (50.3 m) southeast from CHE-32, indicating a nominal flow velocity of 82.5 ft (25.1 m) per hour, or 1,980 ft (604 m) per day at a pressure of 40 psi (276 kPa). This flow velocity converts to a Lu value (under ideal conditions) of roughly 500,000 Lu, whereas the packer test result for this hole at 40 psi (276 kPa) was 1.7 Lu—six orders of magnitude difference.

While this site is certainly not under “ideal conditions,” the comparison still provides some interesting insight into the magnitude of difference using the different methods.

Examining the progression of the red fluorescent dye, as depicted in Figure 12, 30 minutes after being detected in CHE-20, fluorescence was detected in CHE-21 and CHE-22, which are within the dike and on the right (south) side of the dike, respectively. The next positive detection was not until 3.75 hours after initial injection, with additional detections in CHE-25, CHE-30A, and CHE-31, which are downstream from the dam, within the dike complex, and to the right (south) of the dike, respectively. These piezometers were being sampled and 34

examined for red dye or fluorescence every 15 minutes; they continued to test positive for fluorescence until 4.5 hours after initial injection.

At 4.5 hours after initial injection, fluorescence was only detected in CHE-20 and CHE-25, which are screened within the meta-granitic rock and the dike complex, respectively. After 6 hours, fluorescence was detected only in CHE-25. Red dye or fluorescence was never detected in the weir, which collects and channels the underseepage being collected in the drainage blanket beneath the dam. Fluorescence or red dye was not observed in CHE-21A, CHE-23, CHE24, CHE-26, CHE-27, CHE-28, CHE-29, or CHE-30 (Table 1).

The apparent movement of the dyed water demonstrates that water injected on the left (north) side of the dike travels quickly through the meta-granitic rock to the left (north) of the dike, but it does not travel far to the right (south) of the dike, and in fact it appears to mostly stay contained to the north. The exception is the brief detection of fluorescence in CHE-22, CHE-30A, and CHE-31. Flow beyond this point was either too dilute to detect or there is reduced permeability rock prohibiting further flow. Because the dye was not observed in the seepage collection weir, there was much less concern of potentially contaminating the drainage blanket with grout.

Summary of Findings The geologic investigation included several methods for determining the preferential flow of groundwater through fractured meta-granitic rock and cross-cutting dikes: detailed groundwater monitoring, packer testing, falling head tests, and a dye tracer test. Table 3 compares the results of the falling head test, packer tests, and the dye tracer test. Packer testing suggested the meta-granitic rock on either side of the dike is the most permeable under induced pressure and has about the same average Lu value. The falling head tests suggest that the felsic dike is most permeable under low pressures.

Detailed groundwater monitoring along the dam crest and access road suggests that the meta-granitic rock to the left (north) of the dike responds faster to changes in reservoir levels, meaning it is more transmissive under normal reservoir operating conditions than either the dike or the meta-granitic rock to the right (south) of the dike.

The results of the dye tracer test showed that water in the meta-granitic rock to the left (north) of the dike travels much faster than predicted by either the packer test or the falling head test. Red fluorescent dye was never detected in the weir, suggesting that although the dyed water travelled very fast, it did not enter the

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Figure 12. Progression of red fluorescent dye, as observed by hand-bailing water samples (modified from Dean et al., 2016).

drainage blanket. Based on the areal occurrences of red fluorescent dye, there appears to be a semi-impervious boundary condition along the right (south) side of the dike.

Based on the findings of the geologic investigation, it appeared that possibly two of the four potential seepage pathways are likely contributors to the new seepage areas: (1) through or along the dike complex

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Nichols Table 3. Comparison of falling head tests to packer testing. All falling head tests resulted in higher hydraulic conductivity values than did the packer testing.a Falling Head Test (X) Well ID CHE-23 CHE-30A CHE-30 CHE-32 CHE-29 CHE-20 CHE-31 CHE-22

Packer Test (Y)

Percent Change

Screened Interval

(ft/d)

(cm/s)

(Lub )

(ft/d)

(cm/sc )

(X − Y)/Y × 100

Felsic dike Mafic dike Dike (both) North gr/dike North gr North gr South gr South gr

5.3 2.7 — 1980d 0.41 — 0.64 —

0.00187 0.000953 — 0.698500 0.000145 — 0.000226 —

0.04 0.3 0.4 1.8 5.2 4.0 8.1 1.4

0.0015 0.011 0.015 0.07 0.19 0.15 0.30 0.05

0.00000052 0.0000039 0.0000052 0.00002 0.000068 0.000052 0.00011 0.00002

359,462 24,323 — 2,984,945 114 — 114% —

North gr = meta-granitic rock left (north) of the dikes; South gr = meta-granitic rock right (south) of the dikes. a 1 ft/d = 0.000352778 cm/s. 1 Lu = 0.000013 cm/s. b Average Lugeon (Lu) value selected for that hole. c Under ideal conditions (i.e., homogeneous and isotropic) 1 Lu is equivalent to 1.3 × 10−5 cm/sec (Fell et al., 2005). d Flow velocity calculated based on first appearance of dye, not a falling head test.

from the connector channel and (2) through fractures in the meta-granitic rock in the connector ridge. No evidence was found to support the idea that additional or new seepage was travelling through ungrouted fractures in the foundation rock of the original dam. The fourth potential seepage pathway—along potentially dried/cracked materials along the outlet pipeline—is not likely to be a contributor. The piezometer installed in the pipe bedding to assess this potential has always been dry, suggesting that any dried/cracked lake sediments did not create a conduit to allow water to seep into the pipeline backfill.

REMEDIAL GROUTING PLAN As stated at the beginning of this article, the purpose for completing this geologic investigation was to use the results to develop an effective remedial grouting plan to reduce or stop the flow of water coming through new seepage areas. The key components of a successful grouting plan are the locations of the grout holes, the grout mix, and the applied pressures that will inject the grout into the formation. The geologic investigation provided a good indication of where the grout holes should be drilled and the pressures that should be applied to achieve maximum results.

As summarized previously, most of the results of the geologic investigation pointed to the meta-granitic rock as being the most permeable material under normal reservoir operation conditions. However, it became clear that the dike complex also plays a role in the transmission of seepage water, as evidenced by the dye tracer test.

This information was used to select the location for the grout holes, which was along the dam ac36

cess road at the mapped trace of the dike complex.

Two rows of grout holes were placed to cover the extent of the dike, about 60 ft (18 m), with additional grout holes extending to the left (north) side of the dike.

The results of packer testing and falling head tests suggested that to achieve maximum effect in the dike materials, low to no applied pressure should be used to inject the grout. Pressures up to about 40 psi (276 kPa) could be applied to the meta-granitic rock to achieve maximum grout takes; however, in practice, it can be very difficult to discern which type of material is in the actual drilled grout hole using typical drilling methods for this type of work (air hammer or similar tools), because the drill cuttings can look similar. For ease of implementation, the remedial grouting plan specified a maximum pressure of 10 psi (69 kPa).

CONCLUSIONS A multi-faceted geologic investigation was conducted to determine the pathway(s) for new seepage downstream from the original Crafton Hills Dam and to determine the characteristics of the foundation materials with respect to hydraulic conductivity. The investigation included installation of piezometers in the three main rock types (meta-granitic rock, mafic dike, and felsic dike), detailed groundwater monitoring, packer testing, falling head tests, and a dye tracer test.

The detailed groundwater monitoring suggested that the meta-granitic rock to the left (north) of the dike complex is more transmissive under normal reservoir operations than the dikes or in the meta-granitic rock

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to the right (south) of the dikes. Packer testing suggested the meta-granitic rock on either side of the dike complex has higher permeability than does the dike complex itself. However, the falling head tests indicate that the felsic dike has the highest permeability under lower pressures. The dye tracer test, essentially a 2-hour-long packer test, resulted in a calculated hydraulic conductivity of 1.8 Lu. However, based on the observation of red dye at CHE-20 after 2 hours, the observed hydraulic conductivity was calculated to be roughly 500,000 Lu. The results of the geologic investigation suggested that the seepage pathway was a combination of the new conditions created by the construction of the enlargement dam: through fractures in the meta-granitic rock in the connector ridge and through or along the through-going dike complex exposed in the connector channel. Although evidence of the grout curtain was observed in the rock core boring for CHE-20, no direct evidence was found that supports new or additional seepage through ungrouted fractures in the foundation of the original dam. The piezometer installed in the pipe backfill has always been dry, indicating that water is not seeping into the outlet pipeline through dried or cracked lake sediments or rock.

With these results, it became clear that any remedial grouting should target the through-going dike complex and the meta-granitic rock to the north. The hydraulic conductivity testing suggested implementing low pressures for the dike complex and higher pressures for the meta-granitic rock. However, the dye tracer test indicated that subsurface flow velocities could be quite high, so the grouting plan used low pressures.

The results of this geologic investigation underscored the importance of using more than one method to determine hydraulic properties of foundation materials; however, using more than one investigation method can also give contradictory results. But even contradictory results are useful to get a better idea of the range of foundation material properties that may be encountered. Just like cone penetration testing paired with geotechnical drilling and sampling provides a more complete picture of subsurface soils, geologists should consider adding falling head tests, tracer tests, or other types of flow tests to their standard pre-grouting in-

vestigation toolbox to get a more complete picture of how water flows through fractured rock under different pressure conditions.

ACKNOWLEDGMENTS This project would not have been possible without the assistance of several excellent engineering geologists and engineers at the Department of Water Resources, Division of Engineering, Project Geology and Dams and Canals Sections and the Division of Integrated Regional Water Management- Glendale.

REFERENCES BRUCE, T. AND GLICK, F., 2003, Crafton Hills Dam, Additional Results of Geologic/Seepage Investigation During Initial Reservoir Filling: Department of Water Resources, Project Geology Report Number 58-71-15, 24 p.

DEAN, J.; NICHOLS, H.; BUTLER, T.; DUSSELL, B.; HIGHTOWER, N.; MLINAREVIC, A.; PERRY, D.; ZIMMERMAN, M.; AND CURRY, R., 2016, Crafton Hills Dam, Results of 2014 to 2016 Seepage Investigation: Department of Water Resources, Project Geology Report Number 58-71-45, 274 p.

FELL, R.; MACGREGOR, P.; STAPLEDON, D.; AND BELL, G., 2005, Geotechnical Engineering of Dams: Taylor & Francis Group plc, London, U.K. 905 p.

HOULSBY, A. C., 1976, Routine interpretation of the Lugeon watertest: Geological Society London, Vol. 9, No. 4, pp. 303–313.

HYDER, Z.; BUTLER, J. J., Jr.; MCELWEE, C. D.; AND LIU, W., 1994, Slug tests in partially penetrating wells: Water Resources Research, Vol. 30, No. 11, pp. 2945–2957.

MATTI, J.; MORTON, D.; COX, B.; CARSON, S.; AND YETTER, T., 2003, Geologic Map and Digital Database of the Yucaipa 7.5’ Quadrangle, San Bernardino and Riverside Counties, California: U.S. Geological Survey Open-File Report 03-301, http://pubs.usgs.gov/of/2003/0301/ MORTON, D. AND MILLER, F., 2006, Geologic Map of the San Bernardino and Santa Ana 30’ × 60’ Quadrangles, California: U.S. Geological Survey Open-File Report 2006-1217, https://pubs.usgs.gov/of/2006/1217/ NICHOLS, H. J.; MLINAREVIC, A. N.; AND BARRY, G. R., 2014, Traditional cement grout versus stable, high-mobility grout— A case study at the Crafton Hills Dams. In Proceedings of the 34th Annual USSD Conference, Dams and Extreme Events— Reducing Risk of Aging Infrastructure under Extreme Loading Conditions, U.S. Society of Dams, Denver, CO. pp. 1505–1524.

QUINONES-ROZO, C., 2010, Lugeon test interpretation, revisited. In Proceedings of the 30th Annual USSD Conference, Collaborative Management of Integrated Watersheds, U.S. Society of Dams, Denver, CO. pp. 405–414.

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Pedro Miguel Fault Investigations: Borinquen Dam 1E Construction and the Panama Canal Expansion DAVID L. SCHUG∗ AECOM, 401 West A Street, Suite 1200, San Diego, CA 92101

PAUL SALTER CHRISTOPHER GOETZ AECOM, 999 Town and Country Road, Orange, CA 92868

DEREK IRVING Canal de Panama, Corozal Oeste, Edificio 732, Balboa. Panama, Rep de Panama

Key Terms: Dams, Construction, Faulting, Panama Canal ABSTRACT Borinquen Dam 1E is part of the new Pacific Access Channel (PAC) of the Panama Canal Expansion. The 2.3-km-long zoned rockfill dam forms the navigational channel providing navigation access from the Gaillard Cut to the new Post-Panamax Pacific Locks. A key geologic objective during construction was to confirm locations and activity of faults mapped at the dam during design, namely the Pedro Miguel Fault (PMF) and its suspected newly mapped “main trace.” The design allowed for core and filter widening at the anticipated location of the PMF at the south abutment and at a west branch of the PMF (believed to be the main active trace of the fault) mapped along the dam axis about one-third of the way north from the south abutment. As-built geologic mapping revealed complex faulting associated with the PMF crossing the southeast half of the foundation, the PAC, and the nearby Dam 1W foundation along a north-south trend. Trenching and age dating of alluvium overlying the faults crossing the Dam 1E foundation and overlying the PMF at Dam 1W indicated the unfaulted alluvium was latest Pleistocene to early Holocene age.

At Dam 1E, the core and filters were widened to accommodate potential fault rupture on the PMF and a previously unrecognized fault revealed across the width of the dam foundation. The west branch of the PMF (trenched and mapped during design investigations) was determined to not exist at Dam 1E based on mapping the dam foundation and other extensive excavations created for the PAC.

∗ Corresponding

author email: david.schug@aecom.com.

INTRODUCTION Borinquen Dam 1E is part of the new Pacific Access Channel (PAC), a key element in the Panama Canal Expansion project, which also includes new Post-Panamax Pacific and Atlantic Locks and various other canal upgrades (Figure 1). Borinquen Dam 1E, along with several smaller dams (collectively known as the “Borinquen Dams”), retain the water in the canal that provides navigation access from the Gaillard Cut to the new Pacific Locks (Figure 2). The dams retain Gatun Lake, the main waterway of the Panama Canal.

Dam 1E extends approximately 2.3 km from near the northern gates of the 100-year-old Pedro Miguel Locks to the north side of Fabiana Hill (Figure 3). Excavations commenced February 2011 and included geologic mapping of the dam foundation and adjoining areas of the PAC. A cofferdam was installed along the southwest shoreline of Miraflores Lake to facilitate the construction of Dam 1E (Figure 3). Final embankment construction was completed June 2015, and the PAC opened to navigation June 2016.

The purpose of this article is to describe fault investigations carried out during construction of Dam 1E. These investigations were performed to confirm the location and activity of major faults, namely the Pedro Miguel Fault (PMF) and its suspected newly mapped “main trace,” which were reported during design studies to have produced repeated Holocene displacements at the Borinquen Dams. This article summarizes key geologic findings that demonstrate the absence of significant Holocene activity on the PMF at the Borinquen Dams.

Dam 1E retains the PAC water level in Gatun Lake at about elevation 27.1 m, or 10.7 m above the minimum operating elevation of Miraflores Lake. The constructed width and height of Dam 1E vary, based on the depth of excavation required to achieve foundation

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Figure 1. Project vicinity map showing the location of the Pacific Access Channel.

strength and weathering objectives; however, the dam is generally 100–150 m wide at foundation grade and 25–30 m high, with upstream and downstream slopes inclined at 3:1 (H:V). Figure 4 shows the zoned con40

struction of Dam 1E, with rockfill embankments, chimney filters, and a central clay core. Preconstruction groundwater levels at Dam 1E, as high as 18 m above the dam design foundation

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Pedro Miguel Fault, Panama Canal

Figure 2. View looking northwest along the new Third Set of Locks and the Pacific Access Channel. Photo taken January 2016.

level (URS, 2009), were lowered by construction dewatering.

Design of Borinquen Dam 1E took place between 2007 and 2010, as described in Mejia et al. (2011). In the same time frame, paleoseismic fault studies were performed by Earth Consultants International (ECI)

at the Borinquen Dams in support of design, as described in Rockwell et al. (2010). Numerous seismic design challenges existed, including the potential for strike-slip fault rupture up to 3 m across the dam foundation (Mejia et al., 2011). Based on fault trenching near Dam 1W, offsets of paleochannel margins were reported to range from 2.0 to 2.5 m for the most recent event (AD 1621) on the PMF (ECI, 2007a, 2007b).

Based on the paleoseismic data, a displacement of 3 m was calculated to be the 80th-percentile estimate of the potential horizontal offset during surface rupture of the PMF at the PAC (URS, 2008a). The seismic design criteria required that the thickness of the core, chimney filter, and drains within 50 m of the PMF be widened to at least 1.5 times the design horizontal fault displacement of 3 m to provide an adequate margin against full offset of those critical zones and to ensure their continuity to accommodate the estimated fault displacement. The design allowed for widening at the anticipated location of the PMF at the south abutment and at a west branch of the PMF (believed at the time to be the main active trace of the fault) mapped as crossing the axis of Dam

Figure 3. Aerial photo showing the footprint locations of the Borinquen Dams prior to construction. Fabiana Hill forms the south abutment of Dam 1E. Construction of the dam required a cofferdam along the shoreline of Miraflores Lake.

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Figure 4. Cross section of Dam 1E, showing dam foundation zones.

1E about one-third of the way north from the south abutment.

Foundation excavations commenced in February 2011 through initial excavations of the foundation area of the inboard shell (adjacent to the PAC channel), then of the dam core trench, and lastly of the outboard shell (adjacent to Miraflores Lake). Foundation excavations in each of these areas required the removal of overburden fills, alluvium, and weathered materials over broad areas. The core trench was typically excavated at least several meters deeper than the adjacent inboard and outboard shells. Final embankment construction was completed in June 2015 (Figure 5).

Dam foundation geologic mapping was undertaken jointly by Autoridad del Canal Panama (ACP) and URS (now AECOM) geologists. Detailed investigations of faults near Borinquen Dam 1E, within and outside the PAC area, were also conducted to corroborate locations of faults previously mapped and inferred during design.

TECTONIC SETTING Panama is located between the Cocos, Nazca, Caribbean, and South American tectonic plates. Con42

sidering global positioning system (GPS) data, Kellogg and Vega (1995) suggested the existence of a rigid Panama–Costa Rica microplate that is moving northward relative to the stable Caribbean Plate. Seismicity studies by Adamek et al. (1988) and other GPS studies

Figure 5. View northwest along Dam 1E looking toward the Gaillard Cut. The cut slope bordering the PAC is in Miocene basalt and the Pedro Miguel Formation. Miraflores Lake is visible in the lower right side of the photo, and Gatun Lake is visible in the upper right side of the photo. Photo taken March 2016, just prior to flooding the PAC.

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Pedro Miguel Fault, Panama Canal

Figure 6. Geologic map of the canal zone (from Stewart et al., 1980). The Pedro Miguel Fault as mapped is locally known as the “Stewart Trace.” North of the canal, the fault forms the contact between the Miocene Las Cascadas Formation (Tlc) and the Oligocene Panama Formation (Tp). To the South of the canal, the fault was mapped for only a short distance, forming the contact between Miocene basalt (Tb) and Miocene La Boca Formation (Tl).

by Trenkamp et al. (2002) suggest that the Panama– Costa Rica microplate is a low-seismicity province that is presently undergoing minimal internal tectonic deformation. Conversely, Rockwell et al. (2010) suggested that the tectonic deformation that began in the Miocene is ongoing and is distributed among faults throughout Panama and western Colombia, resulting in internal deformation of the isthmus by folding and faulting. Rockwell et al. (2010) suggested that the PMF, ´ Fault, and related faults compose an acthe Limon tive zone of faulting that extends from north-central Panama southward for at least 40 km and likely extends offshore into the Gulf of Panama. The PMF, which crosses the Panama Canal between the Miraflores and Pedro Miguel Locks (Stewart et al., 1980), is the primary tectonic structure that has been the fo-

cus of a seismic hazard characterization for the canal expansion project.

Figure 6 shows the location of the PMF as mapped

by Stewart et al. (1980). The PMF was initially mapped in the Canal Zone as a bedrock fault that formed a structural boundary between several geologic units (Woodring, 1957; Stewart et al., 1980). North of the canal, the fault was mapped as forming the contact between the Miocene Las Cascadas Formation and the Oligocene Panama Formation. To the south of the canal, the fault was mapped for only a short distance, forming the contact between Miocene basalt and the Miocene La Boca Formation.

As part of a seismic hazard characterization for the canal expansion project, the ACP retained ECI to conduct investigations to assess the locations and activity

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Figure 7. Pre-construction generalized bedrock geologic map of Dam 1E. During design, the Pedro Miguel Fault was reported to have east and west branches. The east branch represented the Stewart Trace of the Pedro Miguel Fault; the west branch was considered the Main Trace of the Pedro Miguel Fault.

of the PMF in the area of the proposed Borinquen Dams (ECI, 2007a, 2007b, 2010). Based on their investigations, ECI reported that the fault has a geomorphic signature indicative of a recurrent Holocene active, right-lateral strike slip fault that includes linear valleys, ridges, and escarpments, with several offset and deflected channels.

ECI mapped two strands of the PMF in the area of the proposed dams (Figure 7). The more easterly strand, which coincides with the location mapped by Stewart et al. (1980), is mapped along the northwest side of Fabiana Hill (the south abutment of Borinquen Dam 1E). This strand of the fault is informally referred to herein as the “Stewart Trace.” The more westerly fault strand of the PMF is shown as diverging to the northwest from the Stewart Trace. ECI reported that this trace of the fault has produced 100 ± 20 m of offset of two parallel, east-flowing Late Quaternary tributary drainages of the Coccoli River, just south of Dam 1W. ECI considered this previously unmapped strand of the fault to be the through-going “main strand” of the PMF in the vicinity of the canal. The fault, shown as PMF Main Trace on Figure 7, was inferred to intersect Dam 1E about 600 m northwest of the dam’s southern abutment.

In 2007, ECI performed fault trenching in the Rio Coccoli area, about a kilometer to the southwest of the proposed PAC, to characterize the activity of the PMF and to develop fault slip parameters for design of the proposed Borinquen Dams. The trenching was done in the area of the two parallel offset drainages noted above. Based on the trenching investigations, ECI concluded that the zone of Quaternary deformation is lo44

cally more than 100 m wide, but the majority of the slip was accommodated by a “western strand” of the fault that they named the “Main Trace.” ECI also concluded that the PMF experienced at least two, and possibly up to three, late Holocene earthquakes in the past approximate 1,500 years, and they surmised that the most recent event was the AD 1621 earthquake that devastated the old capital of Panama (Panama Viejo). ECI (2007b) recommended that right-lateral strike slip fault displacement of at least 3 m should be considered for design of the proposed dams. The 3-m displacement was related to the PMF Main Trace. The more easterly trace of the PMF (the Stewart Trace) was judged as active but as only having the potential to produce about one-third of the slip as on the Main Trace (about 1 m).

DAM SITE GEOLOGY Pre-Construction (Design) Geologic Investigations Pre-construction geological investigations carried out at Dam 1E by ACP, URS, and others included reviewing historic data and aerial photographs, geologic mapping, test pits, and borehole investigations (Pinilla et al., 2011); geophysics (Kaufmann et al., 2008); and paleoseismic fault trenching (ECI, 2007a, 2007b, 2010).

The geologic assessment completed for the design of Dam 1E (URS, 2008b, 2009) identified the dam as mostly underlain by the Miocene La Boca Formation (Woodring and Thompson, 1949; Woodring, 1957; and Stewart et al., 1980), consisting of stratified conglomerate, sandstone, siltstone, and clay shale. Sub-aerial volcaniclastic rock agglomerate of the Miocene age

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Pedro Miguel Fault, Panama Canal

Figure 8. View of the excavation at Dam 1E looking east toward Miraflores Lake. The excavation exposed regular dipping beds of La Boca Formation (shown in photo) over broad areas of the foundation.

Pedro Miguel Formation underlies the south end of the dam. Basalt within the dam footprint was mapped as small intrusive dikes and sills in the north and central portions of the dam and as larger intrusive bodies and/or flows in the south abutment area (Fabiana Hill).

The PMF Stewart Trace and Main Trace were mapped across the dam (ECI, 2007a, 2007b). Figure 7 shows the generalized bedrock geology and major faults anticipated during design.

Dam Foundation Geologic Mapping Field methods used to map the foundation geology varied, depending on the dam foundation zone. The contractor excavated defined parcels to an appropriate foundation objective grade per the specifications and undertook final cleaning with a smooth-edged excavator bucket. This resulted in broad exposures of the foundation bedrock geology. ACP survey crews marked a 5 × 5-m survey grid with station and offset numbering on the cleaned foundation. ACP and URS geologists collaborated on identification of key geologic R GPS units to features and then used Trimble Yuma geologically map the cleaned inboard and outboard shell foundations. The core trench was excavated into slightly weathered rock and was mapped by hand using a Brunton compass. A detailed geologic map of the dam foundation was compiled at a scale of 1:300 (URS, 2015). This level of geologic detail allowed documentation of displacement along faults mapped across the dam.

Geologic Units Interbedded La Boca Formation conglomerate, sandstone, siltstone, clay shale, ash flow, and lignite beds were mapped over the northern three-quarters of the dam foundation. La Boca Formation bedding in the dam foundation has an east-west–trending

strike and dips to the south at angles around 15–20◦ (Figure 8).

Thinly to thickly bedded PMF consisting of stratified volcaniclastic sandstone and minor siltstone units were mapped in the dam foundation in the south portion of the dam. The volcaniclastic sandstone units typically include fine to coarse sand-sized tuff and basalt fragments in a dark fine-grained volcaniclastic (tuffaceous) matrix.

Broad exposures of alluvium overlying the bedrock were mapped along the outboard margins of the Dam 1E excavation. The alluvium typically consisted of clayey gravel channel deposits overlain by and interfingered with stratified fine-grained sandy silt units. Alluvium was removed during foundation excavation and cleaning.

Samples of the alluvium were collected for age dating to help constrain fault activity, described below. Laboratory age-dating analyses included optically stimulated luminescence (OSL) dating performed by the Luminescence Dating Laboratory at the University of Cincinnati. Radiocarbon (C-14) accelerator mass spectrometer (AMS) dating was performed by the University of Georgia Center for Applied Isotope Studies.

FAULT INVESTIGATIONS Investigations of fault activity during construction were done to test the design assumption that the PMF Stewart Trace and PMF Main Trace are Holocene active faults that cross the dam. As the foundation excavation at Dam 1E progressed, ongoing geologic mapping and field observations indicated that faults anticipated for dam widening were overlain by potentially datable alluvium. The excavations resulted in extensive cuts exposing alluvium overlying bedrock faults. Observations and mapping of these cuts indicated the alluvium had not been displaced by faulting. The alluvium

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Figure 9. Fault map of Dam 1E showing generalized foundation geology, as mapped during construction. Faults were named for their intersection with the dam centerline station. The dam was widened at the 1 + 860 Fault (approximate mid-area of dam) and the Pedro Miguel Stewart Trace (south abutment). Widened zones are shown as hachure pattern along the dam centerline.

overlying faults was further investigated by trenching and sampling for age dating. Investigations of excavations at Dam 1W were done to supplement the finding of the investigations at Dam 1E. The following summarizes the key findings of these investigations.

Fault Investigations at Dam 1E Numerous faults were mapped in the Dam 1E foundation. These ranged from discontinuous shears with only a few centimeters of displacement to faults that cross the entire width of the dam foundation, with thick (several meters–wide) shear and gouge zones, and displacements of tens of meters, in the Miocene bedrock.

No displacements were observed in the overlying alluvium. Dominant fault orientations regionally trend north to northwest, with steep dips. The through-going faults mapped across the foundation have an apparent right-lateral sense of displacement, with a few exceptions where apparent left-lateral displacements were mapped. Figure 9 shows the principal mapped faults that are continuous over the foundation of Dam 1E. The principal faults observed in the Dam 1E foundation, in terms of continuity, thickness of gouge, and offset, included the PMF Stewart Trace (at the south abutment); the 1 + 860 Fault; 1 + 240 Fault; and the 2 + 380 Fault (named for their intersections at the dam centerline station).

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The PMF Stewart Trace crosses Dam 1E at the south abutment (Figure 9). The fault was mapped at about the same location as mapped by Stewart et al. (1980) at Fabiana Hill. Two strands of the fault were mapped and designated the Stewart Trace “East Branch” and the Stewart Trace “West Branch” (URS, 2014a). The total displacement in early Miocene rocks across the faults is unconstrained at the dam. The dam foundation geology indicated the juxtaposition of dissimilar bedrock types and changes in the orientation of bedding across the faults. Exposures of alluvium were sparse overlying the PMF Stewart Trace at Dam 1E; however, foundation cuts at Dam 1W (along the continuation of the fault to the southwest across the PAC) produced extensive exposures of alluvium overlying the fault.

The 1 + 860 Fault was mapped continuously across the foundation of Dam 1E (Figure 10), with varying thicknesses of sheared rock and gouge. The fault produces apparent right-lateral displacement of approximately 160 m in distinct La Boca Formation beds. At the outboard toe of the dam, the 1 + 860 Fault is overlain by several sequences of unbroken alluvium. Multiple trenches were excavated into the alluvium overlying the 1 + 860 Fault (Figure 11). Alluvial samples from these trenches, which were dated by AMS and OSL, place the age of the unbroken alluvium as latest Pleistocene to early Holocene (URS, 2014b).

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Pedro Miguel Fault, Panama Canal

Figure 10. View south along the strike of 1 + 860 Fault in the Dam 1E core trench excavation. At this location, the fault displaces carbonaceous shale beds (left photo center) within the La Boca Formation.

The 1 + 240 Fault was mapped continuously across the foundation (Figure 9). The strike is generally northsouth, with moderate to steep dips to both the east and west. Approximately 300 m of right-lateral apparent displacement occurs in La Boca Formation beds across the 1 + 240 Fault in the PAC. At the outboard toe of the foundation excavation, strands of the 1 + 240 Fault are overlain by unbroken alluvium. Samples of the alluvium yielded latest Pleistocene dates.

The 2 + 380 Fault was mapped over a length of several hundred meters in the core trench; the fault was mapped as terminating in the outboard shell. Near Fabiana Hill, the dip of the 2 + 380 Fault appeared to flatten, and the fault merged with a bedding plane shear. Relatively pervasive bedding plane shears were mapped in the La Boca Formation at Dam 1E; several faults were observed to be offset by flat-lying bedding plane shears. Faults mapped at the south portion of the dam appear to branch or splay from more continuous faults mapped in the PAC (described below), although

Figure 11. Examples of trenches excavated in alluvium overlying the 1 + 860 Fault at Dam 1E, also showing locations of OSL-dated alluvium within the trenches. Multiple trenches were excavated in alluvium over the width of the foundation. Trench locations are shown on Figure 9.

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Figure 12. View looking west of the initial Dam 1E foundation excavation and the cofferdam along Miraflores Lake. The inboard shell had been excavated to foundation grade, allowing initial placement of rock fill. The foundation excavation exposed laterally continuous beds of La Boca Formation across the width of the foundation. Distinctive layers of red clay shale (shown with a dashed yellow line) and other distinct beds within a thick sequence of the La Boca Formation (Figure 8) were not displaced across the same area where the “Main Trace of the Pedro Miguel Fault” had been mapped at the dam prior to construction. The east-facing cut slope bordering the PAC (mid-portion of photo) was more than 700 m long and up to 70 m deep at the mapped location of the Main Trace of the PMF (shown by the red dashed line). There were no faults in the Miocene columnar basalt. Photo taken December 2011.

none were through-going across the foundation at Dam 1E. The PMF Main Trace mapped by ECI (2007a, 2007b) was not exposed in the foundation at Dam 1E (Figure 12). The anticipated location of the fault was based on trenches excavated at the former ground surface within the proposed footprint of Dam 1E (ECI, 2010). The actual foundation excavation was between 13 and 18 m below the same area where paleoseismic trenches were excavated during design studies. During construction, the dam foundation, inclusive of the previously trenched area, was deepened to remediate adverse bedding in the La Boca Formation, including bedding plane shears and ancient landslide deposits.

These ancient landslides became reactivated during excavation of the PAC; slumping had destabilized a portion of the inboard shell of Dam 1E (Figure 13).

Subsequently, a several hundred meters–long portion of the PAC cut slope, including the Dam 1E foundation, was redesigned and deepened to remove the ancient landslide materials and other weak materials encountered in the cuts. The section redesign increased dam stability by mitigating against potential failure surfaces passing under the shell directly into the PAC channel.

At the elevation of the redesigned deepened founda48

tion, continuous beds of La Boca Formation with distinct marker beds were mapped with a uniform, lowangle dip across the entire width of the dam (Figure 12).

This well-bedded stratigraphy in the Miocene sedimentary rocks precluded the suspected PMF Main Trace as previously mapped. The exposed geologic conditions in the Dam 1E foundation indicated that a fault that could correlate with the Main Trace of the PMF does not exist at Dam 1E. An explanation for this apparent discrepancy is that ancient landslide features may have been incorrectly interpreted as nearly flat fault planes in the relatively shallow paleoseismic fault trenches that were excavated during the design investigations.

The PAC West Bank cut slope produced extensive exposures of Miocene (K-Ar dated at ± 18 mya) columnar basalt across an approximate 700-m-long, up to 70-m-deep excavation at the mapped location of the Main Trace of the PMF (Figure 12). The well-formed uniform cooling joint pattern in the basalt rock mass would be capable of revealing even a minor fault. The absence of any fault in the columnar basalt further confirmed the Main Trace of the PMF does not exist in the Dam 1E foundation (URS, 2012a).

Based on these investigations, faults mapped across the Dam 1E foundation, including the 1 + 860, 1 + 240,

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Pedro Miguel Fault, Panama Canal

Figure 13. View looking east of slope failures along the inboard edge of Dam 1E and the PAC. The extent of instability was over the cut slope forming the inboard edge of the dam foundation. The slope was redesigned and lowered, starting at the approximate location shown by the arrow, continuing northerly over an approximate 500-m-long portion of the foundation along the PAC. Fabiana Hill is at the right photo center. Photo taken November 2011.

and 2 + 380 Faults, were determined to be overlain by, and did not displace, Holocene-dated alluvium. The PMF Stewart Trace was similarly shown to not displace Holocene-dated alluvium at Dam 1W, as discussed below.

Fault Investigations at Dam 1W At Dam 1W, the Pedro Miguel Stewart Trace Fault was well exposed in dental excavations of the foundation at the location mapped by Stewart et al. (1980).

The fault zone observed in the core trench juxtaposes basalt and PMF bedrock. The fault zone could be readily traced from the core trench to the adjacent backcut excavation slope (Figure 14). The 1:1 (H:V) cut slope provided a several meter–thick exposure of alluvium overlying bedrock across more than 300 m of the back cut. The alluvium consisted of a basal gravel overlain by reddish-brown fine sand, silt, and clay. The alluvium/bedrock contact in the back cut at Dam 1W demonstrated that the alluvial deposits were not displaced by the Stewart Trace (Figure 15).

When the outer shell rockfill had been placed to within a few meters below the alluvial contact eleva-

tion, an excavator was used to help clear the slope and better expose the basal alluvial gravel contact. A series of progressively deepened notches and benched trenches were excavated to map the alluvial contact in detail. Samples of the alluvial gravel matrix and overlying fine-grained alluvium were collected for laboratory age dating, including OSL and radiocarbon AMS. This detailed trenching and age dating performed at Dam 1W demonstrated that the Stewart Trace does not displace basal alluvial gravels of early Holocene age (URS, 2012b).

A splay of the PMF was also exposed in the Dam 1W foundation, about 200 m south of the Stewart Trace.

This fault had been recognized in the PAC where it branches southeasterly from the Stewart Trace, continuing into and across the south end of Dam 1W. This fault, named the “Southeast Strand Fault,” was mapped into the dam back cut, where it is overlain by the same alluvial gravel sequence overlying the Stewart Trace Fault. By benching and deepening the exposure with an excavator, it became apparent that the Southeast Strand Fault had displaced (about 0.3-m vertical separation) the lower portion of the basal alluvium (Figure 16).

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Figure 14. View west along the Pedro Miguel “Stewart Trace Fault” (red dashed line) in the core trench and foundation excavations at Dam 1W. The fault displaces tuff and agglomerate of the Pedro Miguel Formation and basalt in the cut and core trench of Dam 1W. Laterally continuous basal alluvial gravels were exposed for over 300 m in the dam back cut (shown by yellow dashed lines). The upper part of the slope is in fill. The top of the slightly weathered columnar basalt is in the lower right photo corner. Columnar basalt extends continuously from the north abutment of Dam 1W over the PAC west bank cut, shown in Figure 12.

The construction excavation at Dam 1W produced a broad exposure of alluvium overlying well-defined faults in the foundation bedrock. This exposure demonstrated that the PMF Stewart Trace was overlain by, and did not displace, basal alluvial gravels of early Holocene age and that the Southeast Strand Fault had produced minor (approximately 0.3-m vertical) displacement in the same dated alluvium.

CONCLUSIONS As-built geologic mapping of the Dam 1E foundation, as well as fault investigations performed outside the PAC boundaries (URS, 2012b, 2014a), revealed a pattern of faulting associated with the PMF crossing the southeast half of Borinquen Dam 1E, the 50

PAC, and Dam 1W along a north-south trend (Figure 17). The PMF forms a system of sub-parallel faults that appear to have distributed tectonic strain in the Miocene bedrock across a zone at least several hundred meters wide in the vicinity of Dams 1E and 1W. This faulting appears to have produced only minor strain in the Holocene and latest Pleistocene alluvium.

South of Dam 1E, the PMF narrows slightly, extends across the PAC, and crosses the footprint of Dam 1W as two strands referred to as the Stewart Trace Fault and the Southeast Strand Fault. Trenching performed at Dam 1W (URS, 2012b) demonstrated that the Stewart Trace Fault does not displace basal alluvial gravels of early Holocene age. Trenching at Dam 1W also demonstrated that the Southeast Strand Fault has

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Pedro Miguel Fault, Panama Canal

Figure 15. View west of the Dam 1W back-cut exposure of Holocene-dated basal alluvial gravels and fine-grained alluvium overlying the Stewart Trace of the Pedro Miguel Fault. When the top of the rockfill was near the level of the alluvial gravels, trenched notches were excavated into the cut slope at the Stewart Trace to confirm the absence of fault displacement in the alluvium. Locations of the fault trenches at Dam 1W are shown on Figure 17. Photo taken December 2013.

produced only minor displacement of the same basal alluvial gravels of early Holocene age.

Considering actual fault locations encountered, and fault displacements measured in the Miocene bedrock, the core and chimney filters and drains of Dam 1E were widened at the south abutment to accommodate the Stewart Trace Fault. The dam core was also widened at the intersection with the 1 + 860 Fault, which was mapped along about 1,200 m of the dam foundation.

Activity on 1 + 860 Fault was unknown at the time the core trench was excavated; therefore, the decision was made to widen the dam’s core at that time. Subsequent mapping efforts in the outboard shell of Dam 1E provided an opportunity to date unbroken alluvial deposits overlying the 1 + 860 Fault. As a result, it was determined that the 1 + 860 Fault was overlain by, and

does not displace, early Holocene–dated alluvium at Dam 1E (URS, 2014b).

The 1 + 240 Fault was mapped extending across the Dam 1E footprint and extending into the PAC. Similar to the 1 + 860 Fault in style, the 1 + 240 Fault produced right-lateral displacement in early Miocene La Boca beds. Trench logging and multiple age dates indicate the 1 + 240 Fault does not displace Holocene and latest Pleistocene alluvium. The core of Dam 1E was not widened at the intersection with the 1 + 240 Fault.

The key geologic investigation objectives during construction were to confirm location and activity of faults mapped at the dam during design, namely the PMF and its suspected newly mapped “main trace.” These investigations demonstrated the following:

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• There was no geologic evidence for multiple late Holocene earthquake events with meters of slip per event at the Borinquen Dams, as suggested by the design investigations.

• The PMF Main Trace does not exist at Dam 1E and the PAC.

Figure 16. View west of trenching the Dam 1W back-cut, where the cut was deepened at the Southeast Strand Fault. At this trench, the Southeast Strand Fault had produced about 0.3 m of vertical separation at the lower portion of the basal alluvium. Photo taken December 2013.

• The Pedro Miguel Stewart Trace Fault does not displace Holocene alluvial gravels at Dam 1W. • The Southeast Strand of the PMF produced minor (about 0.3 m) displacement of early Holocene alluvium at Dam 1W.

The PAC is a nearly 300 m wide and up to 70 m deep at the location of the PMF. This provided an extraordinary opportunity to observe and map the major strands of the PMF at levels substantially deeper than the original grade. During excavation of the PAC, locations of the fault were observed, measured, and surveyed (using GPS) in temporary construction cuts that are now removed, or covered by the Borinquen Dams.

Dam 1E was designed to withstand extreme events with high reliability, including the maximum fault displacement on the PMF. The seismic design criteria required that the thickness of the core, chimney filter, and drains at Dam 1E be widened within 50 m of Holocene active fault locations. It was anticipated during design that widening would be required at the PMF (Stewart Trace Fault) and the PMF Main Trace. Although no evidence for Holocene activity was found at Dam 1E, the dam was widened at the PMF Stewart Trace and the 1 + 860 Fault. Considering the absence of Holocene fault activity at Dam 1E, the decision to widen the dam at the Stewart Trace and the 1 + 860 Fault provides a

Figure 17. Fault map of the PAC between Dam 1W and Dam 1E. The Pedro Miguel Fault was mapped crossing the southeast half of Borinquen Dam 1E, the PAC, and Dam 1W along a north-south trend.

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Figure 18. View north toward the Pedro Miguel Locks (photo center) showing the completed PAC and Dam 1E.

robust mitigation against fault rupture. The PAC opened to navigation June 2016 (Figure 18).

ACKNOWLEDGMENTS ACP and AECOM (formerly URS) personnel and others who participated in geologic field investigations and engineering during construction of Dam 1E follow: ACP: Maximiliano DePuy, Pastora Franceschi, Jamie Arrocha, Roberto Miranda, Francisco Ponce, Romy Coronado, Roberto Campo, Mauricio Lacerda, Eladio Almengor, and E. E. Portugal; URS: Lelio Mejia, James Toose, Daniel Meier, Robert Urban, Derek Rector, Adam Avakian, Jason Moore, and Ja˜ son Castaneda.

Special acknowledgments to Jorge De la Guardia; Director, New Locks Project, and Jorge Fernandez: Project Manager, Dam 1E. The authors appreciate the opportunity to have worked together at Borinquen Dam 1E. The authors also appreciate review comments received on the draft of this article.

The text was reformatted per Journal style.

REFERENCES ADAMEK, S.; FROHLICH, C.; AND PENNINGTON, W., 1988, Seismicity of the Caribbean-Nazca boundary; Constraints on microplate tectonics of the Panama region: Journal Geophysical Research, Vol. 93, pp. 2053–2075.

EARTH CONSULTANTS INTERNATIONAL (ECI), 2007a, Paleoseismic Trenching of the Pedro Miguel Fault in the Coccoli Located Immediately Southwest of the Panama Canal: Report to the Autoridad del Canal Panama, February.

ECI, 2007b, Quantitative Characterization of the Pedro Miguel Fault, Determination of Recency of Activity on the Miraflores Fault, and Detailed Mapping of the Faults through the Proposed Borinquen Dam Location: Report to the Autoridad del Canal Panama, November.

ECI, 2010, Additional Trenching of the Pedro Miguel and Miraflores Faults in Coccoli, Immediately Southwest of the Panama Canal: Report to the Autoridad del Canal Panama, August.

KAUFMANN, R. D.; IRVING, D.; YUHR, L.; AND CASTO, D., 2008, A Geophysical Investigation for the Panama Canal Expansion: American Institute of Professional Geologists; Arvada, CO. KELLOGG, J. N. AND VEGA, V., 1995, Tectonic Development of Panama, Costa Rica, and the Columbian Andes: Constraints from Global Positioning System Geodetic Studies and Gravity: Geological Society of America, Special Paper 295.

MEJIA, L.; ROADIFER, J.; FORREST, M.; ABREGO, A.; AND DE PUY, M., 2011, Design of the dams of the Panama Canal Expansion: Proceedings of the United States Society on Dams 2011 Annual Meeting and Conference, San Diego, CA. PINILLA, R.; SAMPACO, K.; AND IRVING, D., 2011, The Panama Canal’s Third Set of Locks Project: Geologic Setting and Site Characterization: Geotechnical Frontiers 2011, pp. 2356–2365.

ROCKWELL, T.; GATH, E.; GONZALES, T.; MADDEN, C.; VERDUGO, D.; LIPPINCOTT, C.; DAWSON, T.; OWEN, L. A.; FUCHS, M.; CADENA, A.; WILLIAMS, P.; WELDON, E.; AND FRANCESCHI, P., 2010, Neotectonics and paleoseismology of the Limon and Pedro Miguel Faults in Panama: Earthquake hazard to the Panama Canal: Bulletin Seismological Society America, Vol. 100, No. 6, pp. 3097–3129.

STEWART, R. H.; STEWART, J. L.; AND WOODRING, W. P., 1980, Geologic Map of the Panama Canal and Vicinity. Republic of Panama: U.S. Geological Survey Miscellaneous Investigations Series Map I-232, scale 1:100,000, 1 sheet.

TRENKAMP, R.; KELLOGG, J. N.; FREYMUELLER, J. T.; AND MORA, H. P., 2002, Wide plate margin deformation, southern Central America and northwestern South America, CASA GPS observations: Journal South American Earth Sciences, Vol. 15, pp.

157–171.

URS, 2008a, Characterization of Fault Displacement Hazards, Design of the Borinquen Dams: Technical Memorandum prepared for Autoridad del Canal de Panana, February.

URS, 2008b, Geologic Assessment of Design of New Borinquen Dams: Technical Memorandum prepared for Autoridad del Canal de Panana, July.

URS, 2009, Task A.1.4, Foundation Materials—Dam 1E Geotechnical Interpretive Report (GIR): Report prepared for Autoridad del Canal de Panama, January.

URS, 2012a, Geologic Memorandum Fault Investigations, Construction of Borinquen Dam 1E: URS Task Order #21, URS Project No. 26818044.00021, April 2012 (Submittal No. C1106).

URS, 2012b, Geologic Assessment: Faults Exposed at Dams 1E, 1W and the PAC: Task Order #30 (Geologic Support to ACP), URS Project No. 26818044.00030, December 2012 (Submittal No. C1156).

URS, 2014a, Assessment of Recency of Displacement of Pedro Miguel Fault Borinquen Dam 1E: ACP Task Order #35, URS Project No. 26818044.00035, February 2014 (Submittal No. C1240).

URS, 2014b, Additional Fault Investigations, Including Station 2 + 310 to 2 + 715 and South Abutment, Borinquen Dam 1E, Expansion of the Panama Canal: ACP Task Order #41, URS Project No. 26818044.00041, November 2014 (Submittal No. C1272).

URS, 2015, Final Construction Report, Panama Canal Expansion Program, PAC-4 Borinquen Dam 1E: Rev. 1, May 2015 (Submittal No. C1271-003).

WOODRING, W. P., 1957, Geology and Paleontology of Canal Zone and Adjoining Parts of Panama: U.S. Geological Survey Professional Paper 306-A. WOODRING, W. P. AND THOMPSON, T. F., 1949, Tertiary formations of Panama Canal Zone and adjoining parts of Panama: Bulletin American Association Petroleum Geologists, Vol. 33, pp. 223– 247.

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Qualitative Rock Wedge Stability Evaluation Performed for Foundation of Green Peter Dam, Oregon ∗

TODD N. LOAR

U.S. Army Corps of Engineers Risk Management Center, 12596 W. Bayaud Avenue, Suite 400, Lakewood, CO 80228

Key Terms: Dams, Foundations, Engineering Geology, Geotechnical, Rock Mechanics ABSTRACT Green Peter Dam is a concrete gravity structure located in west-central Oregon on the Middle Santiam River within the Willamette Valley Basin. A risk assessment for the project identified continuous and adversely oriented low-angled shear zones underlying portions of the foundation that could potentially facilitate sliding instability of one or more monoliths during earthquake loading. Conceptually, a potential foundation rock wedge could be formed with a shear zone as its sliding surface and joints as the side planes. This wedge, which would otherwise be stable under static conditions, could feasibly be displaced and/or shifted during seismic ground shaking, resulting in significant structural damage and/or breach of the dam. A qualitative evaluation was performed to characterize the geomechanical conditions and geometry of movement (i.e., kinematics) of the dam-foundation system associated with rock wedges.

The study revealed that wedges could indeed be formed by adversely oriented and intersecting rock mass discontinuities. The qualitative evaluation concluded that the displacement geometry and geologic conditions in the foundation collectively suggest that the wedges would likely be stable under even large probabilistic seismic loading. While no concrete dams are known to have failed due to seismic loading, an increased knowledge of higher seismicity in the Pacific Northwest region warranted a careful evaluation to ensure that the risks of foundation rock wedge deformation are well characterized, and that our level of confidence in the available data is acceptable to better constrain the potential risk posed by this failure mode. This paper summarizes the background, findings, and results of the preliminary and qualitative dam-foundation system stability evaluation that was performed for Green Peter Dam.

∗ Corresponding

author email: todd.n.loar@usace.army.mil.

INTRODUCTION Green Peter Dam (GPD) is located on the Middle Santiam River, 4.7 mi (7.6 km) upstream from its confluence with the South Santiam River, a tributary of the Willamette River. The site is approximately 30 mi (48 km) southeast of Albany, in Linn County, OR (Figure 1).

GPD was authorized as a flood risk management project, but it also provides hydropower generation, water supply, and recreation benefits to the region. The concrete gravity dam was constructed between 1963 and 1967 with a maximum height of 380 ft (115.8 m), a crest length of 1,517 ft (462.4 m), 26 monoliths (note: monoliths 12 and 13 were combined when the foundation design was refined, so monolith 13 does not exist) from the right abutment (NW) to the left abutment (SE), and a 20-ft-wide crest. The spillway is controlled by two Tainter gates (45 ft (13.7 m) wide by 48 ft (14.6 m) tall) positioned between monoliths 19 and 21, and an 80 MW powerhouse constructed below monoliths 16–19. The dam is oriented N50W (310 degrees) across the valley, which is oriented N40E (220 degrees) at the dam location. Foundation treatment consisted of excavation to competent bedrock (i.e., removal of poor rock and/or adverse materials); concrete back-filled drifts to improve foundation stability; an upstream-angled grout curtain; and two rows of downstream-angled relief drains advanced from drainage galleries and tunnels in the abutments. Figure 2 presents the overall layout, configuration, and project features at GPD. Background In 2011, a Potential Failure Mode Analysis and Semi-Quantitative Risk Assessment were performed as part of the U.S. Army Corps of Engineers (USACE) Dam Safety Program. The findings from the initial risk screening-level analysis identified a number of potential failure modes at GPD that required more advanced engineering evaluation.

One of the risk-driving failure modes included monolith sliding instability due to deformation of the rock foundation during a large seismic event (approximately 0.5g). This instability mechanism is associated

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Figure 1. Site location map and image of Green Peter Dam.

with the presence of intersecting discontinuities consisting of:

• continuous upstream-downstream, sub-horizontal clayey shears; and • near-vertical bedrock joints, dikes, and faults.

Collectively, these discontinuities could intersect within the dam foundation to form a removable rock wedge positioned such that it could be displaced and/or deformed during a large seismic loading event, causing distress to the structure and subsequent uncontrolled reservoir release. A removable rock wedge configura-

Figure 2. Project layout and features.

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Figure 3. Conceptual diagram of foundation rock wedge. Note: Diagram shows an arch dam, but the wedge concept is similar for a concrete gravity dam (from Londe, 1973; modified by Scott, 1999).

tion requires a combination of intersecting discontinuities that includes a base plane (sliding), side planes, and a release plane positioned geometrically in the foundation so they form a removable wedge (i.e., key block) that reaches the surface downstream of the toe of the dam. Figure 3 presents a conceptual graphical diagram of a rock wedge positioned in a dam foundation (Londe, 1973; Scott and Von Thun, 1993; and Scott, 1999).

GPD was constructed during a period of extensive USACE dam building in the Pacific Northwest with active district, division, and headquarters involvement and support. Subsequently, the design team consisted of highly experienced engineers and geologists who recognized the potential for sliding instability, differential settlement, and erosion issues represented by the subhorizontal clayey shears. Based on their geotechnical assessments, considerable measures were implemented to characterize and treat the foundation during construction. The foundation investigation and treatment consisted of the following:

• detailed geologic mapping and documentation of the foundation excavation, exploration tunnels,

•

• • •

adits/drifts, diversion tunnel, drainage galleries, drill holes, and large-diameter calyx holes; substantial over-excavation of portions of the foundation where shears zones were present near the foundation surface (shallower than 20 ft below rock surface); construction of an additional row of foundation drains; consolidation grouting; and construction of “reinforcing” concrete back-filled tunnels (drifts) where the shear zones were deeper than about 20 ft below the rock surface (Nesbitt and Corns, 1967). See Figure 2 for relative locations and configurations of concrete back-filled drifts.

At the time of dam construction, physical displacement of three-dimensional (3-D) foundation rock wedges was not considered as a failure mechanism. Today, we recognize that failure of rock wedges within the foundation can be a viable failure mechanism for concrete dams (examples: Malpasset Dam, 1959, France [GPD was being built when the failure of this dam was still being analyzed by P. Londe]; St. Francis Dam, 1928, CA; Austin/Bayless Dam, 1911, PA; and Camara Dam, 2004, Brazil).

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Figure 4. Typical 2-D free-body-diagram and foundation failure models (USACE, 1994, 1995).

These notable failures were primarily attributed to elevated pore pressures or uplift pressures acting on the foundation rock wedge; lack of an under-drain system; shallow foundation embedment; unrecognized pre-existing landslide features; erosion of weak materials; and/or changes made to the dam configuration without re-analysis. However, state-of-the-practice in dam foundation engineering now considers high ground accelerations from large earthquake events potentially capable of triggering rock wedge displacement and foundation failure.

Typical Methodology for Gravity Dam Stability Analysis Structural stability analysis for a concrete gravity dam typically assumes the sliding direction is oriented directly downstream (i.e., perpendicular to the dam axis), involving a single monolith without significant 3-D contributions from adjacent monoliths or from the foundation. Preliminary analysis usually consists of a two-dimensional (2-D) limit equilibrium model or a 2-D numerical model using finite element modeling or distinct-element modeling. The failure surface is often input as a single, linear base plane idealized along the concrete-rock contact or through the foundation rock mass (occasionally with a passive wedge at the toe) without influence of the 3-D effects from the entire dam-foundation system (USACE, 1995, 1994; 58

Nicholson, 1983). Figure 4 presents a typical free body diagram and annotated configuration of the different foundation sliding surfaces used in traditional gravity dam foundation stability analyses.

In reality, the resistance of monolith and foundation wedge sliding is strongly influenced by the spatial shear strength, loading, and the geometric configuration of the dam-foundation system involved in deformation, including the following parameters: • shear strength parameters: ◦ conditions between adjacent monolith contraction joints, ◦ rock wedge discontinuity surface shear strengths, ◦ large-amplitude base plane roughness, and ◦ spatial shear strength variability over sliding surface(s); and • kinematic parameters (i.e., geometry of sliding): ◦ loading vectors (reservoir, dam, seismic, uplift pressures) applied to the system, ◦ rock wedge configuration and removability ori entation with respect to the loading vectors and required monolith sliding direction (i.e., typi cally directly downstream), ◦ spatial relationship between the monolith(s) footprint and rock wedge area, ◦ foundation and downstream top of rock topo graphic surface,

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◦ large-scale base plane asymmetric topography, and ◦ lateral confinement of the dam-foundation system.

These factors can combine to induce significant rotation, torsion, dilation, and interlocking between the concrete monolith(s) and surrounding rock mass and within the foundation rock wedge and/or rock mass.

The 3-D aspects of the deformation geometry can mobilize additional shear strength that can be a substantial contributor to increasing sliding resistance and is not fully accounted for in the traditional 2-D structural analysis (Goodman and Bro, 2005; Galic et al., 2007, 2008).

Therefore, development of a conceptual, or qualitative, understanding and characterization of the threedimensionality of the entire dam-foundation failure system relative to the likely loading vectors and sliding direction is essential in evaluating stability of the dam. This characterization should be performed and documented prior to any stability analysis or before complex computer-aided design and drafting models are developed. The geologic evaluation will provide additional information for the risk analysis and/or engineering design team to reduce uncertainties and support conservative assumptions; inform the stability analysis model construction and input parameters; and aid interpretation of the analytical 2-D stability modeling results.

Removable Rock Wedge Criteria Since the 1959 Malpasset Dam disaster in France, the state-of-the-practice has evolved to recognize a number of physical conditions that must be satisfied for foundation rock wedge instability to be possible:

• Geologic discontinuities must exist in proximity to, and intersect each other to form wedges at the scale of the dam foundation.

• Discontinuities, or combinations of discontinuities, must be continuous/persistent enough at the scale of the dam foundation in the upstream-downstream direction (or perhaps in a closely spaced “stepped” or en-echelon configuration) to interact with one another to form a discrete removable block.

• The sliding surface(s) and sliding direction (i.e., line of intersection of the wedge between the base and side planes) of the wedge must reach the surface downstream or be close enough to the top of the rock near the toe of the dam to allow potential for failure.

• The shear strength of the discontinuity surface(s) forming the rock wedge and 3-D effects of the dam-

foundation system providing resistance to sliding must be overcome by a combination of the reservoir, dam, water pressure, and/or seismic loading forces.

When these conditions are met, geotechnical and geological engineering practitioners should consider that foundation rock wedge instability could be a feasible failure mode, and they must thoroughly characterize and document the 3-D aspects of the dam-foundation system. This is often a difficult task to accomplish because the supporting documentation and data may be lacking in sufficient detail, or the foundation rock of interest is inaccessible because it may be buried by the structure or reservoir. Additionally, it can be challenging and expensive to obtain post-construction geotechnical exploration data that fully capture the spatial variability and details of the rock foundation necessary to confidently perform the evaluation.

Scope The scope of work for this evaluation is limited to review of available design and construction documents consisting of: drawings, figures, maps, and data provided in the 1969 Foundation Report (USACE, 1969) and other documents (e.g., construction photos, previous inspection reports, and presentation materials).

No laboratory testing of the shear materials was available to provide a better understanding of the strength and consistency of the sub-horizontal features. No new field investigations or tests were carried out as part of this work.

Fortunately for this study, the foundation at GPD was thoroughly documented with extensive design investigations; construction photos; and detailed, highquality geologic and fracture mapping and geologic descriptions. These investigations focused not only on the rock units, contacts, intrusions, and faults, but also on the geometry, configuration, nature, surface condition, and spatial distribution of discontinuities in the foundation floors and cut-slope walls between monolith sections.

This study would have been considerably more difficult without the detailed foundation mapping performed during construction and preserved in the USACE archives, as post-construction exploration would not likely replace these essential documents. The quality of foundation documentation provided a high level of confidence in the characterization of the rock wedges.

ROCK WEDGE EVALUATION Regional Geology GPD is located within the western portion of the Cascade Geologic Province, the volcanic arc

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Figure 5. Regional physiographic and geologic map.

associated with the Cascadia Subduction Zone. The Cascade Range is subdivided into two parts, the Western Cascades (where GPD is located), which is the older inactive volcanic arc; and the High Cascades, which is the modern active volcanic arc (Sherrod and Smith, 2000).

The Western Cascade Geologic Province is composed of middle Tertiary pyroclastics, lava flows, and igneous intrusions. These rocks were subsequently intruded by dikes, sills, and small stocks of basic to intermediate rock (basalt and andesite). The region is cut by numerous faults and by sets of high-angle stress joints. Many of the older rocks exhibit alteration due to circulation of hydrothermal fluids and corrosive solutions and gases rising along the faults, which have caused intense alteration to clay and zeolite minerals and softening of some rock units. Figure 5 shows the physiographic setting and regional geology at GPD. The foundation geology at GPD consists of multiple lobes and layers of inter-bedded sequences consisting of basaltic to andesitic lava flows, breccia, and lapilli tuff cross-cut by basaltic dikes, sills, and intrusions.

Figure 6 shows a schematic of a typical section of the generalized stratigraphy and annotated photo of an exposure at the site.

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The volcanic flow sequence generally has a low-angle dip of 2 to 12 degrees to the southeast (into the left abutment and slightly upstream), but specific basalt flow or tuff surfaces may exhibit an undulating, largeamplitude (∼5–10 ft) contact over relatively short distances (∼25–40 ft) due to the nature of the irregular lava flow depositional system, consisting of lava flow pressure ridges, lobes, and the hummocky topography of underlying and/or adjacent flow surfaces.

Regional warping, uplift, tectonic stresses, and rebound in addition to stress relief associated with erosion and development of the valley morphology have resulted in deformation and inter-layer slip accommodated preferentially along the weaker altered tuff layers, resulting in the sub-horizontal shear seams that are present throughout the dam foundation.

Site Geotechnical Conditions The dam is oriented N50W (310 degrees) across a relatively narrow portion of the Middle Santiam River where the stream channel is oriented S40W (220 degrees). The river has incised into a layered series of approximately 15 distinctive lava flows and five interbeds of pyroclastic units that are cross-cut by three

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Figure 6. Photo of schematic anatomy of lava flow sequence and excavation in left abutment at GPD.

intrusive dikes, referenced as Albert, Big Ben, and Homer, oriented approximately north-south; and three relatively small faults referenced as Lilly, Audry, and Irene that trend parallel to the dam axis (across the

valley) and that are limited to monoliths 22–26. These layered flow sequences were exposed and mapped in the foundation excavation and are presented on Figure 7.

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Figure 7. Foundation geologic map showing the primary project components and location of geologic cross section A-A (Figure 8).

The basalt layers exhibit a moderate- to welldeveloped cooling joint pattern composed of intersecting sub-vertical fractures that have been previously healed and/or cemented but subsequently re-opened due to regional deformation stresses. This cooling joint system appears to be discontinuous through some of the over-riding breccia and tuff units. The configuration of the multiple cooling joints results in rhombusshaped vertical columns in the basalt and ultimately a very rough, jagged, and irregular surface on vertical exposures across long distances.

The sub-horizontal shear features identified at GPD are strongly associated with preferential deformation parallel to, and along, the weaker lapilli tuff bedding layers with a local orientation of N35E, dipping 5–12 degrees to the SE (dip/dip direction of 5◦ –12◦ /125◦ ) into the left abutment and slightly upstream. These shear features have been interpreted as being laterally continuous upstream and downstream of the dam and across the site. The shears are described as brown to yellowish-brown, medium- to high-plasticity clay (CL/CH) seams, 0.1 to 0.5 ft thick, with noted slickensides, and a platy fabric exhibiting parallel and anastomosing planes of weakness. The material contains angular gravel- and sand-size rock fragments embedded in the clayey matrix.

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Nine sub-horizontal shears or shear zones are mapped across the GPD foundation, referenced as follows: Steen, Kay, Big Kid, Francis, Lena, Kay (present in both the left and right abutments), Hazel, Nesbitt, and Martha, from right to left abutments. The Big Kid shear may not be directly related to the lapilli tuff, but rather a cross-cutting shear between the Lena and Kay shears potentially related to mechanical deformation, and therefore it has a steeper dip angle of approximately >15 degrees to the southeast than the other sub-horizontal shears. Figure 8 presents a cross section cut from the right to left abutment 50 ft downstream of the dam centerline showing these geologic features and foundation configuration (viewed looking upstream).

Conceptualized rock wedges are shown on the section to help visualize their relative positioning relative the foundation topography and dam geometry.

Geologic cross sections cut from upstream to downstream at monoliths 11 and 23 have been annotated to show the subsurface conditions in areas where the Big Kid shear and Kay shear remain in place and underlie portions of the dam foundation at depths of approximately 35 ft. These areas, shown on Figure 9, are where potential wedge instability is considered to be the most likely based on depth of the shear and high dam loadings at monolith 10–11 and 22–23.

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Figure 8. Geologic cross section A-A from right abutment to left abutment, 50 ft downstream of dam center line (approximate elevations are given where each shear intersects the foundation surface), showing conceptualized rock wedges under monoliths 5–6, 10–11, 22–23, and 25–26.

As part of the foundation treatment, 15 tunnels were excavated and subsequently back-filled with concrete to remove portions of the shear zone layers under GPD where they were deeper than about 20 ft and not economical to excavate. Drifts 1–10 are in the right abutment and were excavated along the Big Kid shear parallel to the dam axis under monoliths 10 and 11. In the left abutment, Drifts H-1, K-1, and K-2 were excavated into the Francis/Lena shear and Kay shear under monolith 22, and Drifts M-1 and M-2 were excavated into the Martha shear under monoliths 25 and 26 in an orientation perpendicular to the dam axis. In addition, the diversion tunnel was also back-filled with a concrete plug.

The purpose of the concrete back-filled drifts was to act as additional rock reinforcement in areas where the shear zones were deeper than was economical to remove by over-excavation (deeper than about 20 ft below the designed foundation grade). By removing portions of the shears along the tunnels and replacing the material with concrete, the design stability analysis showed that 2-D sliding of monoliths along those surfaces had an acceptable factor of safety due to the depth of the sliding plane and increased strength of the concrete (Nesbitt and Corns, 1967). The locations and configuration of the concrete back-filled drifts are shown in Figures 2, 7, and 8.

Top of Rock Surface The downstream top of rock topographic surface is an important aspect for evaluating the potential for a foundation rock wedge to intersect with the downstream ground surface and subsequently undergo displacement. Development of a top of rock contour map is often one of the first steps performed in identifying the most critical locations for wedge sliding to occur.

The downstream shape of the top of rock surface, elevation, slope aspect, and side-valley re-entrant swales or ridges control the downstream spatial extent, and “day-lighting” projection (sliding and side plane surfaces), and therefore removability of a rock wedge. The wedge intersection with the top or rock surface helps in assessment of the likelihood of displacement as well as the wedge volume.

Figure 10 presents an annotated plan-view map from the 1969 Foundation Report (USACE, 1969) showing the top of rock contours (in dashed gray) estimated from the borings, road cuts, and outcrops. Black dashed lines identify zones R1, R2, and L1, and a red dashed line identifying zone L2 delineates the downstream areas where foundation rock wedges must reach the surface for each of the monoliths underlain by shears (i.e., 5–6, 10–11, 22–24, and 26). Arrows delineate either topographic ridges where the top of rock surface

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Figure 9. Geologic cross section from upstream to downstream across monoliths 11 and 23, showing the foundation conditions in areas identified as the most susceptible to rock wedge failure. (Note: Green numbers and dashed contacts relate to interpreted rock-quality zones and are not pertinent to this evaluation.)

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Figure 10. Top of rock contour map in the areas downstream of potential wedges. Arrows indicate a ridge (yellow) or swale (blue) topography.

protrudes inward into the valley (yellow arrows), or reentrants (swales) where the top of rock projects away from the valley (blue arrows).

Downstream of monoliths 5–6 and 10–11, on the right side of the valley, the top of rock surface exhibits the morphology of a buried rock ridge that is oriented at azimuth 160 degrees to the south-southeast and approximately 30 degrees to the dam axis, protruding toward the river and creating a top of rock shape that closes toward the valley immediately downstream of monoliths 5–6 and 10–11; these are delineated as Areas R1 and R2 in Figure 10, respectively. Downstream of monolith 26, on the left side of the valley, the top of rock similarly exhibits the morphology of a ridge.

This configuration immediately downstream of monoliths 5–6, 10–11, and 26 tends to be the more favorable configuration because there is no loss of rock mass (in fact, there is an increase in rock mass), and the geometry forces the outcrop pattern of the southeastdipping base plane shears further downstream in the area where foundation wedges would have to move. Additionally, a failed rock wedge would have to rupture upward through a large passive wedge and more intact rock to reach the surface in the immediate vicinity of the toe of monoliths 5–6, 10–11, or 26. This geometry tends to increase stability by creating larger foundation

wedges that must propagate downstream for the base slide plane to reach the surface (daylight). These zones are delineated by black dashed circles labeled as Areas R1, R2, and L1 in Figure 10.

The top of rock surface immediately downstream of the left abutment below monoliths 22–24 is oriented away from the valley orientation, suggesting a buried drainage re-entrant or swale. The top of rock contours are oriented almost 40 degrees from the dam axis into the hillside (south-southeast). This top of rock surface configuration tends to be the least favorable, because there is a reduction of rock mass in the areas immediately downstream from monoliths 22–24 where a foundation wedge base plane could reach the surface closer to the dam toe, creating the potential for a smallervolume foundation wedge. This zone is delineated by a red dashed circle labelled as Area L2 in Figure 10.

Discontinuity Evaluation During the initial 1964 field investigation, several large-diameter calyx core holes were drilled, logged, and photographed down-hole to document the 360 degree in-situ conditions of the primary discontinuities that compose the rock mass in the GPD foundation.

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Figure 11. Geologic foundation mapping highlighting primary discontinuities (monolith wall sections are spread out flat in plan view and labeled with yellow identification boxes).

Images from the calyx holes show that subhorizontal shears are continuous around the perimeter of the boring, but they have significant variability in thickness and exhibit undulations over short distances. This suggests that a likely failure path through the shear at the scale of the foundation would cross through differing conditions and rock fragments. The contact between basalt and breccia is similarly continuous across the boring, has variable aperture, and undulates over short distances. The typical sub-vertical and intersecting jointing pattern in the basalt unit appears to be irregular and stepped with many joint terminations and a high degree of jaggedness over relatively short distances, providing additional evidence that a side plane composed of sub-vertical joints would be a very rough surface.

A detailed discontinuity analysis was performed using the 1969 foundation geologic map. This evaluation involved measuring and compiling a database that consisted of approximately 465 data points describing the following attributes: joint orientation; horizontal joint trace length across the foundation surface; vertical joint trace length mapped on monolith walls 12 and 21; and approximate spacing of sets. Other observational and spatial information was recorded from the 1969 Foundation Report and construction photos (USACE, 1969). Based on this discontinuity analysis, 66

it is recognized that there are lava flow contact surfaces, sub-horizontal shear zones, dikes, faults, and three distinct primary rock joint sets (i.e., sets J1, J2, and J3) that cross-cut the foundation.

Figure 11 presents the compiled foundation mapping (with vertical monolith wall sections flattened and annotated with yellow labels), highlighting the three dominant joint orientations, mapped shear zones (with approximate spot elevations), cross-cutting dikes, concrete back-filled drifts, and construction photos approximating the conceptual rock wedges described in the following sections. It is noted that some of the detail/resolution in Figure 11 is lost in this significantly reduced full-size plot, but the figure is included to reflect the level of detail available to evaluate a complex foundation.

Figure 12 presents statistical results of the horizontal and vertical joint trace length analysis as well as the stereographic projection and tabulated discontinuity data for all the data.

The following information describing the joint sets was derived based on the foundation discontinuity evaluation:

• Joint set J1 (orange color on Figures 11 and 12) is a minor joint set oriented east-northeast, and it strikes obliquely across the dam foundation, has a low

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Figure 12. Histogram of trace length measures for side plane joints across the entire dam foundation, delineated by horizontal length and vertical length. Generalized stereonet and tabulated summary of joint data are also shown. (Data were further sorted based on foundation monoliths being analyzed, but this is not presented herein.)

frequency of occurrence, and has a shorter continuity and relatively short trace length, with an average range of 10–30 ft, and wider average spacing of 15– 20 ft compared to J2 and J3. The J1 set is oriented approximately 56 degrees from the dam axis and 34 degrees from the assumed loading and sliding direction of the monolith (i.e., perpendicular to the dam axis).

• Joint set J2 (green color on Figures 11 and 12) is oriented northwest and strikes sub-parallel to the dam axis. In general, joint set J2 has a high frequency of occurrence across the foundation (and in the vertical monolith walls) and is the most prominent joint orientation in the dam foundation. These joints have the highest degree of continuity (i.e., the longest trace lengths and could be followed up the block walls in some locations), with an average length of approx-

imately 30–50 ft, a maximum trace length range of 125–>200 ft, and a typical spacing of about 5–10 ft.

These J2 joints tend to be sub-parallel to the local faults and seem to have more lateral and vertical continuity than the J1 and J3 joints.

• Joint set J3 (purple color on Figures 11 and 12) is oriented northeast and strikes obliquely across the dam foundation. This set was found to have a high frequency of occurrence across the GPD foundation and a moderate continuity, or medium trace length, with an average range of about 20–55 ft over the entire foundation; however, in the area of monoliths 10 and 11, the average J3 trace length is about 55 ft.

This set has a fairly high continuity (not as high as J2) of 15–45 ft, with a spacing of about 5–10 ft. The J3 set is oriented approximately 70 degrees from the dam axis and 20 degrees from the assumed loading and

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Figure 13. 3-D physical paper model of foundation mapping and conceptual foundation rock wedge. (Does not show the downstream top of rock topography.)

sliding direction of the monolith (i.e., perpendicular to the dam axis).

A 3-D physical paper reconstruction model of the foundation mapping of the left and right abutments was used to visualize and communicate the foundation configuration, monoliths, joints, and shears and conceptualize the foundation rock wedge geometry under monoliths 10–11 and 22–23 (Figure 13). This model emphasizes the importance of pulling the foundation mapping into a 3-D exhibit to better understand the entire foundation as a system and visualize the potential stepping nature of the side plane joint network.

The kinematics and conditions of the joint sets and shear zones relative to the dam configuration and assumed downstream sliding direction, including the physical parameters of the base, release, and side planes forming the foundation rock wedge, are described next:

• Base Plane (shown as red dashed lines in Figure 13): The sub-horizontal clayey shears at GPD dip about 2–12 degrees into the left abutment and slightly up68

stream in the left abutment wedge (Kay shear) and >15 degrees in the right abutment shear (Big Kid shear). The shears are assumed to be continuous from upstream to downstream and clearly represent a possible weak base/sliding surface for a removable rock block in either the left (monoliths 22/23) or right (monoliths 10/11) mid-slope abutments. The orientation of these shears allows them to reach the surface at the top of rock surface and trace across the side slopes and drainage swales downstream of the dam. The shears also have relatively large-scale undulations (up to 5–10 ft over relatively short distances of 25 to 40 ft) associated with the irregular predepositional surface topography, resulting in anastomosing shear surfaces that provide some additional 3-D resistance to deformation.

• Release Plane (shown as green lines in Figure 13): The J2 joint set strikes sub-parallel to the dam axis (similar orientation as the faults) with a high continuity and is ubiquitous across the site and can form the back edge release surface near the heal of the dam for a rock wedge that displaces in a downstream

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Figure 14. Geologic foundation mapping showing abutment wedge configuration and stereographic projection of wedges.

direction. This joint set can also contribute to wedge side plane development as a connecting surface (smaller release planes, in tension) between en-echelon stepping J1 and/or J3 joints in either abutment.

• Side Planes (shown as orange and purple lines in Figure 13): The side planes for a removable rock wedge can only be formed by the J1 (orange) set in the left abutment and J3 (purple) set in the right abutment because the intersection angles of these joints with the base plane project downstream and can reach the surface in the downstream top of rock surface.

The J1 and J3 side planes require the combination with the other discontinuities to create an outwardstepping and potentially very rough and irregular side plane surface. The strike of the side plane is oriented at 56 degrees and 70 degrees to the dam axis for the left and right abutments, respectively, and trends 34 degrees and 20 degrees from the direction of monolith sliding (directly downstream), respectively.

The criteria for foundation rock wedge formation defined above have been met, and a kinematically removable wedge geometry appears to be present in both abutments under monoliths 10–11 and 22–23, as delineated in Figure 14, which shows the basic force vectors for deformation, and as described in the following section. (Note: While rock wedges may also be present under monoliths 5–6 and 26, only the wedges under 10–11 and 22–23 were evaluated in detail because of the larger loading conditions associated with those monoliths.)

Right Abutment: Monoliths 10–11 Based on this discontinuity evaluation, it is kinematically possible for foundation wedges to form with the intersection of the Big Kid shear (base plane) and the J3 joints (side plane). However, due to the limited horizontal and vertical continuity of the J3 joint set, the wedge must combine with the J1 and J2 joints to allow the side plane to step-out in an en-echelon configuration, resulting in a rough surface requiring rupture

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through some portion of intact rock. The trend and plunge of the line of intersection of the base plane shear and the side plane are oriented downstream and about 70 degrees to the dam axis inward toward the valley, and about 20 degrees from the direction of dam and reservoir loading (220 degrees) and assumed failure direction. Based on this geometry, the wedge appears to include monolith 11 and portions of monolith 10.

Based on the orientation of sliding direction and loading vector, a portion of the driving forces, estimated to be approximately 34%, will be applied normal into the side surface of the right wedge, significantly increasing the shear strength of the side plane.

Concrete back-filled drifts 1 through 10 were advanced parallel to the dam axis, following the Big Kid shear at the downstream toe of monoliths 10 and 11, and removing approximately 30% of the Big Kid shear from under the monolith 10–11 footprint. The concrete drifts help to disrupt the continuity of the base surface and force failure either through the concrete back fill or into the more intact rock mass above or below the drifts. This provides additional shearing resistance and strength to the base surface.

Left Abutment: Monoliths 22–23 In the left abutment, it is kinematically possible for a rock wedge to form with the intersection of the Kay shear (base plane) and the J1 set in combination with the J2 and J3 joints (side plane). However, due to the limited horizontal and vertical continuity of the J1 set, it must combine with the J3 and J2 joints to allow the side plane to step out in an en-echelon formation, engaging a highly rough surface and rupturing through some portion of intact rock. Based on the surface geologic mapping and interpretation from the construction photos, it appears that the Kay shear “ramps-up,” or exhibits a large-scale undulation immediately downstream of monolith 22 at the upstream extent of the stilling basin training wall.

The trend and plunge of the line of intersection of the shear and the side plane are 055 snd 005, dipping in an upstream direction and oblique to the axis of the dam. Therefore, the loading forces exerted on the wedge would result in a downstream direction and force it to move upward approximately 5 degrees along this intersection plunge line, which is a favorable configuration for stability.

Based on this geometry, the wedge appears to also include portions of monolith 23. The downstream wedge failure direction is oriented about 32 degrees into the valley from the direction of dam and reservoir loading and assumed sliding direction. Based on the orientations of the sliding direction and loading vector, a portion of the driving forces, estimated to be approx70

imately 55%, would be applied normal into the side surface of the left wedge, significantly increasing the shear strength of the side plane.

Approximately 15% of the base surface shears have been removed along Drifts K1 and K2 in the left abutment. The concrete drifts help to disrupt the continuity of the base surface and force failure either through the concrete back fill or into the more intact rock mass above or below the drifts. This provides additional shearing resistance and strength to the base surface.

This preliminary evaluation assessed the most likely and generalized rock wedge configuration in each abutment. It is recognized that other wedge positions, sizes, or shapes could form and should be further evaluated in more advanced stages of this analysis. The wedge geometry, base plane surfaces projected onto the downstream top of rock surface topography, and approximated positioning of mapped foundation joints and foundation geometry for a number of potential foundation rock wedges are presented in Figure 15.

Pore Pressures The performance of the double-line drain lines across the GPD foundation was evaluated during the stability analysis and was found to significantly reduce the head across the dam from upstream to downstream with a drain efficiency of nearly 80% based on data from uplift cells. The maximum potentiometric surface recorded across the site is above the foundation elevation and above the base plane shears in the abutments along the upstream line of foundation drains during normal dam operations, and the tail-water elevation is well below the base sliding surface of the foundation wedges, as shown in Figure 16. Of particular note, the downstream rows of drains at monoliths 10–11 and 22–23 were constructed from tunnel galleries excavated into the abutments and below the elevation of the base sliding plane, further dropping the potentiometric level below the wedges at these locations (see Figure 9 above).

Therefore, the uplift pressures acting on the wedge planes are expected to be relatively low. The potentiometric data provide a high level of confidence that it is unlikely that appreciable high pore pressures will be developed at a sufficient magnitude within the rock wedge foundation during a seismic event to contribute to dam-foundation system deformation.

However, a foundation sliding analysis assumes that only a minimal potentiometric head drop occurs across the dam foundation (approximately 30%, and up to 50% as a maximum), and that drains can be sheared during deformation, rendering them inoperable (USACE, 1969, 1995). Therefore, the stability analysis should conservatively assume that pore pressures can still be concentrated on the base, side, and release

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Figure 15. Conceptual foundation wedges mapped downstream across the foundation excavation and projected onto the top of rock topography.

planes of the removable rock wedge, but the actual uplift cell data and historically low pore pressures should be considered in interpreting the results of the analysis and assigning a probability of failure during the risk analysis.

FOUNDATION WEDGE STABILITY SUMMARY While there are geometrically capable and removable rock wedges positioned in the GPD foundation, and no numerical or robust kinematic quantitative analyses have been performed, the 3-D geomechanical conditions that were qualitatively evaluated appear to support the preliminary conclusion that the addition of seismic loading would not likely result in consequential deformations of the foundation rock wedges. The following points present the arguments for foundation stability developed to support this conclusion: • The limited horizontal and vertical continuity, and the spatial distribution of the J1 and J3 discontinuities, which together would make up the slide plane of the foundation wedges, result in outwardstepping surfaces. This configuration creates a rough, irregular, blocky/columnar side plane

surface that would require large-scale deformation through portions of intact rock and high-quality rock mass of the lava sequence, thus increasing shear resistance.

• The wedge geometry displacement direction for both the right and left abutment wedges would be oriented inward toward the valley and orthogonal to the dam loading/sliding direction and therefore not directly upstream-downstream. ◦ The dam sliding direction (220 degrees) is oriented directly downstream. The right wedge failure direction is 200 degrees, and the left wedge failure direction is 254 degrees, resulting in a 20 degrees and 34 degrees, respectively, inward angle toward the valley from the dam sliding and loading direction. ◦ The orientation of the side surfaces is orthogonal to the direction of sliding, and a normal force of 34% and 55% of the total driving force from the dam would be imparted on the side surface for the right and left abutments, respectively, significantly increasing the shear strength of the side planes. ◦ Some amount of dilation along both sliding surfaces must occur for deformation to occur, which

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Figure 16. Highest recorded potentiometric surface across monoliths 12 and 16 (inset images) and across the dam at the upstream line of drains. (Note: At the location of the foundation rock wedges under monoliths 10–11 and 22–23, the downstream line of drains was constructed from tunnel galleries that are spatially positioned below the elevation of the base plane; see Figure 9.)

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further increases the shear strength of the side planes.

The base shear surfaces tend to have large-scale undulations on the order of 5–10 ft over relatively short distances (25–40 ft), potentially increasing the strength of the sliding surface at the scale of the rock wedge and increasing the largescale asymmetric variability and resistance of the sliding surface and inducing wedge rotation.

The intersection of the side and base planes of the left abutment foundation wedge dips slightly upstream and into the left abutment. The rock wedge must be displaced at an angle of about 5 degrees upward to move in a downstream direction.

The right abutment has a buried top of rock ridge immediately downstream of the dam oriented into the valley that will significantly buttress and/or extend a foundation wedge.

Based on the triangular-shaped foundation surface area of the rock wedge, it is highly likely that the wedge displacement will include multiple monoliths and portions of concrete monoliths, forcing significant monolith interaction and mo-

bilization of 3-D effects within the dam structure itself. ◦ Significant rotation, torsion, and interlocking within the foundation wedge will result in added 3-D strengthening of the structure and improved resistance.

• Data suggest that low uplift pressure is acting on the foundation rock in the downstream portion of the dam, as the grout curtain and drain efficiency is approximately 80%, and the potentiometric surface is not significantly above the evaluated shears during normal dam operations.

• Focused attention was given during construction to potential sliding and design of shear keys to increase sliding resistance. ◦ Foundation treatment included construction of concrete back-filled drift reinforcement under monoliths 10–11 and monoliths 22–23, where portions of base sliding shears were removed, making them discontinuous under the dam. ◦ Excellent, very high-quality geologic mapping was completed during construction, which provides an equivalent level of confidence in the

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qualitative evaluation of the site geology and geomechanical conditions. â—Ś The designers were very experienced and concerned about potential foundation sliding and made the treatment and reinforcement a focus of the design considerations.

CONCLUSIONS Following the compilation of this data set and geomechanical and kinematic evaluation of the site conditions, the USACE project team developed a 3-D concept of the mechanics and conditions defining the stability of rock wedges positioned in portions of the GPD foundation given the anticipated probabilistic seismic loading conditions. This qualitative evaluation helped to communicate the three-dimensionality of the foundation wedge to the project team, stakeholders, and reviewers; provided insight into the 2-D stability results; and better informed our risk assessment team during failure mode elicitation. When foundation sliding mechanisms represent a potential instability issue for a concrete gravity dam, the project engineering team needs to interpret the traditional 2-D stability modeling results based on qualitative 3-D conceptualization of the actual site-specific failure mechanics and kinematics. A more robust 3-D analysis of the foundation stability is planned to verify the conclusions developed in this preliminary work and to evaluate different wedge sizes and positions in the foundation under anticipated seismic loading conditions.

The value of implementing this simplified geological evaluation utilizing available plans, sections, mapping, reports, and discontinuity analysis cannot be over-stated. This conceptual study provided a better understanding of the failure mechanics and kinematics of the dam-foundation system, and similar evaluations should be performed prior to developing any computer-based 3-D models or performing stability analyses. This preliminary step should not be ignored or under-valued by practitioners in performing gravity structure stability analysis. The methodology presented in this paper for the GPD was invaluable in helping the team to understand the dam-foundation system, reduce the uncertainties in the risks posed by this failure mode, and interpret the relative applicability of the 2-D limit equilibrium stability analysis results. A geologic evaluation should be performed and documented at the early stages of any foundation stability analysis performed for concrete gravity structures.

ACKNOWLEDGMENTS The author would like to express his sincere appreciation to the GPD risk assessment and project devel-

opment team members and is grateful for the opportunity to work through this assessment and contribute to the overall understanding of geomechanical conditions in the foundation at GPD. Special thanks go to Gabriel Lyvers, Josh Corbett, Gregg Scott, Peter Shaffner, Tres Henn, David Scofield, and Jacob Nienaber for performing the significant work of compiling the site characterization report and documentation from which the data were derived, providing insight and direction during the implementation of the foundation evaluation, and delivering a final review of the paper.

REFERENCES ELECTRIC POWER RESEARCH INSTITUTE, 1992, Uplift Pressures, Shear Strengths, and Tensile Strength for Stability Analysis of Concrete Gravity Dams, Volume 1: Electric Power Research Institute (EPRI), Palo Alto, CA, Report TR-100345, Volume 1, Project 2917-05.

GALIC, D.; GLASER, S. D.; AND GOODMAN, R. E., 2008, Calculating the shear strength of a sliding asymmetric block under varying degrees of lateral constraint: International Journal of Rock Mechanics and Mining Sciences, doi:10.1016/ j.ijrmms.2008.01.001.

GALIC, D.; GLASER, S. D.; GOODMAN, R. E.; AND BRO, A., 2008, Laterally controlled shear testing of 1:200 scale model gravity dam monoliths over a foundation with three-dimensional interfacial roughness: American Rock Mechanics Association, ARMA 09-083, 43rd U.S. RockMechanics Symposium, Asheville, NC, June 28-July 1, 2009.

GOODMAN, R. E. AND BRO, A., 2005, Shear strength of a foundation with two dimensional roughness: American Rock Mechanics Association, ARMA 05-799, 70th U.S. Rock Mechanics Symposium, Anchorage, AK. June 25-29, 2005.

LONDE, P., 1973, Rock mechanics and dam foundation design. In International Congress on Large Dams, Paris, CIGB/ICOLD. NESBITT, R. H. AND CORNS, C. F., 1967, Sliding stability of three dams on weak foundations. In Ninth Congress on Large Dams, Istanbul, Q.32, R.29.

NICHOLSON, G. A., 1983, Design of Gravity Dams on Rock Foundation: Sliding Stability Assessment by Limit Equilibrium and Selection of Shear Strength Parameters: U.S. Army Corps of Engineers, Waterways Experiment Station, Technical Report GL-83-13.

SCOTT, G. A., 1999, Guidelines Foundation and Geotechnical Studies for Existing Concrete Dams: U.S. Bureau of Reclamation, Denver, Colorado.

SCOTT, G. A. AND VON THUN, J. L., 1993, Interim Guidelines Geotechnical Studies for Concrete Dams (Draft): U.S. Bureau of Reclamation, Denver, Colorado.

SHERROD, D. A. AND SMITH, J. G., 2000, Geologic Map of Upper Eocene to Holocene Volcanic and Related Rocks of the Cascade Range, Oregon: U.S. Geological Survey Map I-2569.

U.S. ARMY CORPS OF ENGINEERS (USACE), Portland District, 1969, Green Peter Dam, Middle Santiam River, Oregon, Foundation Report: USACE Design Memorandum, Portland, Oregon.

U.S. ARMY CORPS OF ENGINEERS (USACE), 1994, Rock Foundations: USACE Engineer Manual EM 1110-1-2908.

U.S. ARMY CORPS OF ENGINEERS (USACE), 1995, Gravity Dam Design: USACE Engineer Manual EM 1110-1-2200.

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Lessons Learned from USACE Seepage Barrier Wall Construction: Wolf Creek to Present GEORGETTE HLEPAS Geotechnical and Geology Section, Dam Safety Modification Mandatory Center of Expertise, U.S. Army Corps of Engineers

VANESSA BATEMAN∗ Civil Design Branch, Engineering and Construction Division, Nashville District, U.S. Army Corps of Engineers

Key Terms: Dam Safety, Barrier Wall, Seepage Cutoff Wall, Wolf Creek, QA Verification, Data Management ABSTRACT The U.S. Army Corps of Engineers (USACE) maintains a lessons-learned goal for all major projects to capture knowledge gained. The focus of the formal lessonslearned process is to share knowledge and experience nationwide improving USACE contracting methodologies, reducing overall costs, and improving designs. This continuous improvement can be seen in the evolution of USACE barrier wall construction designs and contracting methods. From the first Wolf Creek Dam barrier wall installed in the 1970s to the more recent Bolivar and East Branch Dam barrier wall projects, documentation and sharing of lessons learned in areas such as grouting, data management, and quality assurance procedures have increased the efficiency and effectiveness of barrier wall designs, monitoring, and contract specifications. Contractual philosophy, use of pre-grouting treatment, verification methods, and data management processes have all changed due to lessons learned and have enabled the USACE to improve the overall end product of barrier wall projects.

INTRODUCTION The first major seepage barrier wall installed through a dam by the U.S. Army Corps of Engineers (USACE) was at Wolf Creek Dam, located near rural Jamestown, Kentucky, in the 1970s. This was the first of two barrier walls to be installed at the project. The construction was initiated due to the appearance of sinkholes on the downstream side of the dam, muddy flow, and other signs of distress. This indicated a developing internal erosion problem that was through and along the karst foundation (Simmons, 1982; Zoccola et al., ∗ Corresponding

author email: vanessa.c.bateman@usace.army.mil.

2009). Figure 1 shows the general geological setting of the project, including the large solutions features, particularly those near Section 35 + 00 that caused much of the dam safety concern that presented technical challenges during construction of both Wolf Creek barrier wall projects.

The original wall was installed after a period of intense grouting at the site initiated immediately after the muddy flow of 1968–1969 revealed the presence of sinkholes. Figure 2 shows the location of these distress features and grouting completed prior to barrier wall construction. Only grout lines 1 through 3 were located near the crest of the dam, and these targeted the solution features near station 36 + 00 that were called out later in the project as the “cave and core trench area” and also as “critical area 1 (CA1).” Grouting was initiated at the project as an emergency response to distress features, but then as now, grouting was not considered to be a sufficient permanent solution in a karst environment. Intense grouting during this first phase of wall building was focused primarily downstream of the dam. Grouting at successive wall projects was focused primarily in the areas where a barrier wall was to be constructed.

Unfortunately, the original barrier wall at Wolf Creek, due to technical limitations and trade-off decisions made at the time, proved to be neither deep enough nor long enough, and a second barrier wall was installed starting in 2010. Figure 3 shows the location of the original Wolf Creek wall (often called ICOS wall) and the second Wolf Creek wall along profile. As future seepage barrier walls have been designed and constructed, lessons learned have been captured and shared by USACE personnel in an effort to improve the efficiency and effectiveness of contracting methods, design criteria, verification methods, and data management.

One clear lesson learned from Wolf Creek is that close attention needs to be paid to the barrier wall limits to ensure that the wall is sufficiently sized.

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Figure 1. Wolf Creek Dam geologic section—view looking upstream (Kellberg and Simmons, 1977).

Further, later work demonstrated that there are increased hydraulic gradients at the margins of a wall.

This can cause increased seepage leading to ineffectiveness of an insufficiently sized wall (Rice and Duncan, 2009). Thus, with the example of Wolf Creek and also the later Walter F. George project, increased design attention on subsequent walls has focused on the length, depths, and appropriate tie-ins, particularly for walls located in karst.

Several subsequent USACE cutoff wall construction projects are listed in Figure 4. The geology of their respective foundations differs; however, the purpose of the barrier wall is the same: to reduce the risk associated with internal erosion of the embankment through the foundation.

While some means and methods have been nearly constant, others have vastly evolved with project experience and technological advances that have made it possible to monitor and document projects differently than in the past. Contracting methods have generally progressed from more prescriptive to more performance-based specifications, but none of the walls reviewed to date implement fully performance-based specification. The introduction of the “Best Value” contracting method in the Federal Acquisition Regulations 76

(FAR) has assisted this evolution. A contract requirement for “demonstration sections” has also been somewhat inconsistent but has generally been considered to be a positive benefit to projects.

Pre-grouting prior to barrier wall construction was not typically required for the first three decades of USACE barrier wall projects. However, lessons learned from the slurry loss during the Mississinewa barrier wall construction (2000–2005) indicated that pregrouting was a good standard approach, and this practice has continued (USACE, 2001). Pre-grouting has also proven to be an effective means of exploration and a method to verify barrier wall design assumptions. It has been particularly effective for evaluating depth and length of barrier walls.

Computers, automated monitoring equipment, and Geographic Information Systems (GIS) programs have all had major technological advances since USACE began barrier wall construction. These advances have been used on more recent projects and are showing significant benefits to USACE for both verification of construction and long-term documentation. This paper provides a sampling of the technological improvements and impact on USACE construction verification and data management requirements.

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Figure 2. Grouting completed at Wolf Creek Dam prior to the barrier wall construction.

LESSONS LEARNED: CONTRACTING PHILOSOPHY The 1970s-era Wolf Creek barrier wall was contracted somewhat differently more recent USACE contracts. While a prescriptive specification was used, as will be discussed below, it was applied in a somewhat unusual form. More recent projects have gravitated to a more performance-based specification, culminating with a hybrid performance/prescriptive specification developed along with a Best Value contracting method that allows a contractor to propose, within certain limits, new means and methods of construction. The contracts are awarded based on an evaluation of the technical merits of the contractor proposals. Price permitting, the trade-off among price and technical factors permits an acceptance of a proposal that is not the lowest bid (FAR Subpart 27.404-1). Contractor submittals are ultimately required during construction that include design details, calculations, and methodologies supporting their initial proposal. These submittals must be in compliance with contract specifications; otherwise, a variance is required. Typically, submittal review and approval is completed prior to commencement of each phase of work. While there are some problems with this

approach, as design by submittal also has pitfalls, the advantages of allowing trade-off analysis in contract award has thus far been judged by USACE to be well worth the disadvantages.

Wolf Creek Dam: Unusual Contracting Method Accomplishing Similar Goals The first Wolf Creek barrier wall contract was unusual in that it combined a fully prescriptive contract type with a request for proposals on barrier wall construction designs and techniques (USACE, 1974). Instead of a Best Value contracting method, which was unavailable at the time, it achieved much of the same goals by dividing the project into two phases with several contracts. At the time, USACE did not have much experience with seepage barrier wall construction and was reluctant to solicit proposals based on a prescriptive specification, as was common for most USACE construction projects at the time. As a result, contracting was performed in two phases. In the first phase, a request for technical proposals for a design approach contract was made. Multiple contractors submitted designs, and these were evaluated by USACE along with

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Figure 3. Location of features on the second Wolf Creek barrier wall, including original ICOS wall—view looking downstream.

its Board of Consultants. Two qualified contractors’ proposals were accepted in 1974, and then USACE proceeded to write a prescriptive specification based on these proposals, selecting the secant pile method as the preferred wall construction technique. Thus, while it was a prescriptive contracting method in the end, the contractor awarded the project was essentially bidding on its own technical proposal. This second contract phase was also broken into two parts by barrier wall location (USACE, 1988).

Figure 4. Selected barrier walls constructed by USACE, 1974 to present.

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Best Value Contracting Since the first Wolf Creek barrier wall, USACE has continued to implement more performance-based specifications while still incorporating prescriptive specifications for many elements of the contract. The acceptance of Best Value contracts within the FAR implemented the process of source selection of technical proposals, thus removing the need to let multiple contracts for design and then construction. This method of contracting was allowed by law in the Competition on Contracting Act passed in 1984; thus, all projects completed after Addicks and Barker had current Best Value contracting methodology available. This contracting method has been used on a number of USACE projects, including Mississinewa Dam (2002–2005) and, more recently, Wolf Creek Dam (2008–2014), Center Hill Dam (2012–2015), Bolivar Dam (2014–2016), and East Branch Dam (2015–present).

Rather than prescribing all the means and methods of construction, USACE relies on specifying the geometric minimum extents (effective width, continuity, verticality, length, and overlap of wall elements), required material properties (homogeneity, strength, and

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Figure 5. Location of Wolf Creek walls with contract sections of the original wall identified.

permeability), quality assurance/quality control requirements, and construction verification procedures.

This quasi–performance-based specification method encourages contractors, within certain limits, to propose the means and methods to achieve the design requirements using their distinctive equipment and installation techniques. This pattern has continued in recent contracts, with extensive specifications centered on materials, required geometry, and construction verification/monitoring.

Proposals have been evaluated using the Best Value trade-off method of contract acquisition. By permitting the contractor to propose their means and methods of excavation, embankment stabilization, and backfill mix design, the USACE has opened the door to innovative approaches while maintaining competitive bidding.

This philosophy has been relatively consistent over the past decade, and its continued use in successive contracts indicates USACE acceptance of the methodology as the best currently available contracting method for seepage barrier wall construction.

Demonstration Sections The 1970s Wolf Creek barrier wall contract was ultimately constructed in two phases of work. The first phase included a wall near the downstream switchyard and a 302-m (989-ft) wall along the dam axis starting at the left abutment of the dam (at the concrete interface).

The 302-m wall provided approximately 44 percent of the length of the wall called for in the design. The second phase, contained in a second contract, installed the remainder of the barrier wall length of 381 m (1,250 ft).

This two-phase process effectively provided a means of a test or demonstration section of the project, which ultimately was the final wall. This two-phase approach made the second phase of construction a smoother process, as early construction issues had already been

identified and corrected (USACE, 1988). See Figure 5 for location of the original barrier wall showing these two construction phases.

In retrospect, the order of this should have been changed with the rightmost section of the wall completed first, thus testing the methodology in the less geologically difficult area. This philosophy was followed with the second Wolf Creek barrier wall, as the demonstration section (called the “technique area”) was included near the right end of the wall; see Figure 3 for the location.

Later philosophies on demonstration sections indicated that it is preferable to install a much smaller test section to work out construction issues and develop consistent procedures to be used for the barrier wall construction. Recommended practice is to include a time delay in the contract between submission of this technique area report and start of production areas of the wall. This has been successfully executed on a number of USACE projects. These have sometimes been located completely outside of the new wall area or, as with Wolf Creek, near a less consequential end of the wall.

The Center Hill project is a recent exception to this trend. Due to space limitations, the demonstration sections were incorporated into the middle of the final wall, as shown in Figure 6 as the technique area. While the overall project was successful, the inclusion of the demonstration sections in the middle of the final wall proved to be problematic in both construction monitoring and overall construction schedule. Changes in methodology of concreting operations essentially extended the original technique areas to nearly one-third of the wall.

Another complication of blending the demonstration section into main wall construction is that the resultant verification program was more extensive than originally envisioned by USACE. It also caused

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Figure 6. Profile view of Center Hill barrier wall showing extended technique areas.

difficulties with managing the contract, as the technique area report provided by the contractor did not actually contain the final methodology for concrete placement in the barrier wall elements. Another lesson learned from the Center Hill project was that the schedule needs to incorporate enough time between the completion of the technique and the commencement of production of the wall. The intent of the contract was to have a technique area report produced at completion of the technique areas and for the government to have sufficient time to review; in reality, this “bleed over” caused by time pressure to get to barrier wall completion did not allow for the time envisioned by the original designers (USACE, 2017d).

A lesson learned for future projects is that these demonstration areas should be located in a non-critical part of the final wall and should not be installed in the most difficult geology or in the middle of the wall, which could potentially cause sequencing issues. The second Wolf Creek wall implemented this effectively by including a deep section on the right side that was not needed for barrier; rather, it was used to demonstrate the effectiveness of the contractors’ means and methods.

GROUTING AS PRE-TREATMENT Foundation Grouting Pre-Treatment The 1970s-era Wolf Creek barrier wall was installed because it was determined that grouting would not provide a permanent and positive barrier through karst limestone foundations. The philosophy was that grouting alone would not replace soil infilling in the solution features. Over time, soil infilling has the potential to erode, leading to increasing seepage paths. If grouting pressures were sufficiently applied to remove infilling, these pressures would also have a high likelihood of 80

causing hydrofracturing in the dam and, thus, are not allowed on typical USACE dam safety modification projects.

As can be seen in Figure 2, most of the grouting at Wolf Creek prior to the first wall was performed as a remedial measure to prevent potential failure of the dam, and it was not installed to prevent slurry loss during wall construction as is common on later projects. There have been multiple discussions and disagreements over the years on the necessity of a grouting program prior to wall installations. The original Wolf Creek wall used steel casing on the outside of the primary pile elements, and only the secondary element was excavated into rock without casing (see Figure 7). As only a very short section at the crest of the dam was grouted, the majority of the wall was installed without pre-grouting, and it was difficult to monitor at the time, no major slurry losses were noted (USACE, 1988). Subsequent projects therefore, did not necessarily focus on grouting as a means to control slurry or water loss into the foundation.

This changed somewhat in the early 2000s with Mississinewa Dam, which was also constructed on a karst foundation (USACE, 2001). Due to the perceived difficulties in grouting, a pre-grouting program like that used at Center Hill and the later Wolf Creek wall was considered but not included. During panel installation in the demonstration section, however, there was a large slurry loss. The initial loss was approximately 115 cubic meters (150 cubic yards, or 30,000 gallons) of slurry.

Following this event, there was a subsequent loss of approximately 300,000 liters (80,000 gallons) of bentonite slurry, 60 hay bales, and 460 bags of bentonite.

Approximately 160 cubic meters (208 cubic yards) of sand were eventually placed in the barrier wall panel (Schaeffer 2011). As a result, a grouting program was implemented at the project with a target permeability of 10 Lugeons; a slurry loss on the order of 50–100 gpm was considered to be manageable. Mississinewa

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Figure 8. Typical layout of grouting to support barrier wall construction. Example from Center Hill Dam, 2009, before barrier wall construction began.

Figure 7. Typical plan (top) and section (bottom) of the original Wolf Creek wall. Secondary elements were excavated into rock under slurry.

was the first USACE project to use a combined grout curtain and barrier wall solution (Stare et al., 2013).

Pre-grouting, or foundation grouting, programs began to be included as part of the design for future projects, such as the 2010 barrier wall installation at Wolf Creek and Center Hill. Both of these projects contracted the foundation grouting program separately from the barrier wall installation. The philosophy at the time was that grouting would provide more immediate risk reduction and better define the subsurface conditions while the specifications for the barrier wall were finalized and the contract bidding processes completed. Both grouting contracts were awarded using an Indefinite Quantity Indefinite Delivery (IDIQ) method of contracting. This approach permitted the ability to gain detailed subsurface information and optimize the barrier wall design extents prior to soliciting and awarding the barrier wall contract. Figure 8 shows a typical configuration of grouting when used to support barrier wall construction from Center Hill Dam. However, contracting the grouting program separately from the barrier wall contract came at the cost of additional labor expenses to develop and award two separate contracts and an increase in the overall sched-

ule. It was also not the most efficient means of procuring grouting, as the secondary contractor is not responsible for the previous contractor’s grouting quality and the grouting program may not be optimized for the means and methods of barrier wall construction. This became more of an issue at Wolf Creek, where the grouting was not completed under the IDIQ contract (USACE, 2017a, 2017b). The Contractor at Center Hill dam added additional exploratory borings at each panel location in order to verify that the grouting of the previous contractor was sufficient to prevent/reduce the possibility of slurry loss (USACE, 2017c, 2017d). Furthermore, if the grouting and barrier wall are broken into two contracts, there is a risk of having multiple contractors on-site at the same time and same location. This can impact schedule sequencing, especially on projects with work space limitations.

The lesson learned from these experiences was that it may be better to include both grouting and barrier wall installation in one contract. This was not done at these projects due to concerns of the length of time it would take to execute a full grouting and barrier wall contract.

At both Center Hill and Wolf Creek, the grouting information was used extensively to evaluate the proper geometric extents of the wall. During a risk assessment process that occurred at Center Hill, the length of the wall was altered to include a section of wall with an extended depth. The basis of that decision was the geological and grouting information that would not have been available had grouting not been performed (USACE, 2015).

The Bolivar Dam barrier wall modification included a partial depth (i.e., hanging) wall that did not extend to rock; however, there was a requirement for grouting of the rock at the left abutment. Based on the lessons learned at Mississinewa, a pre-grouting program was included in the design. Based on the experiences at Wolf Creek and since the grouting of the abutment

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alignment intersected the barrier wall installation, the grouting program and barrier wall installation were included in the same contract. This reduced the risk of issues with having two separate contractors on-site at the same time. It also provided an effective means for altering the length of the barrier wall into the left abutment, if the grouting program dictated that it was required, by a simple contract modification.

East Branch Dam, like Wolf Creek, Center Hill, and Mississinewa, has a barrier wall design that extends through the foundation rock and was designed to intercept discontinuities and voids (albeit not in a karst environment). Similar to Bolivar and based on the lessons learned from Wolf Creek, Center Hill, and Mississinewa barrier wall installations, the contract specifications included pre-grouting of the foundation and barrier wall construction in the same contract. A final evaluation of the success of that project cannot be completed at this time, as the demonstration section has not begun. However, by including the grouting and barrier wall in the same contract, the contractor that was awarded the project had involvement in foundation geology exploration and grouting results. Thus, costs associated with scheduling two separate consecutive contracts and the potential for schedule impacts with transition from one contractor to another are expected to be avoided.

Whether or not pre-grouting is contracted separately from the barrier wall contract or within the same contract, pre-grouting provides the subsurface information needed to verify that the barrier wall design extents meet the design intent. It also provides a means of reducing the risk of slurry loss in cases where slurry is used. It is noted that both Bolivar and East Branch did not have similar geology and that the amount of grouting was relatively limited, and therefore the projects are not entirely analogous. Nonetheless, the lessons learned are that pre-foundation grouting provides incredibly useful information in guiding the design extents of barrier walls. The need for and contracting methods for the pre-foundation treatment need to be carefully weighed based on the advantages and disadvantages of various contracting mechanisms and the technical requirements of the project.

BARRIER WALL VERIFICATION REQUIREMENTS In order to verify the barrier wall construction, typical USACE contract specifications require the contractor to demonstrate the following: 1) barrier wall geometry (depth, length, width, verticality, overlap) 82

Figure 9. Adjacent panels not meeting continuity requirements of minimum width and overlap due to twist or tilt of panel.

2) barrier wall joint tightness (at bedrock and interpanel/pile interface) 3) backfill material properties (homogeneity, strength, permeability) Specific tools used to verify the final geometry may vary depending on the shape of the barrier wall elements being used at a particular site.

Geometry One of the main features of the barrier walls include geometric extents and verticality. Methods of verifying barrier wall geometry are generally specified to be supplied by the contractor. The barrier wall width (perpendicular to the axis) is readily verified, at least by neat-line tool dimensions, when the excavation equipment meets or exceeds the width of the excavation.

The neat-line tool dimensions measure the actual position and dimensions of the tool and where it passes through the excavation. Depth can readily be measured by excavation equipment and verified subsequently by soundings. However, verifying continuity of minimum width and that minimum overlap between adjacent elements throughout the depth of the excavation is met is a more difficult proposition.

Where an element deviates too far out of alignment or, as with rectangular elements, too much twist occurs, openings can be left in the barrier wall between elements. This can also result in a reduced thickness even when there is no opening. Figure 9 is an example of the continuous width and minimum overlap not being met due to twist or tilt of adjacent elements.

Experience from the Bureau of Reclamations Fontanelle Dam in the 1980s indicated that three methods to ensure verticality were needed. The bureau’s logic in including three methods was that there could be disagreement between two methods of verticality measurement and that a third method would be required as a tiebreaker. However, measurement methods and equipment have improved since that time. While implementing multiple methods for element geometry verification has benefits, some methods are inherently more accurate than others.

The 2000s Mississinewa Dam specifications did not require three methods of element geometry verification.

Nevertheless, three methods were implemented in the

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field. The specifications called for the excavation equipment as a primary measurement method and the Koden DM-682/684, or equal, as the secondary method.

The excavation equipment measures neat-line tool dimensions, while the Koden utilizes an ultrasonic bidirectional echo sensing system to record the distance between itself and the walls of the excavation. As the Koden probe is lowered by winch cable into the excavation, measurements are recorded. There are, however, limitations to the distance the device can achieve a returned a signal and record a measurement. The third method that was executed in the field was an inverted pendulum, which is a pendulum that has its mass located above the pivot point rather than below.

The 2010s Wolf Creek barrier wall specifications initially required two methods of verticality measurement, while Center Hill required three. Multiple methods included neat-line excavation dimensions determined by excavation equipment equipped with inclinometers, and through the use of a down-the-hole probe system (which measures neat-line dimensions as well as the orientation of the pilot hole at depth), measurements of pilot holes for circular elements were used at Wolf Creek in addition to the Koden. Once the deviation of the pilot hole was measured with the down-the hole-probe, CADD software was used to project the dimensions of the barrier wall element for comparison to the other survey methods. Excavation tools equipped with inclinometers as well as the Koden were also used at Center Hill. The contractor at Center Hill also implemented what it termed the Cable Inclination System (CIS). The CIS is a system used in conjunction with the onboard inclinometers in the hydromill tool to correct for actual position. Survey measurements are taken at points along the cable in order to provide an exact position in space of the hydromill tool. Center Hill also used the Sonic Caliper to measure the actual excavation dimensions. The Sonic Caliper method provided excellent results in circular excavations but very poor results in rectangular excavations due at least partially to the computer processing assumptions on the shape of the excavation. As most of the barrier wall was constructed of rectangular wall elements, its use was ultimately abandoned. Wolf Creek and Center Hill verified excavation with a free-fall pass of the tool in the excavation after it was complete, using the inclinometers within the tool (USACE 2017b, 2017d).

Analysis of the data at Center Hill showed that there was significant benefit to having more than one method for measuring verticality, and the Koden device gave the most accurate measurement of actual verticality and excavation dimensions for both rectangular and circular barrier wall elements. Rectangular panels whose horizontal extents surpassed the abilities of the Koden to measure were measured by position-

ing the device at multiple locations within the excavation. Ultimately, the down-the-hole orientation system, inclinometers in the excavation tools, and the CIS were used to verify the Koden measurements instead of being the tiebreaker. Where there was disagreement, measurements were retaken using the Koden. Functionally, even though the specifications required three methods, the project ended up with two methods of verticality measurements, as both the free-fall pass and the CIS made use of the onboard tool inclinometers (USACE 2017d).

Bolivar Dam project specifications heeded the lessons learned from Center Hill and Wolf Creek and required a minimum of two proven technologies for verifying verticality. The down-the-hole orientation system and the Koden (or other, similar technology that has been proven to be reliable) were included as options in the specifications. Ultimately, the contractor proposed the use of the automated monitoring equipment of their excavation tool and the Koden for verticality measurements (USACE 2014a).

East Branch Dam specifications did not reduce the number of required measurements for verticality to a minimum of two like those at Bolivar Dam did. Instead, they required the contractor to propose three methods and offered additional potential verticality measurement methods of “including, but not limited to Automatic Monitoring Devices on the Contractor’s Excavation Equipment, Optical Surveys, Koden DM-604, and Loadtest SoniCaliper devices” (USACE 2014b).

These specifications identified specific additional options, such that if the Koden was not selected as one of the three methods, there would be sufficient proof that verticality was met. The Koden was chosen for the East Branch project; however, its success on the project cannot be verified until construction is complete.

Joint Permeability In the 1990s, Mud Mountain Dam employed colored primary elements and non-colored secondary elements to distinguish the joint conditions at depth. The use of dye within the concrete mix of the primary panels permitted the ease of visually identifying the condition of the joints, as shown in Figure 10. This became a recommended standard practice for the USACE and was employed in the 2010s Wolf Creek and the Center Hill barrier walls. This was of significant benefit to both projects, as some joints were so tight that it was difficult to distinguish between primary and secondary wall elements in the areas where the colored concrete was not used. East Branch Dam also included the requirement of concrete coloring specifically to identify the interface between two adjacent barrier wall elements.

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Figure 10. Photo of verification core at barrier wall joint depicting contrasting colored backfill material of adjacent elements (March 31, 2015).

Uncolored concrete was specified for primary panels and a minimum of two other colors with an appropriate level of contrast. Neither Center Hill nor Bolivar Dam incorporated the colored concrete as a requirement in the original specification; however, during the construction project, the contract was modified to include the requirement.

Backfill Material Properties Backfill material properties are generally verified by two methods. One method is testing of the batch material prior to backfilling the excavation, and the other is testing of the in-place barrier wall. In the case of the 1970s Wolf Creek barrier wall, material properties were verified by performing unconfined compressive testing on the batch concrete. Mississinewa Dam required both batch testing prior to placement and verification core drilling within concrete elements and at concrete joints.

Cores were visually inspected for homogeneity, and unconfined compressive testing was performed. The verification hole was used for water pressure testing and included both a flow test and a duration test. Acceptance or rejection criteria based on these pressure tests were not included in the specifications.

By the 2010s, Wolf Creek barrier wall project, verification methods were changed substantially from those at Mississinewa. This was likely due to the fact that the Mississinewa specification was prescriptive when it came to the backfill materials, while the 2010s Wolf Creek was more performance based. In addition, technological advances provided more resources to verify that construction specifications were met. Wolf Creek required the contractor to design the concrete mix to meet criteria, such as minimum 28-day strength, temperature, slump, and air content (USACE, 2017b). The Wolf Creek concrete was required to have batch testing at the batch plant and on the platform prior to placement as well as verification drilling and testing.

There were strict time limits on use of the concrete from initial batching to placement. Verification core was sampled and tested to ensure that it met minimum strength requirements. However, water pressure testing 84

was not required or employed, as there was concern that the pressure test could negatively impact the wall.

Alternatively, technological advances permitted the less risky method of inspecting holes with optical televiewers (OPTV) and closed-circuit television (CCTV) video camera technologies. The OPTV and CCTV were vital tools for assessing the in situ concrete quality on the project and gave much better results than coring alone. Lessons learned were that coring periodically indicated incorrectly that concrete material was of poor quality and did not meet homogeneity requirements or that there were voids in the wall. However, OPTV and CCTV revealed that in many of these cases the inplace concrete met specifications and that the drilling techniques resulted in damage to the core.

The contractor was also required to propose a method to calculate the permeability of the barrier wall, which was not a requirement at Mississinewa. Both falling and rising head permeability tests were ultimately performed at Wolf Creek to verify that concrete permeability requirements were met (USACE, 2017b).

Center Hill Dam, Bolivar Dam, and East Branch Dam subsequently followed the same verification criteria as the 2010s Wolf Creek wall requirements except for final methodology of permeability testing. This included the batch testing of materials at the batch plant and on the platform prior to placement as well as verification coring of the placed material for visual inspection of the core; compressive strength and permeability test results on the core samples obtained; use of OPTV to verify homogeneity within the hole and to identify any voids, segregation, or leakage within the verification hole; and falling/rising head tests within the holes.

Center Hill used a different methodology for calculating the permeability of the barrier wall than Wolf Creek, and this was a source of some controversy on the project. The main source of the controversy was that the boundary conditions of both methods are somewhat different than an actual installed barrier wall.

This was recognized during the specification process, but there was a need to have a “yardstick” with designated value to assess barrier wall permeability in order to provide a clear evaluation guideline during construction. The Wolf Creek contractor proposed the Lefranc method, though there was extensive discussion on whether to accept this method, and Center Hill used Hvorslev, method G (Hvorslev, 1951; Canada–Bureau de Normalisation du Qu´ebec [CAN/BNQ] 2008; and USACE 2017b, 2017d). The different calculations can give somewhat different answers, though both projects were judged during assessments to have provided acceptable permeability results. As both Wolf Creek and Center Hill projects used a permeability of 1 × 106 cm/s in the specifications, a lesson learned is that designers should always be aware of the differences in the

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methods and carefully assess what basis will be used to judge the permeability placed in the specification.

DATA MANAGEMENT SOLUTIONS The 1970s cutoff wall construction was laboriously documented by hand-drawn sketches and tables due to the equipment available at the time. The early 2000 Mississinewa method of managing data was more computer based, but management of files incorporated primarily hard copies of information and computerized files in multiple locations and in a variety of formats.

Mississinewa was the first major USACE barrier wall project to use automated grout monitoring (Stare et al., 2013). The same was true of the grouting projects in preparation for wall construction at Center Hill and Wolf Creek. These grouting projects generated large quantities of data files that were organized by hand into folders. This was a fundamental change in managing a project due to the volume of data.

Neither the Wolf Creek nor the Center Hill project in the grouting phase was prepared for the quantity of data, nor had either the government or the contractor fully adapted to the new requirements. For example, the grouting data at Center Hill consisted of six grout lines with more than 1,300 holes with over 10,000 unique stages. The file library for the raw grouting files was more than 20 gigabytes in size. Task orders written for these IDIQ contracts were not written with access to the raw data and database in mind. The contractor and government eventually came to an agreement, and the government received those raw data files in an Excel format. However, the quality assurance staff at both the Wolf Creek and the Center Hill projects manually typed the data into Excel files and later incorporated them into an Access database. This was a considerable duplication of effort on-site, as most of those data had already been gathered by an automated grout monitoring program (USACE, 2017a, 2017c).

Thus, it was obvious to USACE that updated contracting language and methods for dealing with the volume of data were needed. While it was too late to incorporate into the original specifications, the cutoff wall contract at Wolf Creek was modified to include a data management solution via a Request for Proposal after the original contract was awarded. Using a Structure Query Language database, File Transfer Protocol site, and GIS to visually display the data, the Wolf Creek information management system was used to store and organize virtually all the project cutoff wall construction records. It was used to visually display progress at the site and was used extensively by the government to query and display and analyze project data. The GIS model included both plan and profile views of the cutoff wall, site geology, site instrumenta-

tion, and resultant verification test results. This system was also used extensively in post-construction assessments of the project. The pool raise meeting, to decide on an interim pool raise, occurred 6 days after the completion of the last barrier wall pile, a feat that would not have been possible without a well-organized data management system. Having the project records in a database made querying various results more effective and efficient. Figure 11 illustrates the type of data that can be pulled and compiled quickly.

The visualization with the GIS provided a means to readily assess the project status as well as evaluate the acceptability of each barrier wall element. For example, if an area of concrete was segregated or if a falling head test indicated that permeability requirements were not achieved, the location of the failing results were overlaid on the existing geology and evaluated in relation to neighboring test results. This permitted the ability to determine the impact of a failing result on the overall performance of the wall before rejection of an element was determined. During construction, the geospatial data management system permitted timely access to data and the ability to monitor construction progress near real time and facilitated verification that project specifications were met. Figure 12 shows some compiled data for evaluation of a pile that deviated out of alignment at the bottom and did not meet specification.

Post-construction, the system aided in verifying that the project design intent was met. In the long term, the system will be maintained as a living document of the project with a historical accounting of the original project construction and dam modifications for future use by both design engineers and operations personnel.

Due to the successful implementation of the data management solution at Wolf Creek Dam, implementation of data management systems began to be included in subsequent projects. Because of the timing of the Center Hill contract, not all of the lessons learned from the Wolf Creek data management system could be incorporated into the contract specifications. Fortuitously, the data management subcontractor was the same on both Wolf Creek and Center Hill; thus, the contractor’s proposal included a system very much like that at Wolf Creek. Ultimately, the system was designed by the data management subcontractor with input from the USACE, and lessons learned from Wolf Creek were implemented at Center Hill even though some of those lessons were not reflected in the specifications. These data management innovations allowed the USACE to effectively communicate results quickly both at the site level and up the vertical chain. The Wolf Creek Dam project had a major technical meeting to assess the quality of the barrier wall 6 days after the installation

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Concrete Quantities

Encasement wall Barrier wall Miscellaneous Job total

Volume of Concrete Placed (CY)

Number of Batches of Concrete

Number of Concrete Cylinders Cast

Number of Slump Tests Performed

Number of Temperature Readings Taken

43,422 30,099 560 74,081

4,412 2,172 29 6,613

1,711 604 46 2,361

4,412 2,171 29 6,612

4,412 2,172 29 6,613

Figure 11. Data table (top) and plot (bottom) of concrete data from Center Hill Dam.

of the last barrier wall pile, an efficiency that would have been impossible with earlier methods. The successful implementation of the geospatial data management solution on both projects led to the development of a USACE guide specification that was then implemented partially at Pine Creek Dam and was included in both the Bolivar and the East Branch Dam cutoff wall projects as a separate data management specification section.

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Lessons learned from Center Hill and Wolf Creek have also been incorporated into other areas of USACE work, as it was the basis for the SIMDAMS (Site Information Modeling and Data Management Solutions) program, which provides data management and GIS models of dam projects to support risk assessments. Both the Center Hill and the Wolf Creek models were extensively used in the post-construction risk assessments and in the production of the completion

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Figure 12. Use of GIS and data management system to evaluate a pile that did not meet minimum overlap specification of 2.0 ft.

reports and have been implemented on the Web inside the USACE network.

CONCLUSIONS Communication of lessons learned and implementation of improvements on future products provide for an environment of enhancement over time. The evolution of barrier wall contract specifications has provided for more effective contracting mechanisms and the ability to keep the door open to innovation while still meeting the project needs. Methodologies for verification of barrier wall final geometry and material properties have been modified over time as a result of proven technologies and gained confidence in the available tools. Most recently, by including data management solutions as part of the contract requirements, the USACE is able to have fully documented products for each project that aid in quality assurance and verification of design that is unprecedented in historical USACE barrier wall projects. The data management system provides for both near-real-time verification of the project as construction progresses and long-term effective documentation for future design and operations use. As barrier wall projects continue on into the

future, the contract methodologies and design specifications will likely continue to evolve with increasing shared bodies of knowledge and new available technological advances.

REFERENCES CAN/BNQ, 2008, Soils—Determination of the coefficient of permeability by the Lefranc method. CAN/BNQ Standard 2501-135. https://www.bnq.qc.ca/en/standardization/civilengineering-and-urban-infrastructure/soils/soilsdetermination-of-the-coefficient-of-permeability-by-thelefranc-method.html FAR SUBPART 27.404-1. Chapter 1, Title 48 Code of Federal Regulations. https://www.acquisition.gov/far/current/html/ Subpart%2027_4.html HVORSLEV, M., 1951, Time Lag and Soil Permeability in GroundWater Observations: Bulletin No. 36, Waterways Experiment Station, USACE, Vicksburg, Mississippi, 50 pp. 3.

KELLBERG, J. and SIMMONS, M., 1977, Geology of the Cumberland River Basin and the Wolf Creek Damsite, Kentucky: Bulletin of the Association of Engineering Geologists, Vol. 14, No. 4, pp.

245–269.

RICE, J. and DUNCAN, J., 2009, Findings of case histories on the long term performance of seepage barriers in dams: Journal of Geotechncial and Geoenvironmental Engineering, Vol. 136, No. 1, pp. 2–15.

SCHAEFFER, J., 2011, Mississinewa Lake dam internal erosion case history: RMC PowerPoint presentation.

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STARE, D.; DREESE, T.; AND BRUCE, D., 2013, Contemporary drilling and grouting methods. In XXXX (Editors), Specialty Construction Techniques for Dam and Levee Remediation: Boca Raton, FL, pp. 15–106.

USACE, 1974, Wolf Creek Dam request for technical proposals for the construction of concrete diaphragm wall. USACENashville District Library, Nashville, TN. USACE, 1988, Wolf Creek diaphragm wall final completion reports, phase I and II, USACE-Nashville District Library, Nashville, TN, December 1988.

USACE, 2001, Dam foundation remediation contract specifications, Mississinewa Lake, Indiana, Contract No. DACW27-01-R0003, USACE-Louisville District, Louisville, KY, February 23, 2001.

USACE, 2014a, Big Sandy Creek of Tuscarawas River, Bolivar Dam seepage barrier as-awarded certified final, Contract No. W91237-14-C-0003, USACE-Huntington District, Huntington, WV, February 2014.

USACE, 2014b, East Branch Lake, East Branch Clarion River Dam safety modifications/barrier wall as-awarded certified fi-

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nal technical specifications, Vol. 1, Contract No. W911WN-14C-0002, USACE-Pittsburgh District, Pittsburgh, PA, August 2014.

USACE, 2015, Center Hill Dam, supplemental major rehabilitation report. USACE-Nashville District Library, Nashville, TN. USACE, 2017a, Geotechnical and concrete materials completion report for the Wolf Creek. USACE-Nashville District Library, Nashville, TN. USACE, 2017b, Geotechnical and concrete materials completion report for the installation of barrier walls and grout curtains, Wolf Creek Dam. USACE-Nashville District Library, Nashville, TN. USACE, 2017c, Geotechnical and concrete materials completion report for installation of grout curtain. USACE-Nashville District Library, Nashville, TN. USACE, 2017d, Geotechnical and concrete materials completion report for installation of barrier wall, Center Hill Dam (draft).

ZOCCOLA, M.; HASKINS, T.; AND JACKSON, D., 2009, Seepage, piping and remediation in a Karst foundation at Wolf Creek Dam: In XXXX (Editors), 29th Annual USSD Conference: April 20-24, 2009, Nashville, TN, pp.

1464–1474. https://www.ussdams.org/wp-content/uploads/ 2016/05/USSDProceedings2006.pdf

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Failure of the Alexander Dam Embankment and Reconstruction Using Drainage Mitigation on Kauai, Hawaii, 1930–1932 KERRY D. CATO∗ California State University, San Bernardino, CA 92407

J. DAVID ROGERS Missouri University of Science and Technology, Rolla, MO 65409

Key Terms: Dams, Construction, Landslides, Materials of Construction, Hydraulic Fill, Dam Failure ABSTRACT Alexander Dam is a hydraulic fill earth dam and the second-highest embankment dam in Hawaii, having been built in 1929–1932 on the south side of the Hawaiian island of Kauai to provide irrigation for McBryde Sugar Company Ltd. It was constructed across Wahiawa Stream mauka (Hawaiian for “stream that comes from the mountains,” literally “toward the mountains”), upstream of Kalaheo, to store 800 million gallons (5 million m3 ) of water to irrigate sugarcane fields. The embankment dam was intended to have a maximum height of 125 ft (38 m), a crest length of 620 ft (189 m), and a maximum base thickness of 640 ft (195 m). The total design volume was 580,000 yd3 (443,120 m3 ) and consisted of hydraulic fill sluiced to the dam site and supporting shell material. On March 23, 1930, a 60-ft(18.3-m) wide section of the core pool suddenly dropped ∼30 ft (9.1 m) and moved downstream, rapidly draining the core pool and enlarging the mass. The embankment was at a height of 95 ft (29 m) and 78 percent complete when the failure occurred. The failure occurred so quickly that it killed six workers and injured two others on the downstream face. The volume of slide debris was ∼275,000 yd3 (210,100 m3 ). Thirty feet (9.1 m) of the embankment’s core stood near vertical after the failure, leading engineers to believe that the materials making up the downstream shell had consolidated sufficiently to inhibit internal drainage. The embankment was rebuilt by emplacing a 40-ft- (12.2-m) high rock buttress across the downstream toe, widening the downstream shell, and installing tile drains to facilitate internal drainage. The retrofitted structure was completed in December 1932 and remains in service some 85 years later.

∗ Corresponding

author email: kerry.cato@csusb.edu.

INTRODUCTION Among the many important lessons to be learned from the Alexander Dam failure are the necessity of effective internal drainage in the sloping shells of a hydraulic fill embankments during construction and the unique properties of residual soils developed in volcanic rocks in a tropical climate. By current design standards, hydraulic fill dams are considered inherently unstable in terms of seismic shaking because of their low relative density (Seed, 1979). What was not widely appreciated in 1930 was just how unstable and susceptible to rapid, catastrophic failures hydraulic fill embankments could be. In the case of Alexander Dam, the failure occurred so rapidly that six workers standing on the embankment were swept away in the failure and buried in the debris. While this failure and loss of life were tragic, it should be noted that reservoir water was not released during this event; no villages existed downstream of the dam along Wahiawa Stream, only individual houses and sugarcane work yards. The catastrophic failure provided one of the earliest case histories demonstrating how internal drainage of the supporting shells is a critical aspect of the performance of hydraulic fill embankments (Gilboy, 1934).

The use of rolled fill embankments became more common after the construction of the Bouquet Canyon embankments near Los Angeles in 1933–1934, which introduced mechanical compaction using the Proctor test (Proctor, 1933). The last hydraulic fill embankment of any significance was the Fort Peck Dam constructed by the Army Corps of Engineers, which experienced a massive liquefaction failure (9.94 million yd3 , or 7.6 million m3 ) of the dam’s right upstream shell in September 1938 (Middlebrooks, 1942). In the post-war years, rolled fill embankments became more economical with the proliferation of better earthmoving equipment and the realization that mechanical compaction at 85 percent saturation was less expensive and more conservative than hydraulic fill methods (Casagrande, 1950). Nevertheless, hydraulic fills are still widely

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employed for tailing dams in the mining and power industries.

The failure of Alexander Dam demonstrated that during construction, particular attention should be given to the hydraulic conductivity of the supporting shells, which are subject to increases in effective normal stress as the embankment rises. The fill materials used in the supporting shell contained too many fines to maintain effective subdrainage, and this defect led to the failure during construction. The embankment was repaired by employing an innovative system of subdrainage that, even without filters, allowed successful completion of the dam.

A few years later, Casagrande (1934) suggested that the construction of hydraulic fill dams should “be limited to such cases where an abundant supply of coarsegrained soils is available for the construction of the outer sections, or whether a satisfactory hydraulic-fill dam can also be built with very limited quantities of coarse grained materials.”

From the standpoint of materials, this case points out that deeply weathered volcanic materials impart unique characteristics that generally inhibit drainage.

The fill was sluiced from deeply weathered residuum that contained fine-grained materials that inhibited subdrainage and allowed entrained pore water pressure to rise as filling progressed. As the embankment rose in elevation, the material volume of each new lift diminished substantially. As less and less fill was required for each lift, the embankment height rose much more quickly. This rapid heightening of the embankment would have increased pore water pressure because of increased pressure head. The increased head would serve to reduce the effective stress within the embankment and thus reduce its shear resistance.

The success of the mitigation measures employed on Alexander Dam in 1931–1932 was subsequently modeled by Professor Glennon Gilboy at the Massachusetts Institute of Technology (MIT) (Gilboy, 1934). Gilboy’s work was appreciated by his mentor Karl Terzaghi, who subsequently advanced his theory of effective stress in soil mechanics (Terzaghi, 1943; Goodman, 1999).

SETTING AND BACKGROUND Alexander Dam is located on Kauai, the most western and northern of the principal Hawaiian Islands (Figure 1). The dam is situated on the southwestern side of the island, about 2 mi (3 km) from Kalaheo, which has a population of 3,900. Today, Kauai is mainly rural with the economy based on tourism, retirees, coffee plantations, and wineries. The largest town on the island is Kapa’a, with a population 9,500.

The dam site is located about 9.5 mi (15 km) from the town of Waimea. Kauai is the oldest inhabited of 90

the Hawaiian Islands, having been settled by Polynesians around 400–500 CE (Kauai Historical Society, http://www.kauaihistoricalsociety.org).

Background In 1928, construction began on Alexander Dam for the McBryde Sugar Company Ltd. McBryde began planting sugarcane in 1899 and was acquired by Alexander & Baldwin Ltd in 1909. By 1920, McBryde Sugar Company employed approximately 1,500 workers, and the company facilities included a store and hospital (University of Hawaii, 2017). The McBryde Sugar Company continued operating until 1987 when another subsidiary company of Alexander and Baldwin, the Kauai Coffee Company, took over operations, including operation of Alexander Dam (Cultural Surveys Hawaii, 2005). When Alexander Dam was built, only three sugar companies were operating plantations on Kauai (see Figure 2A), down from nine commercial sugarcane plantations on Kauai between 1835 and 1907. Production of sugarcane increased significantly in 1875 when the Treaty of Reciprocity between the Republic of Hawaii and the United States allowed sugar imports with low tariffs (Figure 2B).

The Hawaiian Islands are subject to northeasterly trade winds that dump large volumes of precipitation on the northeast-facing slopes of the islands. The leeward side of Kauai faces south and west, and those slopes and valleys are typically warmer, drier, and sunnier than the slopes and valleys on the opposite side of the island. Sugarcane requires continuous moisture for 6 to 7 months each year. While Kauai receives the most rainfall of any of the eight Hawaiian Islands, over 300 in. (762 cm) per year at the highest elevations (Kauasian Institute, 2011), the McBryde Plantation was situated on the leeward side, which receives only between 40 and 80 in. (100–200 cm) per year (Figure 3A and B).

Typical rainfall also varies seasonally from 4 to 6 in.

(10–15 cm) per month during the winter to only 1 to 2 in. (2.5 to 5 cm) per month during the summer. As a consequence, crops planted on the leeward side need additional moisture, which can be supplied only by irrigation. By 1920, sugarcane plantations on Kauai were diverting more than 800 million gallons (3 million m3 ) of surface water per day while pumping an additional 400 million gallons (1.5 million m3 ) per day out of the ground (Wilcox, 1997). Alexander Dam was constructed to store 750 million gallons (2.8 million m3 ) of runoff that could be utilized to augment the irrigation of nearby sugarcane fields.

Geologic Setting Kauai lies on the Pacific Plate, which is moving about 6 mm/yr toward the northwest over a pair of hot spots

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Figure 1. Location map of Alexander Dam on the island of Kauai, Hawaii (modified from Google Earth Pro, 2017, and USGS, 2017a).

in the earth’s mantle. The hot spots currently lie beneath the big island of Hawaii. As a result of the northwestern movement of the Pacific Plate relative to the hot spots, the volcanic rocks making up the Hawaiian Is-

lands decrease in age to the southeast. For example, the volcanics on the island of Hawaii range from 430,000 ybp (years before present, or 430 ka) at the western side to present-day age at Kilauea Volcano on the eastern

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Figure 2. (A) Sugar plantations that operated on Kauai, Hawaii. McBryde Sugar Company Ltd began in 1899 and operated until 1959. In 1930, when Alexander Dam was built, three sugar companies were still operating (modified from Wilcox, 1997). (B) 1875 Treaty of Reciprocity allowed low-tax/tariff trade with the United States and triggered an era of growth in sugar production in Hawaii that continued until around 1965 (University of Hawaii, 2001).

side. These active volcanics are located about 320 mi (515 km) southeast of the Alexander Dam site. The areal extent of the Hawaiian Islands diminishes with time due to the relentless surf erosion. Kauai is the fourth largest of the main Hawaiian Islands and mea92

sures approximately 25 mi (40 km) north to south and 30 mi (48 km) east to west.

Figure 4A and B presents a physiographic map of Kauai (A, top) and a geologic map of the site area (B, bottom). The arrow indicates the location of

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Figure 3. (A) Kauai precipitation isohyets shown in inches per year. The McBryde Sugar Plantation (near the black circle showing location of Alexander Dam) receives annual rainfall ranging from 60–80 in. (153–203 cm) per year (Kauaian Institute, 2011). (B) Monthly average rainfall in Kauai is much greater in the October–March winter months, when it ranges from 4 to 6 in. (10–15 cm) per month, than in the April–September summer months, when rainfall averages only 1 to 2 in. (2.5 to 5 cm) per month (modified from Wilcox, 1997).

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Figure 4. (A) Physiographic map of Kauai. Arrow indicates reservoir location (modified from Wilcox, 1997). (B) Geologic map of site area (USGS, 2007). Arrow indicates dam location.

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Figure 5. Typical physiographic features of Hawaiian slopes, which include dikes, buried stream channels, lava tubes, and buried residual soils (modified from Deere and Patton, 1971).

Alexander Dam just east of Hanapepe Canyon. The age of the volcanism on Kauai is early Pliocene age (5 million ybp, or Ma).

Hawaiian volcanics are typically comprised of interbedded lava flows, volcanic ash, and eroded surfaces and gravel-filled stream channels filled with younger flows (because the stream channels existed as topographic lows at time of volcanism) (Macdonald et al., 1960). Figure 5 shows a typical slope in Hawaii, which can be underlain by aa breccia flows, vesicular basalts in pahoehoe flows, ash, buried stream channels, buried residual soils, and younger cross-cutting dikes.

Buried soils are very common on Kauai and develop during periods of dormancy or after deposition and re-working of ashfall deposits. Incised channels are often filled with alluvium, which become quasi– “underground rivers.” In 1927, McBryde Sugar was excavating water supply tunnels under the Hanapepe River when they intercepted an underground channel that yielded between 2 million and 32 million gallons (7,500–120,000 m3 ) per day for many years thereafter (Wilcox, 1997).

Most of the slopes on Kauai are structurally influenced by the gentle dips of the lava flows, which are

usually oriented subparallel to the slopes. The aa flows generally consist of highly permeable breccia or clinkers resembling scoria, with intermediate layers of dense aphanitic basalt. The breccia beds and seams can easily transmit water but not always for long distances, depending on the presence of other discontinuities, such as dikes (Figure 5). The pahoehoe flows were more fluid and vary in thickness from a few centimeters to as much as a few meters. The exposed surface of a pahoehoe flow typically forms a vesicular crust that is 1 to 5 cm thick.

Rainfall tends to percolate the intensely fractured mass with seeming ease, but the horizontal permeability is generally far greater. The groundwater table usually lies at some depth beneath the exposed slopes, as shown in Figure 5. Sudden shifts in groundwater levels can be caused by aphanitic dikes with baked zones on either side comprised of colloidal materials of low hydraulic conductivity (Figure 5).

Alexander Dam was constructed across Wahiawa Stream mauka, upstream of Kalaheo, which drains to the Pacific Ocean (USGS, 2017b). This stream trends south-southwest, subparallel to the larger Hanapepe Canyon, the adjacent watershed to the west (Figure 6A). Wahiawa Stream developed on an old volcanic

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Figure 6. (A) 1913 drainage map of East Kauai, prepared by the USGS (modified from Wilcox, 1997). Arrow indicates reservoir location.

(B) Aerial photograph shows the Alexander Dam embankment and emergency spillway of the rebuilt dam (modified from Google Earth Pro, 2017).

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shield surface and, having a smaller watershed, has not incised itself as deeply as Hanapepe Canyon. Maximum local relief at the dam site is on the order of 125 ft (38 m), which controlled the maximum height of the dam.

SITE CONDITIONS AND DAM DESIGN Site Conditions Only limited details about the original dam are known because as-built plans and drawings of the original embankment and foundation are scarce. Reports of the failure include published and unpublished news stories and reports. Additional details about the original embankment were collected during the failure investigation and planning for the remedial work. The dam’s name came from Alexander & Baldwin Ltd of San Francisco and Honolulu, who controlled McBryde’s stock and served as the company’s agents.

The dam and reservoir are shown in a 2004 aerial photograph presented in Figure 6B. The reservoir is low, as about 400 ft (122 m) of the dam’s total 640-ft (195-m) upstream-to-downstream embankment width are exposed in the photo. The spillway is located on the right abutment and is controlled by an ogee weir that drains to a narrow concrete channel. The outlet works consist of two 8 × 8-ft (2.4 × 2.4-m) excavated rock tunnels that are not visible in the photo. The embankment dates from the repairs made in 1932 and has successfully operated for 85 years.

Technical Concerns of Hydraulic Fill Dams Most embankment dams constructed prior to the mid-1930s were of the hydraulic fill, sluiced fill, or “puddled fill” type, which employed hydraulic nozzles to sluice fill materials to fashion a crude sort of zoned fill embankment. This was accomplished by constructing rockfill containment dikes at the upstream and downstream toes of the proposed embankment, from which fill material would be sluiced by nozzles into the depression formed between the two dikes. The fill material could be conveyed to the dikes by wagons, conveyor belts, or side-dumping railcars or through hydraulic flumes. The fill materials usually came from excavations (borrow pits) located on either abutment, usually within a short distance from the embankment and ideally at sufficient elevation to permit sluicing of the material onto the embankment. In some cases, the valley fill would be excavated with dredges and then pumped upward onto the containment dikes and thence into the embankment; however, this method was more expensive because of the additional pumping costs.

The goal of the sluicing was to hydraulically sort coarse-grained, free-draining materials (which tend to drop out first), called the “beach deposits,” from the finer-grained silts and clay, which tended to remain in suspension within the turbid “core pool” that formed in the lowest area, between the containment dikes (Figure 7A). The dam’s core was intended to be of low permeability to preclude seepage from the reservoir.

Core material of the lowest permeability was preferred in order to reduce hydraulic uplift and thereby increase overall stability (Saville, 1908). The supporting shells, or “beaches” as they were then described (Figure 7B), were situated on either side of the fine-grained (siltclay) core. The term “beach” was intended to describe free-draining materials, such as sand and gravel, which allow drainage and retain shear strength even when submerged.

Some of the most common problems with hydraulic fills included the following: 1. When construction began, borrow pits usually concentrated on weathered materials exposed in the upper portions of either abutment so that finergrained source materials were initially being placed.

As borrow pits proceeded deeper and deeper into the abutments and yielded fewer fine-grained materials, coarser-grained source materials were being placed.

2. As the source materials became increasingly coarse, fewer fine-grained materials were available to construct the dam’s low-permeability core.

3. If the source materials became increasingly fine grained, then the shells of the dam would not drain easily and would possess lower shear strength and bearing capacity.

4. Rapid filling of the embankment could trigger localized bearing failures and slides, such as those recorded during construction of Gatun Dam in 1908–1912, Belle Fourche Dam in 1909, Lafayette Dam in 1928, Saluda Dam in 1930, and Alexander Dam in 1930.

5. Rapid placement of fill could trigger static liquefaction failures of the upstream shells (Calaveras Dam in 1918) or liquefaction of the foundation materials (Fort Peck Dam in 1938).

6. Seismic shaking could trigger liquefaction of cohesionless fill materials or of uncemented sands and gravels in the dam’s foundation, as occurred at Sheffield Dam in 1925, Chatsworth Dam in 1930, Lower San Fernando Dam in 1971, and Garvey Reservoir in 1985–1987.

7. “Topping off” with rolled or tamped fills was necessary to complete the trapezoidal section at the crest of hydraulic fill dams, resulting in composite sections of varying permeability and relative density.

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Figure 7. Common designs, techniques, and assumptions used in hydraulic fill embankments. (A) Typical hydraulic fill placement technique showing how sluiced material is separated and sent to both the upstream and the downstream areas. From there, sediment is discharged from opposing levees onto “beaches” to form the shoulders (i.e., shells). The fine-grained materials slowly settle out in a central pool to form the core pool. The clay core of the embankment serves as the key element of any earthen dam (Plummer and Dore, 1940). (B) As material is discharged away from the upstream and downstream levee, the coarse-grained materials settle first with the fine-grained materials pooling and settling out in the core area. The assumption is that coarse, well-drained sand and gravel will form the shells (Morgan, 1951).

Original Design of the Dam The first dam constructed for sugarcane irrigation in Hawaii was on the Waialua Sugar Plantation on Oahu in 1904–1906 (Wegmann, 1922). It was an earth and 98

rockfill embankment 98 ft (30 m) high with a base width of 580 ft (177 m). Puddled clay was sluiced against the upstream side of the rockfill placed against the core wall. The crest length was 460 ft (140 m), and the embankment volume was 141,000 yd3 (107,724 m3 ).

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Figure 8. Embankment dimensions and zonation for the original design; cutoff wall detail is shown at bottom (modified from ENR, 1930a).

When the time came to consider what type of dam to build along the Wahiawa Stream on Kauai in 1928, five reasons were cited for employing a hydraulic fill embankment:

• At that time (1928), rolled-earthfill embankments had a poor performance record in Hawaii due to “formation of water channels through compacted fill” (Stearns, 1964).

• Laboratory tests indicated an in situ void ratio of 0.6 to 0.68 for roller compacted soils, providing very low permeability as compared to other earthfill dams (Cox, 1938).

• It was costly to transport construction equipment to this remote site with steep terrain without any established road network, and there were also the daily maintenance costs and supplies, such as fuel.

• There was a belief at the time that with rolled embankments, failure could occur anytime during the dam’s operating life but that the failure of hydraulic fill embankments tended to occur only during construction.

• The valley bottom would not support the weight of a mass concrete structure.

A cross section of the original embankment is shown in Figure 8. The upstream slope was constructed at

a 3H:1V (horizontal to vertical) slope (18.1◦ ), and the downstream face was sloped at 2H:1V (26.6◦ ).

The maximum embankment height was about 125 ft (38 m), with a maximum base width of 640 ft (195 m). The total embankment volume was to be 580,000 yd3 (430,120 m3 ), and the completed reservoir was intended to impound 750 million gallons (2,300 ac-ft, or 2,837,000 m3 ). While the reservoir storage seems small in regard to the embankment volume, it was pointed out that this site was better than most on the islands, where 1,000 yd3 of earthfill is usually required to provide a million gallons (3.1 ac-ft, or 3,785 m3 ) of storage (Engineering News Record [ENR], 1930b).

A small, 50-ft- (15.2-m) high, hydraulic fill “pioneer dam” was built at the upstream toe of the main embankment to serve as a cofferdam (Figure 8). This structure was to be incorporated into the main embankment. The purpose of the pioneer dam was to impound a sufficient reservoir to supply water for the hydraulic fill operations for the main embankment and to generate 1,500 kW of hydroelectric power to run pumps that lifted the lake water up to the borrow areas on either side of the stream valley.

To avoid seepage that had occurred in a “porous stratum” beneath the pioneer dam, a cutoff trench was ex-

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cavated beneath the main embankment through some “porous layers” and founded on “hard” lava rock. A concrete cutoff wall was placed in the center of this trapezoidal-shaped trench and filled to the full depth of the trench. On either side of the concrete wall, a triangular area was backfilled with “alkali-treated earth” (Figure 8). About 40–50 pounds (18–22.7 kg) of sodium carbonate (Na2 CO3 ) per ton (908 kg) of soil was added as chemical treatment of the backfill to reduce its permeability. This was the first example of chemical stabilization of earth materials to be used in a dam (Mitchell, 1993). It was designed by Dr. Francis E. Hance, who was an agricultural scientist working for the Hawaiian Sugar Planters Association Experiment Station (Hance, 1929; Cox, 1932a).

The upstream cofferdam (pioneer dam) was constructed between August 1928 and January 1929. It was allowed to drain for many months before the main embankment rose above its height. As a consequence, it did not fail when the downstream shell failed catastrophically in March 1930, likely because it did not suffer from excess pore water pressures even though submerged. The main embankment was constructed at a much faster rate. Its construction began in May 1929, and within 10 months, it had reached a height of 95 ft (29 m) before it failed on March 23, 1930.

Fill material was transported to the main embankment by a 2 × 2-ft (0.6 × 0.6-m) sluice on a gradient of about 3 12 percent. Water for the sluicing operation was supplied by two 1,500-gpm (5,700-L/min)capacity pumps that operated under 300 ft (91.4 m) of head. At the dam, the sluice split into two parallel branches to supply material to both the upstream and the downstream beaches. Figure 7A and B presents schematic views of the hydraulic fill technique. At Alexander Dam, the saprolite materials used to fill the sloping “beaches” were disaggregated and contained far too many fines to allow effective drainage. Fortyfive percent of the fill mixture was finer than 0.01 mm (fine silt and clay). In his reports, Cox (1932a, 1932b) referred to these materials as laterites, but we are referring to the weathered bedrock material, not simply the residual soil horizons; thus, the term “saprolite” is used here. Most of the materials used to construct hydraulic fill embankments of that era fell into a range of values presented graphically in Figure 9, which shows the sieve analyses for all American-built hydraulic fill dams up through 1940 (Plummer and Dore, 1940). Note how the sieve analysis for Alexander Dam is an outlier, far finer than any other dam shown.

Material for the embankment was obtained from residual soil deposits above the dam site that formed from weathering of the volcanic material to depths of approximately 100 ft (30.5 m). (Note: At many sites, embankment material can be transported from dis100

tant borrow sources [Figure 10], but because of the remoteness and absence of vehicular site access at the time, all borrow materials for Alexander Dam were, of necessity, obtained locally using hydraulic excavation and conveyance techniques [Figure 11].). Vesicular lavas formed the dam’s right (north) abutment. Ash and coarse pyroclastic materials from a nearby cone, with some evidence of fumarole activity, formed the left (south) abutment. Both of these materials were deeply weathered. The materials consisted of “with the exception of residual boulders on the north ridge... heavy earth, claylike in character and variable in color, with red predominating” (ENR, 1930b). The red color is indicative of oxidation.

Blasting with black powder was used to dilate the in situ materials so that they could be excavated and sluiced to the dam site in order to minimize the finegrained percentage being deposited on the embankment (as opposed to hydraulic excavation). At the embankment, the sluiced material was observed to arrive in “12 inch (30 cm) lumps down to colloidal mud” (ENR, 1930a) The coarser fraction settled out quickly, forming the upstream and downstream beaches, with the finer-grained and colloidal material forming the dam’s core.

Joel B. Cox was the chief engineer for the project.

He was a 1915 graduate of Stanford University in civil engineering and worked in Hawaii his entire career, except for 10 months during World War I (1917–1919).

EMBANKMENT FAILURE EVENT The failure of the Alexander Dam embankment occurred at 3:45 p.m. on March 23, 1930 (ENR, 1930a).

At the time of failure, the embankment crest was 95 ft (29 m) above the streambed, or an elevation of 1,575 ft (480 m). The embankment was 78 percent complete with 453,000 yd3 (346,000 m3 ) of hydraulic fill placed.

The reservoir pool was 40 ft (12.2 m) below the top of the embankment at elevation 1,535 ft (468 m) and below the crest of the pioneer dam (see Figure 12).

The failure commenced without warning. According to interviews with workers after the failure, plastic deformation and the flow of mud (dilation) began some time before the mass movement was noticed (Figure 12). Shortly thereafter, a 60-ft- (18.3-m) wide section (graben) of the core pool suddenly dropped 30 ft (9.1 m) vertically and translated downstream. This movement rapidly drained the core pool and enlarged the slide mass (Figure 13). Approximately 257,000 yd3 (196,350 m3 ), or about 57 percent of the embankment, was involved in the slide. Water stored in the temporary reservoir was unaffected by the failure, remaining at elevation 1535 ft (468 m).

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Figure 9. Sieve analyses for all hydraulic fill dams built in the United States through 1940 (modified from Plummer and Dore, 1940). Most of the materials used to construct hydraulic fill embankments fell into a range shown on this grain size analysis; however, Alexander Dam’s sieve analyses (#12) are the outliers, far finer than any other dams shown.

Figure 10. This is the hydraulic fill Haiwee Dam in California, under construction. Here, beach fill was brought in on rail, which means that fill can be brought in from entirely different borrow sources, unlike Alexander Dam, where borrow came from on-site sources.

At Alexander Dam, when the saprolitic material arrived at the dam, it ranged from 12-in. (30-cm) lumps to fine colloidal mud (from Los Angeles Department of Water and Power).

Figure 11. View looking upstream during the early stage of construction of the dam in 1930. Fill was hydraulically sluiced along the inclined wooden flumes. Note loose-dumped containment dike in foreground (Wilcox, 1997).

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Figure 12. This is the embankment just before (A) and then after (B) the failure (modified from ENR, 1930b). (A) This was the embankment at the time of failure on March 23, 1930, at 3:45 p.m. At that point, the embankment was 95 ft (29 m) high and at elevation 1,575 ft (480 m).

The embankment was 78 percent complete in terms of the 453,000 yd3 (346,000 m3 ) of fill volume that had been placed. The reservoir level was at elevation 1,535 ft (468 m), or 40 ft (12.2 m) below the constructed crest, and still not over the top of the pioneer dam. The bold black line shows the landslide failure plane. (B) Failure occurred so quickly that it killed six workers and injured two on the downstream face. Volume of slide debris was approximately 275,000 yd3 (210,000 m3 ), or 47 percent of the volume of the placed embankment. The 30 vertical feet (9.1 m) of the embankment’s clay core that stood near vertical led investigating engineers to believe that the hydraulic fill materials deposited in the downstream shell had consolidated and thereby failed to allow internal drainage. Reservoir level at elevation 1,535 ft (468 m) was not affected by the failure.

Investigation of the Failure At the time, ideas about the cause of the embankment failure centered on two main issues. The first issue was the depth of weathering the volcanic source rocks on Kauai. Kauai is the oldest of the surviving Hawaiian Islands and receives the highest rainfall (Cox, 1924).

The deeply weathered nature of the saprolites was appreciated by the designers, and this information was allegedly considered in developing suitable borrow areas as well as in the design assumptions for the embankment. The dam was one of the first in the United States or its territories where consolidation tests were performed during the design stage to estimate the expected settlement of the embankment (Cox, 1936). Unfortunately, the saprolite materials with their preponderance of fine-grained materials restricted the amount of freedraining materials to a dangerously low percentage, with only 12 percent coarser than 0.075 mm (fine sand size), as shown in Figure 9. This led to serious problems with drainage of the beaches or supporting shells of the embankment. Even with the use of blasting and limit102

ing the hydraulic mining in a futile attempt to reduce the percentage of fines, there was just too much finegrained material in the embankment shell to allow sufficient drainage for consolidation. The saprolitic soils were disaggregated during excavation, hydraulic transport, and sluicing, with an abnormally high percentage of colloidal particles (Figure 14). This led to a much lower hydraulic conductivity in the shells than the designer (Joel B. Cox) assumed beforehand.

The lower hydraulic conductivity prevented the downstream shell from draining internally. At the time of the failure, it was speculated that overburden surcharging “promoted crushing, which produced smaller particles” (ENR, 1930b), but this was never verified.

The sieve analyses basically tell the story (Figure 9).

Without easy drainage, it is very likely that the internal pore water pressures became elevated, lowering the effective stress developed in the downstream shell, which could not support the much softer and saturated (lower-density) core. This was an inherent problem of hydraulic fill dams identified by Gilboy (1934), whose model studies showed that construction details and

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Figure 13. (A) Photograph of the embankment just after the failure; vie is toward the south and at the left abutment; upstream is on the left side of the photo, where there is a near-vertical face of in-place fill. On the left abutment, the core trench is visible above the level of where fill had been placed to a height of 95 ft (29 m). The landslide movement direction was from left to right; 53 percent of the embankment was involved in the landslide (modified from ENR, 1930b). (B) The graph at the bottom shows settlement that was recorded on the downstream face of the placed embankment. Note that it shows settlement beginning on September 16, 1929, and continuing through March 11, 1930 (the last record). Failure occurred 15 days later on March 23, 1930 (modified from ENR, 1930b).

geometry of the shells were critical to short-term stability of such structures during construction.

The permeability of any compact soil mixture tends to be controlled by the relative percentage of finegrained material (silt and clay). Mixtures with greater than 3 percent fines passing the No. 200 sieve could be expected to retard drainage capability. Mixtures with greater than 15 percent fines were considered to be non-draining. The materials being sluiced into Alexander Dam averaged about 88 percent silt and clay, which was an outlier when compared to other hydraulic fill embankments of that era (Figure 9). Post-failure interviews with workers who were on the dam confirmed that they had observed a cessation of drainage through the downstream shell in the vicinity of the slide area about 10 days prior to the failure. The buildup of pore water pressure within the downstream shell would have

reduced the effective stress acting on the soil, reducing its shear strength markedly.

Interaction between Joel Cox, Glennon Gilboy, and Karl Terzaghi While it is easy to look back with today’s knowledge and judge what has occurred in the past, it is instructive to understand how engineers of that era analyzed the failure and prepared an appropriate scheme to mitigate the problem. Detailed descriptions and photos of the failure were published in ENR at the time, which had the largest circulation of any serial engineering publication in the USA. This media coverage led Karl Terzaghi and his first doctoral student, Glennon Gilboy, to take interest in the situation. Gilboy’s master’s thesis under Terzaghi at MIT had focused on analyzing short-term

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Figure 14. The embankment material presented two design problems. First, the saprolitic soils were disaggregated during excavation, hydraulic transport, and surcharge. This is shown by the lower curve (typical natural material) moving toward the fine-grained part of the graph on agitation (natural material after stirring). Thus, the material breaks down more easily than predicted, creating an overabundance of fine-grained material and a shortage of coarsegrained materials (modified from ENR, 1930b). Second, all of this fine-grained material in the embankment leads to a much lower hydraulic conductivity than the designers imagined possible beforehand. The ability for the shell to drain is significantly lowered. This, in turn, lowered the effective stress of the embankment materials.

stability of hydraulic fill embankments. When Terzaghi departed in 1930, he named Gilboy as his replacement on the MIT faculty.

Karl Terzaghi had published several articles in ENR in 1926–1927 on the new field of soil mechanics. Those articles prompted Cox to evaluate the change in void ratio with loading for the borrow material, which was without precedent in Hawaii at that time (1928). His crude tests allowed him to measure material consolidation under the proposed load of the embankment.

When compared with actual settlement data (Figure 15), the amount of consolidation turned out to be much greater than predicted by the oedometer tests.

Cox (1932a) was aware of drainage problems in the embankment prior to its failure: “The response to the cessation of sluicing operations was prompt in the upper level, slower lower down. The lag due to a height of 100 ft (30.5 m) was about 3 weeks. As the dam increased in height and [pore] pressures increased, the permeability of the material surrounding the drains in the downstream shell decreased, so that less water was to be handled, the pressure head required remained nearly constant, only a moderate decrease in [pressure] head was observable.” It should be noted that these observations are the outward manifestations of internal drainage of the embankment shell becoming compromised.

Partly because of this failure and his collaboration with Joel Cox, Gilboy (1934) penned a convincing article that showed why the upstream and downstream shells of hydraulic fills were particularly vulnerable to

Figure 15. Karl Terzaghi’s articles in ENR in 1926–1927 prompted an analysis of soil void ratio with time, as the material consolidated under the surcharge load of the embankment. The actual consolidation of the core material turned out to be considerably greater than predicted by the lab oedometer test data, shown here (Cox, 1936). It was reported on by the design engineer Joel B. Cox at the First International Conference on Soil Mechanics and Foundation Engineering at Harvard University in 1936 through the personal invitation of Karl Terzaghi.

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Figure 16. (a) Workers constructing a free-draining rockfill prism on a slope of 1.5H:1V (33.6◦ ). This stacked rockfill served as the downstream containment dike, where the dam was highest (Wilcox, 1997). (b) View looking upstream at the 40-ft (12.2-m) high rockfill toe dike during reconstruction of the dam in 1931 (Wilcox, 1997).

pore pressure–induced and geometry-induced slope instabilities during construction. This led to the gradual demise in the use of hydraulic fill embankments after 1935.

A few years later, Karl Terzaghi invited Joel Cox to present his work on the Alexander Dam failure and repair at the First International Conference on Soil Mechanics and Foundation Engineering at Harvard University in June 1936 (Cox, 1936). At that meeting, Cox stated that he was aware that the saprolitic soils would become disaggregated during excavation, transport, sluicing, and surcharge. However, he pointed out that the much lower hydraulic conductivity observed in the 10 days preceding the failure was “a change in material properties” that was unanticipated. What merited Cox’s invitation to the international conference was his installation of a coherent system of internal subdrainage while rebuilding the embankment, which was an elegant solution.

REMEDIATION At the time of the failure, Joel Cox recognized that the lack of drainage in the downstream shell was the trigger mechanism for the failure, but no one fully appreciated the concept of effective stress and the debilitating role of elevated pore pressures on shear strength.

Fortunately, the McBryde Sugar Company allowed Cox to effect the repairs instead of hiring someone else.

Cox began the remedial work by constructing a 40ft- (12.2-m) high rockfill dike inclined at 1.5:1 (33.6◦ ) at the downstream toe of the dam (Figure 16A and B).

Cox then focused his attention on a method to increase internal drainage of the supporting shell. His design was novel at the time and, in retrospect, the most appropriate remedial measure given the low permeability of the source material (Figures 17 and 18). The embedded drainage network consisted of a crisscrossing series

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Figure 17. This plan view of the downstream shell shows the drainage remedial measures that were placed in the downstream shell to increase effective stress (Cox, 1936). These consisted of a crisscrossing series of tile drains laid within rock-filled trenches beneath the downstream shell of the dam. The largest trenches were 36 × 48 in. (0.9 × 1.2 m) conveying 8 in.- (20.3-cm) diameter tile drains, while the smaller trenches were 24 × 24 in. (0.61 × 0.61 m) conveying 3 in.- (7.6-cm) diameter tile drains. These tile-formed drains were referred to as “pipes” on this graphic as well as Figure 18. Furthermore, the drains were installed at a relatively steep gradient so that if vertical settlement occurred, the drains would continue flowing.

of tile drains placed in rock-filled trenches within the downstream shell of the dam. The largest were 36 × 48in. (91 × 122 cm) trenches conveying 8-in.- (20.3-cm) diameter tile drains, while the smaller trenches were 24

in × 24 in (61 × 61 cm) with 3-in.- (7.6-cm) diameter tile drains.

The subdrains and collector pipes were laid out in a manner that would allow flow paths of less than 8 ft

Figure 18. Section of Alexander Dam showing core, beach, and toe sections; progress of hydraulic fill; locations of drains; and penetration depths of 8-in. (20.3-cm) balls. The ball test was used to estimate undrained and remolded shear strengths based on the amount of penetration of the sphere (Cox, 1932b). Cross-section view showing the subdrains that were installed in the rebuilt downstream shell at Alexander Dam. The system included 562 linear feet (171 m) of gravel-filled drainage trenches, 1,027 linear feet (313 m) of 8-in.- (20.3-cm) diameter drain tiles, and 4,331 linear feet 1,320 m) of 3-in.- (7.6-cm) diameter drain tile.

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Figure 19. View along axis of the dam as the new core pool of the rebuilt structure was topped off in November 1931. Downstream is to left, reservoir to the right. Note that the core pool becomes smaller with increasing height but also promotes increased hydraulic pressure head, commensurate with such height (courtesy of Richard Cox, PE [son of Joel Cox]).

(2.4 m) from any location in the downstream shell.

These drains were installed on a noticeable gradient so they would continue to flow as the embankment drained and settled. The system of subdrains included 562 linear feet (171 m) of gravel-filled drainage trenches, 1,027 linear feet (313 m) of 8-in.- (20.3-cm) diameter drain tiles, and 4,331 linear feet (1,320 m) of 3-in.- (7.6cm) diameter drain tiles. The drain tiles of that era were typically made of ceramic or terra-cotta and equipped with holes or slits to allow collection and conveyance of seepage without the hazard of corrosion. The natural soils at the dam site were recognized as being strongly acidic, typical of the tropics (Hance, 1929).

This integrated system of subdrainage has served as a surrogate for the more traditional system of subdrainage in zoned fill embankments, which employ sand and gravel filters along the upstream and downstream margins of the low-permeability clay core.

Alexander Dam has performed well for 85 years since being retrofitted with the subdrains (Figure 18).

The filling operations proceeded without any problems, as shown in Figures 19, 20, and 21. When the rebuilt embankment reached an elevation of 1,593 ft (486 m), 12 ft (3.7 m) below the design crest, Cox decided to delay further filling. Cox left a 12-ft- (3.7-m) high cofferdam at the upstream crest of the dam (Figure 21) while waiting an entire year for the internal stresses and pore water pressure to equilibrate and the embankment to settle before adding the embankment’s crown.

As the reservoir filled, Cox was pleased to learn that the effective storage (absent flood storage) was 800 million gallons (2455 ac-ft, or 3 million m3 ), about 14 percent greater than originally envisioned. When the project was completed in December 1932, the total cost was U.S.$2.2 million, which included the 1,100-kW Kalaheo hydroelectric power plant, which supplied power to the three major wells in the Hanapepe Valley. For comparison, using changes in the consumer price in-

Figure 20. View looking north at right abutment showing the central box flume and the sloping lateral flumes, which dumped sluiced material onto the “beaches” at the distal edges of the embankment, and flowing back toward the central core pool (Wilcox, 1997).

dex, $2.2 million in 1932 would be worth approximately $41.5 million in 2017 (U.S. Bureau of Labor Statistics, 2017).

CONCLUSIONS This case history serves to document some of the problems with deeply weathered saprolitic soils in tropical climes and the short-term instability of hydraulic fill dams during construction. Today, we are constantly reminded about the seismic instability of hydraulic fills because of their low relative density and susceptibility to liquefaction and lateral spreading. However, the real lesson of this study was that the source material was bereft of sufficient coarse-grained, cohesionless material, which was needed for drainage and short-term strength. The case history also demonstrates how pore pressures tend to elevate with the heightening of the embankment, which accelerates as the dam approaches its design crest elevation, because less and less fill is required for each lift due to the trapezoidal shape of the embankment.

Figure 21. Cross section of the rebuilt dam as it appeared in 1931– 1932. The embankment’s crown was constructed as a rolled fill in December 1932 (Cox, 1932a).

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This example shows the effects on an embankment with insufficient internal drainage. While materials such as glacial outwash deposits are particularly well suited to hydraulic fill structures, other materials with poorly graded soil mixtures seldom perform well.

Saprolitic or other clay-rich materials would also be suspect.

As the Alexander Dam embankment shows, fill using volcanic residuum is problematic because of disaggregation and breakdown of clods, creating a semiimpervious and low-strength fill. This example points out the need for thorough testing of materials, in both wet and dry states, to gauge how material properties change with increasing effective stress and percent saturation.

The mode of failure remains relevant for hydraulic fill embankments being constructed today. Hydraulic fills are still used for tailings dams across the United States and internationally. These embankments, like many levees, become “legacy structures” because their cross sections have been modified repeatedly, complicating their systems of internal drainage and consolidation. The predominance of low-strength, fine-grained, and low-permeability materials in tailings dams makes them particularly susceptible to pore pressure entrapment if the integrity of the interconnecting system of subdrainage ever breaks down. The success of using a drain system to increase permeabilities and the downstream shell was a major accomplishment that allowed the dam to be completed and function effectively for 85 years. This mitigation may have applications to other hydraulic structures, such as tailings ponds.

When the dam failed, Cox allowed many critical details to be published in ENR, which led to professional collaborations with learned people, such as Professor Glennon Gilboy at MIT, who had spent his professional career studying hydraulic fills. These collaborations led to Cox publishing numerous articles on the Alexander Dam studies, sharing his laboratory and field data (Cox, 1934). He received invitations to lecture on the dam at both MIT in November 1932 and the First International Conference on Soil Mechanics at Harvard University in June 1936. Cox remained with the McBryde Sugar Company and their parent company Alexander & Baldwin Ltd. Cox joined the engineering faculty at the University of Hawaii in Manoa in the fall of 1941, became engineering department head in mid-1944, and retired in 1956.

REFERENCES CASAGRANDE, A., 1934, Discussion of mechanics of hydraulic fill dams: Journal of the Boston Society of Civil Engineers, Vol. 21, No. 3, pp. 206–207.

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CASAGRANDE, A., 1950, Notes on the design of earth dams: Journal Boston Society Civil Engineers, Vol. 37, No. 4, pp. 405–429.

COX, J. B., 1924, Periodic fluctuations of rainfall in Hawaii: ASCE Transactions, Vol. 87, pp. 461–490.

COX, J. B., 1932a, The Alexander Dam in Hawaii: Technology Review, Vol. 35, No. 2, pp. 53–55.

COX, J. B., 1932b, Beach drainage safeguards, Alexander Dam: Engineering News Record, Vol. 109 (October 20), pp. 466–469.

COX, J. B., 1934, Discussion on hydraulic fill dams: ASCE Transactions, Vol. 99, pp. 263–277.

COX, J. B., 1936, The Alexander Dam, soil studies and settlement observations: In Proceedings, First International Conference Soil Mechanics and Foundation Engineering, Vol. II, Paper Z-12: Harvard University, Cambridge, MA, pp. 296–298.

COX, J. B., 1938, Discussion on rolled-fill earth dams: ASCE Transactions, Vol. 103, pp. 28–30.

CULTURAL SURVEYS HAWAII, 2005, An Archaeological Inventory Survey to Support the Restoration of Alexander Dam Irrigation Ditch Project, Wahiaw Ahupua’a, Lolao-Poipu District, Kaua’i: Cultural Surveys Hawaii, Inc., Kailua, HI. 58 p.

DEERE, D. U. AND F. D. PATTON, 1971, Slope stability in residual soils, In Proceedings 4th International Conference Soil Mechanics and Foundation Engineering, Vol. 1: Amer. Soc. of Civil Engineers, NY, NY. San Juan, Puerto Rico, pp. 87–170.

ENGINEERING NEWS RECORD, 1930a, Collapsed Alexander Dam a notable structure, cutoff trench sealed by chemically treated earth fill—Unusual design and construction methods: Engineering News Record, Vol. 104, No. 17, p. 703.

ENGINEERING NEWS RECORD, 1930b, Hydraulic-fill dam of fine volcanic ash fails disastrously: Engineering News Record, Vol. 104, No. 21, pp. 869–871.

GILBOY, G., 1934, Mechanics of hydraulic-fill dams: Journal of the Boston Society of Civil Engineers, Vol. 21, No. 3, pp. 185–205.

GOODMAN, R. E., 1999, Karl Terzaghi, The Engineer as Artist: ASCE Press. 352 pp.

GOOGLE EARTH PRO, 2017, 22.0479 ◦ N and −159.05354◦ E, December 2016 Digital Globe Imagery.

HANCE, F. E., 1929, Chemical treatment of hydraulic dam cores, Engineering News Record, (October 3, 1929), pp. 542–543.

KAUAIAN INSTITUTE, 2011, Map of Kauai Precipitation: Kauasian Institute, Kapa’a, Kaua’i.

MACDONALD, G. A.; DAVIS, D. A.; AND COX, O. C., 1960, Geology and Ground Water Resources of the Island of Kauai, Hawaii, Bulletin 13, Hawaii Division of Hydrography, 212 p.

MIDDLEBROOKS, T. A., 1942, Fort Peck Slide: ASCE Transactions, Vol. 107, pp. 723–764.

MITCHELL, J. K., 1993, Fundamentals of Soil Behavior, 2nd ed.: John Wiley & Sons, New York. 437 p.

MORGAN, A., 1951, The Miami Conservancy District: John Wiley & Sons, New York.

PLUMMER, F. L. AND DORE, S. M., 1940, Soil Mechanics and Foundations: Pitman Publishing Corp., New York. 473 p.

PROCTOR, R. R., 1933, New principles applied to actual dambuilding: Engineering News Record, Vol. 111, No. 12, pp. 372– 376.

SAVILLE, C. M., 1908, Gatun Dam investigations, appendix E: In Annual Report of the Isthmian Canal Commission: Government Printing Office, Washington, DC. pp. 127–196, pls. 62– 173.

SEED, H. B., 1979, Considerations in the earthquake-resistant design of earth and rockfill dams: Geotechnique, Vol. 29, No. 3, pp. 215–263.

STEARNS, H. T., 1964, Ground Water Supplies for Pioneer Paper Mill Co.: Report for Amfac Sugar Company, Honolulu, HI. TERZAGHI, K., 1943, Theoretical Soil Mechanics: John Wiley & Sons, New York.

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Failure of the Alexander Dam UNIVERSITY OF HAWAII, 2001, Economic History of Hawaii: University of Hawaii, Honolulu.

UNIVERSITY OF HAWAII, 2017, McBryde Sugar Company, Kauai: Hawaiian Sugar Planters’ Association Plantation Archives, Hawaiian Collection, M˜anoa Library, Honolulu: Electronic document, available at http://www2.hawaii.edu/∼speccoll/ p_mcbryde.html U.S. BUREAU OF LABOR STATISTICS, 2017, Consumer Price Index (CPI) Inflation Calculator: Electronic document, available at https://www.bls.gov/data/inflation_calculator.htm U.S. GEOLOGICAL SURVEY, 2007, Geologic Map of the State of Hawai‘i, Sheet 2—Island of Kaua‘i: Open

File Report 2007-1089, Sheet 2 of 8, by Sherrod, D. R., Sinton, J. M., Watkins, S. E., and Brunt, K. M.: Electronic document, available at http://pubs. usgs.gov/of/2007/1089 U.S. GEOLOGICAL SURVEY, 2017a, Kauai Topographic Map, 1:100,000.

U.S. GEOLOGICAL SURVEY, 2017b, Koloa, HA Quadrangle, 7.5 Minute Topographic Quadrangle Map, 1:24,000.

WEGMANN, E., 1922, The Design and Construction of Dams: John Wiley & Sons, New York. 740 p.

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Discovering the Polaris Fault, Martis Creek Dam, Truckee, California LEWIS E. HUNTER U.S. Army Corps of Engineers, Sacramento, CA 95814

RONN S. ROSE U.S. Army Corps of Engineers, Sacramento, CA 95814

BRUCE HILTON∗ Kleinfelder, Sacramento, CA 95827

WILLIAM MCCORMICK Kleinfelder, Santa Rosa, CA 95407

TODD CRAMPTON GEI Consultants Inc., Oakland, CA 94612

Key Terms: Dams, Engineering Geology, Foundations, Geotechnical, Remote Sensing, Site Investigations ABSTRACT Martis Creek Dam, located in the Truckee Basin north of Lake Tahoe, CA, was initially rated as one of the U.S. Army Corps of Engineers’ highest risk dams in the United States. While the dam has performed its flood control purpose, a history of excessive seepage during even moderate reservoir levels has prevented it from also fulfilling its potential water storage function. During seepage and seismic studies to assess and mitigate deficiencies, high-resolution light detection and ranging (LiDAR) data were obtained. This imagery provides an unprecedented representation of the ground surface that allows evaluation of geomorphology even in areas with a dense vegetation canopy. At Martis Creek Dam, this geomorphic analysis resulted in the recognition of a previously unknown and through-going lineament between the spillway and dam embankment. This feature extends to the southeast, where several lineament splays are exposed on the East Martis Creek Fan. These lineaments were subsequently explored by paleo-seismic trenching at two locations and confirmed as faults with Late Quaternary to Holocene displacement. Faulting was confirmed in both trenches as unique splays of a fault zone with several feet of apparent normal (vertical) slip and an un-

∗ Corresponding

author email: BHilton@Kleinfelder.com.

known magnitude, but a potentially significant, strikeslip component. Faulting was observed near the ground surface in both cases, and multiple fault events (a minimum of two) are interpreted as at least latest Pleistocene in age, and probably active in the Holocene.

INTRODUCTION Martis Creek Dam is owned and operated by the Sacramento District, U.S. Army Corps of Engineers (USACE). The dam and reservoir are located approximately 3 mi (4.8 km) east of Truckee, CA, on Martis Creek, a tributary to the Truckee River (Figure 1). The function of the dam is to provide flood control and possibly future water supply. However, heavy seepage coincident with even moderate pool elevations during test fillings have not allowed the dam to be used as a water supply source. Recent studies by the USACE have been designed to address deficiencies including seepage, spillway inadequacy, and potential liquefaction/stability issues. Preliminary modeling of dam breach scenarios concluded that there is potential for large impacts to downstream areas and facilities, including inundating large portions of the RenoSparks metropolitan area, railroads, bridges, and an interstate highway. The USACE began evaluation of their U.S. portfolio of dams based on performance, probability of failure, and consequence of failure in 2006.

The evaluations are referred to as Screening Portfolio Risk Assessments (SPRA), and as a result of this preliminary screening, the USACE designated

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Figure 1. Site location map.

Martis Creek Dam as one of the highest risk Corpsadministered dams in the United States. Martis Creek Dam was ranked Dam Safety Action Class (DSAC) 1, requiring “urgent and compelling” action. Further evaluations have led to the dam now being classified as a DSAC 3 dam. In response to these seepage deficiencies, the dam is now operated to allow only flood control operation and no water storage function.

Martis Creek Dam is a 113-ft-high (34-m-high) zoned, rolled earth-fill dam constructed in 1972. It has a maximum storage capacity of 20,400 acre-feet (25,162,992 m3 ) at the spillway crest (gross pool elevation 5,842.0 ft [1,780.6 m], North American Vertical Datum [NGVD] 29) and 34,600 acre-feet (42,678,408 m3 ) at the maximum spillway design flood pool elevation (Figure 2). The dam consists of three zones: (1) an upstream “impervious” zone, (2) a downstream zone (zone 2) identified as “random” fill, and (3) a vertical to horizontal drain as shown in Figure 3. The design includes no seepage cutoff. Instead, a clay blanket was constructed upstream of the dam at the base of the reservoir as a barrier to strengthen the seepage pathway against under-seepage.

The embankment foundation was stripped to a nominal depth of about 12 in. (30.5 cm) on the abutments, with the soft alluvium in the valley floor removed to a maximum depth of about 6 ft (1.8 m) and an average depth of about 3 ft (∼1 m; USACE, 1972) to a lowpermeability foundation layer referred to as the “Blue Silt” zone.

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A 250-ft-long (76-m-long) concrete emergency spillway is located at the left abutment. The uncontrolled spillway sill is at elevation 5,842 ft (1,780 m). The dam crest at elevation 5,862 ft (1786 m) provides 20 ft (6 m) of freeboard against a probable maximum flood event.

The outlet works consist of a vertical concrete intake shaft, a gatehouse containing two gated passages and operating chamber, a 4 ft (1.2 m) square reinforced concrete conduit combined with a 4 ft by 8 ft (1.2 m by 2.4 m) access gallery, a utility house, a stilling basin, and a return channel to carry discharges to the natural streambed.

Figure 2. Aerial photograph showing Martis Creek Dam facilities.

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Polaris Fault, Martis Creek Dam, CA

Figure 3. Maximum cross section of Martis Creek Dam.

REGIONAL GEOLOGY

TECTONIC SETTING

The Truckee Basin, including Martis Valley and the main branch of the Truckee River east of the town of Truckee, CA, is generally composed of a series of Pleistocene glacial deposits (Birkeland, 1964) overlying partially indurated Pliocene to Pleistocene age sediments.

Figure 4 shows the regional distribution of these geologic units and mapped major, regional faults (Saucedo, 2005).

Middle to Late Pleistocene moraine deposits flank the Truckee River in the western half of the basin, while contemporaneous glacial outwash terraces dominate the eastern half. The oldest recognized glaciogenic deposits and corresponding outwash in the Truckee Basin are the locally named “Donner Lake” age deposits, followed by “Tahoe” age deposits, and finally the youngest “Tioga” age deposits. During the various glacial stages, sediment-laden braided streams aggraded thick sequences of alluvium (glacial outwash terraces) that blanketed the bottom of the basin downstream of the corresponding glacial margin, leaving a relatively smooth outwash plain. The highest terrace on the basin margins corresponds with the oldest Donner Lake glaciation, and the subsequent Tahoe and Tioga glacial deposits are present at progressively lower elevations (Birkeland, 1964). The left abutment of the dam rests on Donner Lake age outwash deposits, and the right abutment of the dam is situated on Pliocene age andesite and dacite volcanic flows dated at 1.36 Ma (Sylvester et al., 2008). Prosser Creek Formation sediments underlie the outwash deposits and interfinger with the Pliocene and Pleistocene volcanic flows beneath the right abutment. A thin veneer of alluvium was also present beneath the reservoir and along Martis Creek beneath the maximum section of the dam prior to removal during construction.

The Truckee Basin straddles the boundary between two geologic provinces: the Sierra Nevada micro-plate and the Basin and Range Province. The Sierra Nevada–Great Basin boundary zone (SNGBBZ; van Wormer and Ryall, 1980) is a complex zone of active northwest-southeast–oriented dextral shear (Hammond and Thatcher, 2007) that accommodates 20 to 25 percent of the Pacific–North American plate motion at this latitude. Spatially overlapping domains of north-striking normal faults (typically associated with the Sierra Nevada Frontal Fault Zone) and northweststriking dextral strike-slip faults (typically associated with the Walker Lane Belt) occur within the SNGBBZ. The oblique dextral shear or dextral transtension is accommodated by alternating modes of faulting with east-west extension on normal faults and north-south contraction on the strike-slip faults (Schweickert et al., 2004).

In addition to mapping of regional geology, Saucedo (2005) compiled available regional fault mapping (Figure 4) that showed the through-going Dollar Point Fault System west of the site and a due-north fault system passing through the reservoir and main section of the dam. Figure 5 shows the historic earthquake activity in the northern Lake Tahoe Basin area. This figure shows the relatively low seismicity, with some clustering evident in the Agate Point area along the north shore of Lake Tahoe and northwest of Martis Creek Dam. The Advanced National Seismic System is an earthquake database that provides data for the U.S. Geological Survey (USGS) National Fault Database (USGS Earthquake Hazards Program, 2009), also shown in Figure 6, and it shows the very low frequency and magnitude of earthquake activity in the area of the dam. Recorded faulting in the north Tahoe Basin

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Figure 4. Regional geology (Saucedo, 2005).

identifies little to no faulting in the vicinity of Martis Creek Dam. GEOLOGIC AND GEOTECHNICAL SITE INVESTIGATIONS The Martis Creek Dam site was initially investigated in 1928, when two test pits were hand dug by the U.S. Bureau of Reclamation. The logs for these trenches are unavailable. The USACE took a greater interest in the site starting in 1958, drilling three diamond core holes. Site foundation investigations began in 1964 and consisted of 64 boreholes using various methods and 24 trenches.

Site geologic and foundation investigations identified potential foundation seepage issues beneath the left abutment of the dam as well as at the spillway. Drainage 114

control measures were incorporated both internal to the dam and downstream of the dam to control seepage, and an impervious clay blanket was placed upstream of the dam to limit seepage. Other controls included encapsulating natural springs that would have been buried beneath the left embankment. These seepage control measures were apparently inadequate, because uncontrolled seepage was judged as excessive during test fills in the 1970s and 1980s. Modifications to the left abutment seepage control measures were made from 1972 to 1995 but were not completely successful, although seepage improvements were observed.

Because of repeated seepage events, even at lower pool elevations, investigations were undertaken by the USACE to identify potential seepage routes and potential mitigations. Subsurface investigations as part of the deficiency investigation began in 2007. These

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Polaris Fault, Martis Creek Dam, CA

Figure 5. Historic earthquakes since 1900 in Tahoe Basin (USGS Earthquake Hazards Program, 2009).

investigations have consisted of hollow stem auger and sonic bore holes at various locations around and on the dam. Geophysical investigations have also been conducted by the USGS and GeoVision using seismic refraction/reflection, direct current (DC) resistivity, selfpotential, and time-domain electromagnetic methods.

The results generally enhance stratigraphic interpretations from the 1964 investigations.

Another significant observation made during the subsurface investigation in 2008 was the presence of vertical stratigraphic displacements of approximately 45 to over 55 ft (13.7 to >16.8 m) in the otherwise relatively flat-lying “Blue Silt” at the locations shown on Figure 7. These vertical “offsets” occurred in the approximate location of the Polaris Fault (see geomorphology discussions), where it is projected to pass between the spillway and the dam. At least three fault strands would be necessary to account for the observed vertical displacement of the “Blue Silt” marker bed.

Disruptions in the DC resistivity data also correspond

to the inferred locations of the fault between the spillway and left abutment (Figure 8).

GEOMORPHOLOGY AND PALEO-SEISMIC TRENCHING In January of 2008, the USACE purchased nine 1 mi2 (2.6 km2 ) tiles of light detection and ranging (LiDAR) data from the Truckee-Donner Public Utility District. These were acquired with the intent of using these data to generate 2 ft (0.6 m) contour maps of the USACE property around Martis Creek Dam (Figure 9a). These included LiDAR data in the form of first return, bare earth, canopy, and intensity data sets. In addition, ortho-rectified aerial photography was provided for the same grids. During the initial inspection of these data, a prominent lineament was observed to cross a prominent alluvial fan. This alluvial fan was considered to be relatively young and locally known as the East Martis Creek Fan (EMCF). The lineament

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Figure 6. Existing mapping of faults in the northern Lake Tahoe region (USGS Earthquake Hazards Program, 2009).

Figure 7. Offsets in “Blue Silt” horizon during geotechnical investigations.

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Figure 8. Locations of geophysical (resistivity and refraction) anomalies along fault.

was judged to likely be a fault-related scarp (Figure 9b) based on vegetation contrasts and apparent surface deformation. These lineaments could be traced from the southern boundary of the imagery and projected northwards towards Martis Creek Dam. Reconnaissance-scale Quaternary geologic mapping and geomorphology analyses were also performed in the area east and south of Martis Creek Reservoir and extending across the EMCF to the adjacent highway to the south (Figure 10).

This new information led to further evaluations by the USACE and USGS (Hunter et al., 2009). On the bases of imagery, LiDAR data, and geomorphology, several candidate sites were identified for paleo-seismic trenching. Two were unavailable on the basis of right of entry and/or access issues, leaving two key sites for evaluation. Trench site T-1 was coincident with an area of the distal portion of the EMCF at a location where a strong vegetation contrast was visible in both aerial imagery (Figure 11) and on the ground (Figure 12).

Trench site T-2 was located further up the fan near the creek and was coincident with a well-defined scarp strongly evident on the LiDAR data. Both sites corresponded to anomalies observed in reconnaissance geophysical data. These were expressed as sharp boundaries in the magnetic data and disruptions in reflectors observed in the ground-penetrating radar data.

In the fall of 2008, the Kleinfelder/Geomatrix Joint Venture executed trenching operations on two of the locations identified by the USACE (Figure 13).

Paleo-seismic trenches were excavated and logged at the two sites in an effort to characterize the Polaris Fault with respect to its recency of activity, sense of slip, slip rate, and recurrence. Limited samples from pertinent stratigraphic units within the trenches were collected and submitted for laboratory bulk accelerometer mass spectrometry C-14 dating. However, the resulting age dates were judged to be unreliable due to potential contamination from near-surface organic sources.

Figure 9. (a) LiDAR image of Martis Creek Dam and vicinity without interpretation. (b) LiDAR image of Martis Creek Dam and vicinity with interpretive lineament.

The trench exposures confirmed the existence of faults coincident with the surface geomorphic features previously identified at each of the locations. The fault exposure in T-1 offset a thin basalt flow interbedded with alluvial and debris flow deposits of the EMCF in an apparent down-to-the-west (i.e., normal) sense of displacement (Figure 14). The total apparent vertical offset across the fault zone was about 2 ft (0.6 m). Two distinct earthquake events could be identified in T-1, with the most recent event rupturing the basal contact of the uppermost soil unit. Apparent vertical offset during this event was about 0.25 ft (0.08 m), whereas the penultimate event produced an estimated 1 ft (0.3 m) of apparent vertical offset. The horizontal offset could not be determined for either event.

The fault exposure in T-2 offset volcanic bedrock and the overlying fan deposits, also in an apparent downto-the-west (i.e., normal) sense of displacement (Figure 15). The minimum apparent vertical offset across the fault zone in T-2 was 5 ft (1.5 m). A minimum of four distinct earthquake events could be identified in T-2. Interpretation of the trench exposures suggests the sense of slip during the most recent event was predominantly lateral (strike-slip). Approximately 0.5 ft (0.15 m) of apparent vertical displacement occurred

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Figure 10. Quaternary geology and geomorphic mapping of Martis Creek fans and dam vicinity showing proposed paleo-seismic trench locations.

during this event; the horizontal displacement could not be determined. The maximum apparent vertical or horizontal offset occurring during the preceding three events could not be determined. Apparent thickening of the surface soil unit was observed across the fault zone, with no visible offset. Approximate slip rate or recurrence interval could not be determined from trench exposures. The results of these investigations were provided by Kleinfelder-Geomatrix (2009).

Relative topographic relationships between the EMCF and surrounding older alluvium and glacial 118

outwash surfaces, in conjunction with pedogenic soil development exposed in the trenches, suggest the faults in T-1 and T-2 are at least latest Pleistocene in age, and probably active in the Holocene. The stratigraphic and structural relationships exposed in the trenches provide evidence for a predominantly strike-slip sense of displacement on the Polaris Fault, with an apparent component of vertical down-to-the-west (i.e., normal) displacement. Prominent lineaments evident in the LiDAR images also suggest predominantly right-lateral strike-slip displacement, based on apparent dextral

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Figure 11. Aerial view of vegetation lineament and proposed T-1 paleo-seismic trench site.

offset of the Truckee River and a paleo-channel located nearby.

Figure 13. Paleo-seismic trenching and logging efforts at T-1 trench site.

CONCLUSIONS This article was prepared to highlight detailed studies by the USACE, Kleinfelder, and Geomatrix at Martis Creek Dam as part of a risk-informed hazards evaluation under the recent SPRA effort. Due to a prolonged history of seepage and concomitant concerns regarding the dam’s past performance, it was initially designated as a DSAC 1 dam, requiring “urgent and compelling” action. The discovery of the “Polaris Fault” is of major importance in terms of re-evaluating regional tectonic processes and, to a lesser extent, identifying a new hazard to the dam (seismic risk is low due to the lack of a permanent pool of water). In this light, the evaluation of Martis Creek Dam represents the state-of-the

Figure 12. Ground view of vegetation lineament and proposed T-1 paleo-seismic trench site.

art process being executed by the Sacramento District, USACE, through its Dam Safety Program in order to preserve safety of its infrastructure and the public. If deemed beneficial to further evaluation of the safety of Martis Dam, future investigations could be targeted to provide further definition of recency of faulting at locations shown on Figure 16.

Bulk soil charcoal samples were obtained from the fault trenches, and radiocarbon data were obtained.

Unfortunately, bulk results were subject to crosscontamination and thus are less reliable than discrete sampling results from charcoal samples. Quaternary

Figure 14. Trench T-1 north wall log relative to observed vegetation lineament.

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gations at or near the dam site. In addition to the USACE and Kleinfelder-Geomatrix team, the USGS and other students and geologists at UNR and NBM&G have contributed significantly to this effort. We specifically want to thank Jim Howle of the USGS for his assistance in the field and in the interpretation of the LiDAR imagery. We wish to express our appreciation to the Truckee-Donner Land Trust for their willingness to work with us to gain access to protected conservation lands. With their partnership, we were able to access trench sites that provided valuable paleo-seismic data with the least amount of disturbance to protected environmental sites.

Figure 15. Trench T-2 north wall log relative to observed scarp.

geologic mapping and geomorphic analysis of the overall Polaris Fault Zone will be used to identify candidate sites for future paleo-seismic trenching that may provide better age dating, as well as evidence of multiple events that can be used to compute slip rate estimates and better constrain the displacement direction and magnitude per event.

ACKNOWLEDGMENTS The Martis Creek Dam studies discussed in this article are the result of a large number of pre-existing geologic and tectonic findings and more recent investi-

Figure 16. Future potential sites for paleo-seismic trenching.

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REFERENCES BIRKELAND, P. W., 1964, Pleistocene glaciation of the northern Sierra Nevada, north of Lake Tahoe, California: Journal of Geology, Vol. 72, pp. 810–825.

HAMMOND, W. C. AND THATCHER, W., 2007, Crustal deformation across the Sierra Nevada, northern Walker Lane, Basin and Range transition, western United States measured with GPS, 2000–2004: Journal of Geophysical Research, Vol. 112, pp. B05411.

HUNTER, L. E.; HOWLE, J. F.; ROSE, R. S.; AND BAWDEN, G. W., 2009a, The “Polaris Fault”: A previously unmapped fault discovered using LiDAR near Martis Creek Dam, Truckee, CA: Seismological Research Letters, Vol. 80, No. 2, pp. 305.

HUNTER, L. E.; ROSE, R. S.; HOWLE, J. F.; BROWNE, V. W.; POPWERS, M. H.; HILTON, B. R.; AND HUBBARD, E., 2009b, Geotechnical and paleoseismic investigations of the Martis Creek Dam, Truckee, California. In AEG 2009 Field Trip Guide Book.

KLEINFELDER-GEOMATRIX, 2009, Martis Creek DSAC Fault Trenching Interim Report, Nevada County, California. Sacramento, California: Kleinfelder-Geomatrix Joint-Venture Report, prepared for Sacramento District, U.S. Army Corps of Engineers under Contract No. W91238-08-D-0015, Task Order 0009, July 2009.

SAUCEDO, G. J., 2005, Geologic Map of the Lake Tahoe Basin, California and Nevada: California Geological Survey Regional Geologic Map No. 4, scale 1:100,000.

SCHWEICKERT, R. A.; LAHREN, M. M.; KARLIN, R. E.; SMITH, K. D.; HOWLE, J. F.; AND ICHINOSE, G., 2004, Transtensional deformation in the Lake Tahoe region, California and Nevada, USA: Tectonophysics, Vol. 392, pp. 303–323.

SYLVESTER, A. G.; WISE, W. S.; HASTINGS, J. T.; AND MOYER, L. A., 2008, Digital Geologic Map of the Tahoe-Donner Pass Region, Northern Sierra Nevada, California: U.S. Geological Survey Data Series 08-xxxx (Draft Version 6L07).

U.S. Army Corps of Engineers (USACE), 1972, Martis Creek Dam and Appurtenances, Truckee River Basin, Nevada and California: Foundation Report, November 1972..

U.S. Geological Survey (USGS) Earthquake Hazards Program, 2009, United States Fault Database: Electronic document, available at https://earthquake.usgs.gov/hazards/ qfaults/ VAN WORMER, J. D. AND RYALL, A. D., 1980, Sierra Nevada–Great Basin Boundary Zone: Earthquake hazard related to structures, active tectonic processes, and anomalous patterns of earthquake occurrence: Seismological Society of America Bulletin, Vol. 70, pp. 1557–1572.

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Grout Curtain Construction at Bolivar Dam, Ohio MICHAEL C. NIELD∗ U.S. Army Corps of Engineers, 502 Eighth Street, Huntington, WV 25701

Key Terms: Grout Curtain, Foundation, Dam, Engineering Geology, Bedrock Permeability, Grout, Seepage Barrier Wall, Internal Erosion ABSTRACT Bolivar Dam, located in eastern Ohio, is an embankment dam constructed by the U.S. Army Corps of Engineers in 1937 as part of the Muskingum River Basin Project for flood control. As a result of seepage concerns observed at the dam during several flood events over the life of the project, seepage reduction measures, including a partial-depth seepage barrier wall through the embankment and a grout curtain in the left abutment, were designed and constructed. These dam safety modifications were constructed between 2014 and 2016. During flood events, Bolivar Dam experiences excessive seepage through the glacial outwash foundation as well as through a network of open joints within the left bedrock abutment.

Seepage in the bedrock abutment could erode/scour the dam embankment at the bedrock contact, potentially leading to dam failure. To lower this potential risk of dam failure in the left abutment, a grout curtain was constructed between the new seepage barrier wall and an existing grout curtain across the emergency spillway.

The new grout curtain is designed to impede groundwater seepage, resulting in reduced groundwater velocity/ energy downstream of the grout curtain, thereby decreasing its potential to scour or transport fine-grained embankment material. The double-line grout curtain is approximately 65 ft (19.8 m) deep and 400 ft (121.9 m) long and was completed in November 2015 as part of a major dam safety modification project. Two thin limestone units encountered during drilling proved to be problematic and posed various challenges during construction.

It was common during drilling to lose water circulation within the vicinity of these limestone units, which then required the use of downstaged grouting methods. The majority of the grout volume for the project was placed within these downstaged intervals. This article presents the risk-informed decisions that were made during both design and construction of the grout curtain and includes various lessons that were learned during this process.

∗ Corresponding

author email: michael.c.nield@usace.army.mil.

SITE INFORMATION Bolivar Dam is located on Sandy Creek, a tributary within the Muskingum River Watershed (Figure 1) in Ohio. Bolivar Dam is part of a system of 16 dams located within the Muskingum River Watershed that were built by the U.S. Army Corps of Engineers (USACE), mostly in the 1930s. This system of dams was the nation’s first to impound water for regional flood control. Bolivar Dam does not maintain a permanent pool and is considered a “Run of River” structure, such that the river flows unhindered through the outlet works during normal flows. However, during flood events, Bolivar Dam retains a pool to reduce flooding downstream.

Bolivar Dam is an embankment dam, having an approximate length of 6,300 ft (1,920 m) and a maximum height of 87 ft (26.5 m), with a crest elevation of 982 ft (299.3 m) (Figure 2). The dam is a two-zoned earth fill embankment with an impervious core and pervious upstream and downstream shells. The dam is founded mostly on thick deposits of glacial outwash throughout

Figure 1. Project location map.

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Figure 2. Aerial view—site plan.

the valley bottom; however, the dam’s left abutment is founded on bedrock. The outlet works (including two tunnels constructed in bedrock) and spillway on the left abutment and are also founded on (or in) bedrock.

SITE GEOLOGY The dam site is located within the unglaciated Appalachian Plateau physiographic province. Although the dam site is unglaciated, it is located less than 5 miles downstream from the remnant ice margins of both the Illinoian and Wisconsin glacial stages of the Pleistocene Epoch. As a result of the proximity to these continental glaciers, the 190-ft (57.9-m)–thick alluvial sediments within the valley bottom consist primarily

of glacial outwash with occasional evidence of glacial lake deposits. The glacial outwash material consists primarily of pervious sands of variable amounts of silt and interlayered gravel strata with variable thicknesses and extents. Gravel strata consist of both poorly and well-graded materials. The dam foundation throughout the valley bottom is primarily founded on this glacial outwash material. Glacial lake deposits consist of occasional, discontinuous, thin layers of clays and silts.

The top-of-rock surface rises steeply at the left abutment to within several feet of the ground surface. Thin soil, 1-17 ft (1.3 to 5.2 m) thick, on the left abutment consists of fine-grained colluvium and fill material. The bedrock at the left abutment consists of near-horizontal sedimentary rock of the Pennsylvanian-aged Pottsville Group (Figure 3). The Pottsville Group at the dam site consist of interbedded members of sandstone, shale, siltstone, claystone, and thin beds of coal and two thin limestone units. Bedrock discontinuities consisted primarily of near-horizontal bedding planes and highangle joints with surfaces that are typically smooth and planar. The constructed grout curtain, the subject of this article, is located through the left abutment.

Two limestone members proved to be problematic during foundation grouting, the Upper Mercer Limestone, at an approximate elevation 965 ft (294.1 m), and the Lower Mercer Limestone, at an approximate elevation 935 ft (285 m). These limestone units are thin (3 to 5 ft thick, or 0.9 to 1.5 m), laterally continuous, dark gray, occasionally fossiliferous, very hard, and exhibit a maximum unconfined compressive strength of 32,000 psi (221 MPa). The limestone units are highly jointed, with high-angled joint spacing typically ranging from 3 to 5 ft (.091 to 1.5 m) and with a distinct 90◦ or

Figure 3. Geologic profile at left abutment, viewed looking upstream.

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Figure 4. Photo of Lower Mercer Limestone outcrop and optical televiewer image.

orthogonal joint set orientation (general upstreamdownstream direction and parallel to the dam axis) (Figure 4). High-angled joints and near-horizontal bedding planes are often solutioned with apertures ranging from 0.5 to 4 in. (1.2 to 10.2 cm) (Figure 4).

These open discontinuities are often interconnected, forming conduits for significant groundwater flow during high-pool events (Figure 5).

POOR PERFORMANCE—SEEPAGE During three historical high-pool events, excessive uncontrolled seepage occurred within the bedrock

above the downstream outlet works, located on the left abutment of the dam. These three events included one in winter of 1991, with maximum pool elevation of 949 ft (289.3 m); the pool of record during the winter of 2005, with maximum pool at an elevation of 952 ft (290.2 m); and one in spring of 2011, with maximum pool at an elevation of 946.5 ft (288.5 m). Initial seepage in the vicinity of the outlet works was noticed once the pool reached an elevation of 943 ft (287.4 m) and above (Figure 6). The seepage path is suspected to travel through a series of interconnected bedrock discontinuities within the Lower Mercer Limestone. This groundwater movement posed a safety risk because of its

Figure 5. Hypothetical seepage through bedrock joints.

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Figure 6. Photos of left abutment seepage during high-pool events.

potential to erode the embankment, where it is in contact with open discontinuities, then transporting this embankment material through the network of open joints to the ground surface (Figure 5). It was determined that foundation treatment would be implemented to reduce this dam safety concern; this treatment is the subject of this article.

Excessive seepage downstream of the dam was also observed within the valley bottom during several highpool events. During these events, large boils and uncontrolled seepage, from the foundation glacial outwash filling the valley bottom, were observed downstream of the dam and at the toe of the dam embankment. This seepage necessitated an analysis of the dam for potential internal erosion of the glacial outwash foundation below the dam embankment, which could lead to dam failure. This analysis determined that sufficient risk of dam failure existed as to warrant corrective action, and a USACE Dam Safety Action Classification of 2 (“high urgency”) was assigned.

DAM SAFETY MODIFICATION Previously completed remedial measures constructed to address and monitor seepage concerns through the glacial outwash foundation consist of relief wells, filters, blankets, and in situ instrumentation. To further address seepage concerns through the glacial outwash, within the valley bottom, a partial-depth seepage barrier wall was constructed from 2014 to 2016 to retard groundwater flow though the highly pervious sands and gravels, eliminate potential piping pathways, and reduce excess hydraulic head. The seepage barrier has an approximate 144-ft (43.9-m) depth, 4,518.5-ft (1,377-m) length, and minimum 2-ft (0.7-m) width. The backfill material was required to be continuous and homogeneous, with a hydraulic conductivity of less than 1 × 10−6 cm/s, and a minimum unconfined compressive strength of 750 psi (5.2 MPa). This partial-depth seepage barrier wall constitutes a significant portion of the work effort for this project, relative to the grout curtain.

The left terminus of the seepage barrier was designed to be socketed into the left abutment bedrock for an approximate 140-ft (42.7-m) length, with a wall height 124

ranging from elevation 924 ft (281.6 m) to the top-ofground surface. This additional length of the seepage barrier is intended to cut off potential seepage pathways through bedrock that are located near the top-of-rock surface and in close proximity to the dam embankment.

The grout curtain overlaps the terminus of the seepage barrier from grout-line stationing −0 + 10 to 0 + 00.

The grout curtain alignment extends from the seepage barrier, through the left abutment bedrock, and terminates at the existing emergency spillway at grout-line stationing 3 + 84.

GROUT CURTAIN DESIGN To address seepage concerns through bedrock, a grout curtain was proposed and constructed, which is the subject of this article. The intent of this grout curtain was to inhibit groundwater flow through bedrock discontinuities, downstream of the grout curtain, during certain high-pool events. The new grout curtain was designed to reduce the groundwater velocity and, therefore, its capacity to erode/scour embankment material near the embankment/rock interface, and it should lower the potential for groundwater to transport finegrained material through the network of open joints to the surface.

A double-line grout curtain design was selected, with 10-ft (3-m) spacing between opposing lines, to provide redundancy and to increase assurance of creating a low-permeability curtain. The grout curtain alignment was selected to connect the terminus of the newly constructed seepage barrier wall with an existing grout curtain (constructed in 1989) that is located along the spillway sill (Figure 2). This alignment results in a grout curtain length of 395 ft (120 m), with a 10-ft (3-m) overlap with both the seepage barrier and the existing spillway grout curtain. Grout curtain depth is approximately 65 ft (19.8 m) from the top-of-ground surface to the intended bottom elevation of 924 ft (281.6 m), so that the lowest limestone unit would be penetrated by the grout holes.

The grout curtain is separated vertically into two zones designed to isolate the two limestone units and more adequately grout these two units separately: Zone 1 is located between top-of-rock surface to elevation 950 ft (289.6 m), and Zone 2 is located from elevation 950 ft (289.6 m) to the bottom of the grout curtain at elevation 924 ft (281.6 m). Grout holes were typically inclined at 30◦ from vertical to increase the probability of intersecting high-angled bedrock joints. The holes were inclined in opposition to one another along the two grout lines, such that holes along the upstream grout line are inclined toward the right abutment and those along the downstream line toward the left abutment.

This opposing inclination of the holes increases the

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Figure 7. Photos of drilling operations.

probability of intersecting a near-vertical joint within the bedrock. The minimum grout hole spacing was established at 10 ft (3 m). This decreased hole spacing was necessary to increase the probability of intersecting high-angled fractures within the thin limestone units. See Figures 11 and 12 for a profile of the grout curtain showing hole orientations for both upstream and downstream grout lines with hole spacing. A splitspaced method was used to locate each higher order of grout holes along the grout line (Tertiary, Quaternary, and Quinary). This method represents standard practice to promote adequate grout curtain closure.

CONSTRUCTION CONTRACT A firm-fixed price contract to construct the seepage barrier wall and grout curtain at Bolivar Dam was advertised, with the successful bidder selected using best-value procedures. The contract was awarded to Treviicos South, Inc., with an initial construction cost of $44.2 million and a “Notice to Proceed” date of May 2014. A performance-based specification was used for the seepage barrier wall construction, which is the primary component of the project. Treviicos South, Inc., selected a panel method of construction for the seepage barrier wall, which was primarily excavated utilizing hydromills. Construction of the grout curtain on the left dam abutment, the subject of this article, was subcontracted to TerraFirm Construction, LLC. GROUT HOLE DRILLING The grouting subcontractor, TerraFirm Construction LLC, began drilling in June 2015. Grout hole inclinations were typically drilled at 30◦ ; however, inclinations ranged from 0◦ to 70◦ at the locations of the radiating fan of holes at the terminus of the grout lines. Grout hole lengths ranged from 25 ft (7.6 m) to 110 ft (33.5 m), with typical hole depth being approximately 65 ft (19.8 m). Drilling was performed using one of three crawler-mounted rotary drill rigs (Figure 7). Upon completion of drilling, the grout hole alignments were measured primarily using a Gyro-Shot tool.

The upstage method of grouting was specified, which required the hole to be drilled to full depth and grouted from the bottom to the top of the hole in various stages, which typically increases productivity and lowers costs.

However, if excessive drill water loss (>50 percent) was encountered, the downstage grouting method was utilized. The downstage method requires drilling of the hole to be stopped at the point of excessive drill water loss; it also requires that the hole be hydraulically pressure tested and then grouted prior to extending the hole to the designed bottom elevation.

To reduce the risk of hydrofracturing within the soil, grout holes were drilled through soil without the use of drilling fluids. To accomplish this, the contractor utilized hollow-stem auger drilling methods to drill through the 15- to 21-ft (4.6- to 6.4-m)–thick soil and socket a minimum of 2 ft (0.6 m) into bedrock. Casing was installed to prevent collapse of the hole, prevent contamination from surface water, and to isolate the soil from drilling fluids and injected grout. A grout seal was placed between the casing and bedrock, and the annulus between the casing and soil was grouted in two separate lifts to reduce the risk of hydrofracturing. A total of 2,122 linear feet (646.8 m) was drilled through overburden for the grout curtain.

Grout hole drilling through bedrock was accomplished utilizing crawler-mounted drills with wateractuated down-hole hammers. Grout holes in bedrock were specified to have a minimum 3-in. (7.6-cm) diameter. Grout holes were logged by a qualified geologist and were incorporated into the data management system and displayed using a Web-based GIS format.

Once the hole was drilled to the design tip elevation, the bedrock portion of the hole was washed with water during a separate operation using a high-pressure nozzle. Upon completion of drilling and cleaning, the grout hole was pressure tested and grouted. Down-hole optical televiewer images of grout hole sidewalls were obtained from select holes, which typically included holes with large grout takes or those adjacent (Figure 9). A total of 8,889 linear feet (2,709.4 m) of bedrock were drilled and grouted on this project.

DRILLING CHALLENGES Upstage grouting was the principal method specified for this project. However, along the upstream grout line, a large number (90 percent) of primary order grout holes encountered significant loss of drill water return prior to reaching the design bottom depth, which necessitated a shift to downstage grouting. Many of these primary order holes had multiple downstages prior to reaching the design hole depths. The excessive loss of drill water was frequently the result of drilling into the Upper and Lower Mercer Limestone units, which

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proved problematic. These units were penetrated by a network of open discontinuities, which led to loss of drill fluid circulation and required utilizing downstage drilling and grouting methods. Drill water loss was most notable at the near-horizontal lithologic contact between the uppermost claystone and Upper Mercer Limestone.

Bedrock encountered at Bolivar Dam included sedimentary rock of varying strengths and degrees of weathering. In addition to high-strength limestone and sandstone members, soft claystone units and thin coal seams were also encountered. Some of these units were soft, highly fractured, and severely weathered. Occasionally, the grout hole sidewalls along these weaker units would partially cave in or slough into the angled holes during drilling or washing operations. This cavein material was removed by either re-drilling through the material or by washing out the loose material.

DRILLING LESSONS LEARNED In open foundation conditions, consideration should be given to specifying the primary order hole being advanced utilizing downstage grouting methods in short stages/zones (5–15 ft or 1.5 to 4.6 m). However, secondary and higher-order holes could then be grouted using the upstage grouting method to increase productivity. In addition, consideration should be given to reducing or eliminating future requirements for a separate washing operation through weak seams, which did not significantly benefit from such operation.

PRESSURE TESTING Once the grout holes were drilled and washed, the grout holes were then hydraulically pressure tested.

Pressure testing consisted of injecting water under a designated pressure into an isolated length of grout hole while measuring the volume and rate of water being injected. Pressure testing was required to contribute toward cleaning discontinuities prior to grouting, assist in determining characteristics of bedrock discontinuities, assist in estimating the secondary permeability of the bedrock along the grout hole, and help anticipate grouting behavior. Pressure testing was completed by utilizing one of two track-mounted grout carts equipped with a hose, hose coil, down-hole packers, pressure transducers, and flow meters (Figure 8). Water was delivered by progressive cavity pump (Moyno pump).

Typically, while using upstage grouting methods, Zones 1 and 2 would be hydraulically pressure tested separately. However, where downstage grouting methods were required, separate pressure tests were performed for each downstaged interval. The designated interval along the grout hole, either zone or stage, was 126

Figure 8. Photos of cart used for pressure testing and grouting and automated system.

isolated for testing utilizing a single or double packer.

Water was injected into this interval under a predetermined pressure. The pressures used were established to prevent potential hydrofracturing of the bedrock.

The maximum allowable effective pressure was established at the mid-point of the stage or zone and was set at the approximate confining pressure of the overlying materials, 0.5 psi/ft (3.4 kPa) for overlying soil and 1 psi/ft (6.9 kPa) for overlying bedrock. The gauge pressure was then determined by considering the allowable effective pressure, groundwater level, hydraulic pressure exerted by the column of water within the grout hole, and the frictional losses within the hoses and lines of the pressure testing apparatus. Pressures were measured by pressure transducers, and water flows were measured using a flow meter, both located on the grout cart apparatus (header).

Water pressure and flow data were recorded using an automated grouting control and data collection system. These data were collected into a Cinaut 15 system, which is manufactured by Jean Lutz SA. This system controls the pumps and regulates flow by utilizing the data from pressure transducers. The pressure and flow data are then managed and recorded with the Cinaut 15 system, which provides a continuous data read-out of gauge pressure, effective pressure, water flow, and Lugeon values relative to time. Pressure test data were incorporated in the data management system and displayed in a Web-based GIS format.

PRESSURE TESTING: LESSONS LEARNED Typically, pressure test duration at Bolivar Dam was only 2 minutes for an individual stage or zone. Consideration could have been given to increasing the duration of the test to better determine variations in flow rate, as recorded by the automated data collection system. For pressure tests that result in higher water takes, multiple tests can then be performed at varying pressures while comparing Lugeon values (Houlsby, 1976; ˜ Quinones-Rozo, 2010). These test results could assist

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Figure 9. Optical televiewer images of grout holes showing grout-filled discontinuities.

in determining the condition of the fractures that will be grouted.

GROUTING METHODS Once the holes were drilled and pressure tested, all of the holes were then grouted. Grouting consisted of injecting balanced and stabilized grout mixtures into the bedrock foundation while utilizing an automated system to collect the grouting data. This cementitious grout was used to fill, or partly fill, the open discontinuities within the bedrock that intersected grout holes along the curtain alignment (Figure 9).

Grout was mixed using a hydraulic grout plant equipped with a 100-gallon (0.38-m3 ) mixing tub with a colloidal mixer, 100-gallon (0.38-m3 ) holding tub with a paddle mixer, cement hopper with bulk feeder, sand hopper with screw feeder, bentonite mixer with storage tanks, scales, and flow meters (Figure 10). Grout was delivered by progressive cavity pumps (Monyo pump) to a track-mounted grout cart equipped with grout

lines, coil, pressure transducers, flow meters, and packers. Grout pressures and flows were monitored using a Jean Lutz computer grout monitoring system. The grouting data were then incorporated in the data management system and displayed in a Web-based GIS format.

Figure 10. Photo of grout plant during operation.

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Figure 11. Profile of upstream grout line showing all grout-hole orders.

Upstage grouting was the preferred method, as specified, unless excessive drill water loss was encountered, at which point the downstage method was utilized, as mentioned above. Grout was injected into an interval (zone or stage) along the grout hole, which was isolated utilizing single or double packers. Grout was injected at pre-determined pressures to prevent hydrofracturing within the bedrock. The allowable effective pressure and target gauge pressures used during grouting were calculated utilizing methods that were similar to pressure testing, described above. Typically, grouting was initiated within a zone or stage using the thinnest grout mix and changing to thicker grout mixes as foundation and grouting conditions warranted. Grouting was continued in a stage or zone until refusal criteria were achieved, which was defined as a grouting flow rate of 0.5 gallon/min (1.89 L/min), with this flow being held for 10 minutes at the targeted “effective pressure.” These refusal criteria typically resulted in “apparent” Lugeon values ranging from 1.1 to 3.5 Lu. Grouting started with primary order grout holes, located on 20-ft (6.1-m) centers, along the upstream grout line. Once primary grout holes were completed within a section of the upstream grout line, then secondary grout holes were drilled. Secondary holes were split-spaced between the primary holes or spaced 10 ft (3 m) from completed primary holes. All primary and secondary holes were required to be drilled and grouted. Most primary and secondary holes were grouted along the upstream grout line prior to drilling the first primary hole located on the downstream grout line.

Tertiary and higher-order grout holes were required if initial split-spacing criteria were encountered during grouting. These split-spaced criteria were specified as 100 gallons (378 L) of grout placed beyond 128

the theoretical volume of the hole, or as otherwise directed. If a secondary hole exceeded this 100-gallon (378-L) criteria, then tertiary holes were located on either side of the secondary hole, split-spaced 5 ft (1.5 m) from the secondary hole. If a tertiary hole exceeded the 100-gallon (378-L) criteria, then quaternary holes were split-spaced at 2.5 ft (0.8 m) between the tertiary holes. Only two quinary holes were required to be split-spaced on either side of a quaternary hole, spaced 1.25 ft (0.4 m) from the quaternary hole. See Figure 11 for all grout-hole orders drilled along the upstream grout line, and see Figure 12 for all grout-hole orders along the downstream grout line.

Six balanced stable grout mixes were designed for use at the Bolivar project; these mixes were identified as mixes A through F. These balanced stable grout mixes exhibit minimal bleed (specified to not exceed 1 percent bleed), and the mix rheology was designed to remain consistent throughout the grout injection process (specified pressure filtration coefficient value to not exceed 0.05 kPa). In addition, the grout mixes were designed to satisfy specified requirements for viscosity, final set time, and temperature. See Table 1 for field and lab grout test results and specified limits. Grout mix components consisted of varying amounts of water, Portland cement—Type III, bentonite, diutan gum, and super-plasticizer. The grout mixes were design to meet a specified range of water/cement ratio (specified range of 0.6 to 2.0) and bentonite content (0 percent to 8 percent). Grout mixes A through D varied sequentially in water/cement ratio and viscosities. Grout mix E contained a relatively higher quantity of diutan gum, but was rarely used. Sand was an added component to the F mix, which was primarily injected in grout holes that had open foundation conditions that took large volumes of grout.

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Figure 12. Profile of downstream grout line showing all grout-hole orders.

Table 1. Field and lab grout testing results.

Typically, grouting of a hole was initiated by injecting the thinnest grout mix (A mix). During grouting, if the target grouting pressure was not obtained or the grout take remained high with little to no decrease in grout flow rate then a sequence of thicker grout mixes was injected. Grout mix A was the primary mix used at Bolivar Dam, with sequential reduction of total grout volume for the thicker grout mixes. See Figure 13 for a graph showing the total volumes of grout injected by mix and grout-hole order. The sanded grout mix, mix F, was primarily used during downstage grouting with open foundation conditions.

Downstage grouting methods were utilized when open bedrock foundation conditions were encountered during drilling and significant drill water was loss was encountered (>50 percent). After pressure testing, the typical grouting sequence for these downstage conditions included injecting up to 200 gallons of A mix and then, if grout flow rates did not decrease, injecting up to 200 gallons of B mix, potentially followed by up to 200 gallons of C mix, up to 200 gallons of D mix, and up to 1,000 gallons of F sanded mix. If the grout flow

rate had not significantly decreased after placing the sanded grout mix, then 200 gallons of E mix was injected, and grouting would cease for a minimum of 12 hours. After this resting period, a determination was made to either resume grouting or continue drilling.

Figure 13. Graph of total grout injected by grout mix and grout-hole order.

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Figure 14. Location of downstage grouting relative to bedrock lithology.

GROUTING RESULTS The majority of the grout volumes injected at Bolivar Dam were placed in the downstaged intervals, many of which were located in the vicinity of the two thin limestone units (Upper and Lower Mercer Limestone).

See Figure 14 for locations of downstage grouting for both upstream and downstream grout lines combined.

An approximate total of 131,200 gallons of grout was injected into the foundation at Bolivar Dam, resulting in an average of 14.8 gallons per linear foot of bedrock drilling. See Figure 15 for a profile showing grout takes along the upstream grout line. See Figure 16 for a profile showing grout takes along the downstream grout line. The effectiveness of filling discontinuities and voids was documented with an optical televiewer (Figure 10).

Review of grouting data revealed that grout takes were reduced sequentially for higher-order grout holes, which is an indirect indicator that the bedrock permeability was reduced as grouting progressed. Sequential

reductions in grout takes were observed from the first order of holes that were grouted (primary order holes along the upstream grout line) at 63.7 gallons/linear foot (0.79 m3 /m), to the last order of holes that were grouted (quaternary order holes along the downstream grout line) at 0.3 gallons/linear foot (0.004 m3 /m) (Figure 17). Likewise, pressure test data also revealed sequential reduction for higher-order grout holes, which is a good indicator of bedrock permeability reduction as grouting progressed. Maximum reductions in pressure test results were observed between the first order of holes that were tested (primary order holes along the upstream grout line), having a median Lugeon value of 41.5 Lu, and the last order of holes that were grouted (quaternary order holes along the downstream grout line), at 1.5 Lu (Figure 17). The sequential reduction in grout takes and pressure testing results in higherorder grout holes was evident throughout the project, including along both the upstream and downstream grout lines and within both Zones 1 and 2 (Figure 18).

In addition, a sequential reduction in the number of

Figure 15. Grout takes along upstream grout line.

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Figure 16. Grout takes along downstream grout line.

intervals that required downstage grouting was also evident in higher-order grout holes (Figure 20). The upstream grout line was generally grouted prior to the downstream line and required a considerably greater number of grout holes and total gallons of grout injected (102 holes and 118,200 gallons, or 447.44 m3 ) compared to the downstream grout line (69 holes and 13,000 gallons or 49.21 m3 ). The sequential reduction in grout take and pressure test results, as grouting progressed, is another indicator that the foundation permeability was also reducing.

Bedrock permeability in the vicinity of the grout curtain was significantly reduced as a result of grouting.

Reduction of bedrock permeability, inferred from average Lugeon values, was determined for the various grout lines and zones. To achieve this, the bedrock permeability of the first-order grout holes (primary grout holes along the upstream grout line), representing the original bedrock permeability, was compared to that of the highest-order grout holes, representing the permeability of the completed grout curtain. Reduction of

bedrock permeability for Zone 2 was approximately 1 order of magnitude, from a permeability of 4 × 10−4 cm/s before grouting, compared with both the completed upstream grout line at 5 × 10−5 cm/s and the downstream grout line at 6 × 10−5 cm/s. Reduction of bedrock permeability for Zone 1 (from station −0 + 10 to 2 + 00) was approximately 1.5 orders of magnitude, from 5 × 10−3 cm/s before grouting, compared to both the completed upstream line at 2 × 10−4 cm/s and the downstream line at 6 × 10−5 cm/s. Lugeon and permeability values by grout zones and grout lines are shown in Table 2.

GROUTING: LESSONS LEARNED Grout takes into the limestone discontinuities resulted in a higher-than-anticipated volume of grout placed in the foundation. The original pre-construction estimated grout quantities were based on applying a 1.5 multiplier to the average grout quantities that were derived from several previously completed

Figure 17. Graphs of grout takes and pressure test results by grout lines.

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Figure 18. Graphs of grout takes by zone and number of downstages.

grouting projects in the region. Although these projects included bedrock materials similar in lithology and stratigraphic sequences to those involved in the Bolivar project, and although a relatively significant multiplier was adopted, quantities for grout components (cement, sand, bentonite, and admixtures) were underestimated and required modifications to the construction contract. A more accurate estimate of grout quantity requirements might have been derived by estimating the potential void space based on joint spacing, orientation, and aperture information from outcrops and borings at the site.

RISK-INFORMED CLOSURE CRITERIA Closure criteria for completion of the grout curtain were established to ensure that the design intent for Table 2. Lugeon and permeability values by grout zones and lines.

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the curtain was satisfied; that intent was to reduce the probability of dam failure due to erosion/scour of dam embankment by groundwater seepage through bedrock discontinuities. To accomplish this design intent, the grout curtain must significantly reduce groundwater flow/velocity to inhibit the erosion and transportation of dam embankment material. To reduce groundwater flow/velocity, the injected grout must significantly reduce the aperture or void area of the interconnected discontinuities and thereby reduce the bedrock permeability in the vicinity of the grout curtain. To evaluate grout curtain closure, a cadre of experienced grouting professionals was established to review grouting data and to determine if the design intent of the grout curtain was satisfied. The grout curtain closure criteria, established by this cadre, primarily included 1) review of data to assure that specified methods were successfully implemented for drilling, pressure testing, and grouting; 2) ensuring that a sequential reduction of grout take and pressure test values for higher-order grout holes was obtained; and 3) review of pressure test results of the highest-order grout holes to determine if a suitable reduction in bedrock permeability was actually obtained.

The results of reviewed field logs and test data suggested that specified methods for drilling, pressure testing, and grouting were successfully implemented, including satisfying individual hole refusal criteria (0.5 gallon/min) for each zone, meeting criteria for adding split-spaced higher-order grout holes (grout takes greater than 100 gallons or 378 L of the hole volume), and meeting grout quality requirements. Both grouting and pressure test data revealed that grout takes and pressure test results were reduced sequentially for higher-order grout holes, which is an indicator that the bedrock permeability was reduced as grouting progressed, as discussed earlier.

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Figure 19. Profile showing Lugeon values within Zone 2, along upstream line.

Risk-informed closure criteria were established to ensure efficient risk reduction. To develop these criteria, pressure test results were reviewed to determine if the bedrock permeability was reduced sufficiently in areas that posed the highest risks. It was determined that seepage through the Lower Mercer Limestone, located within grout Zone 2, posed a higher risk for dam failure relative to seepage through the Upper Mercer Limestone, which is located in grout Zone 1. This increased risk is due, in part, to increased frequency of pools that would provide sources for groundwater seepage through the Lower Mercer Limestone, as well as to longer pool durations and higher hydraulic gradients than those exposed to the Upper Mercer Limestone.

Therefore, more restrictive risk-informed criteria were established for Zone 2, which contains the Lower Mercer Limestone, which included the requirement to obtain an average pressure test value below 5 Lu for the

highest-order grout holes, with an upper bound value of 10 Lu for individual holes. See Figure 19 for Lugeon values within Zone 2 along the upstream line, and see Figure 20 for Zone 2 along the downstream line. Risk-informed criteria for the highest-order grout holes within Zone 1, which contains the Upper Mercer Limestone, included obtaining an average pressure test value below 20 Lu, with an upper bound of 35 Lu for individual holes. However, the risk-informed criteria for Zone 1 were applied to a limited length along the grout line from station −0 + 10 to 2 + 00, which is closest to the dam embankment. This limitation is due to outcrop exposures of the Upper Mercer Limestone downstream of the grout curtain, along the spillway sidewalls, which would allow groundwater flow to outflank the grout curtain during extreme flood events. See Figure 21 for Lugeon values within Zone 1 along the upstream line, and see Figure 22 for Zone 1 along the

Figure 20. Profile showing Lugeon values within Zone 2, along downstream line.

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Figure 21. Profile showing Lugeon values within Zone 1, along upstream line.

downstream line. As a result of reviewing pressure test and grouting data, an additional 22 grout holes were required to ensure that the grout curtain satisfied the risk-informed closure criteria.

CLOSURE CRITERIA: LESSONS LEARNED The contractor was instructed to add 22 grout holes during the final weeks of the project, which were required to meet risk-informed closure criteria. Riskinformed criteria were established on the review of pressure test data from the highest-order grout holes, which was not undertaken until the grout curtain was nearly completed. Impacts to the work schedule during the final weeks of the project could have been reduced with the different specified criteria for adding split-

spaced higher-order grout holes. The specified criteria for this project required additional holes to be drilled and grouted (split-spaced) if individual grout hole takes were 100 gallons (378 L) greater than the volume of the hole. Consideration could have been given to reduce this volume to 50 gallons (189 L) or even 20 gallons (76 L) of grout take greater than the volume of the hole. This change in specified criteria would have resulted in a foundation with lower permeability prior to review of the data.

SEEPAGE BARRIER—SOCKETED INTO BEDROCK The partial-depth seepage barrier wall was constructed primarily to address seepage within the glacial

Figure 22. Profile showing Lugeon values within Zone 1, along downstream line.

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Figure 23. Seepage barrier slurry loss and outbreak location.

outwash foundation within the valley bottom. However, a small portion of the seepage barrier wall is socketed into bedrock at the left abutment. The terminus of the seepage barrier wall overlaps with the grout curtain to provide a continuous barrier to groundwater seepage. This socketed portion of the barrier wall also provides a positive cutoff within the bedrock at the critical location near the top-of-rock surface, in the vicinity of the dam embankment. The seepage barrier wall was successfully constructed utilizing an overlapping panel method. Excavation for the wall was primary undertaken by a hydromill with a temporary bentonite slurry backfill to maintain sidewall integrity. The permanent panel backfill consisted of a high-slump, low-strength concrete mix, placed using the tremie method.

Open discontinuities were encountered while excavating bedrock at the seepage barrier wall panel P6467.

This panel is located in the small portion of the barrier wall that is socketed into bedrock along the dam’s left abutment. During excavation, rapid loss of a portion of the temporary bentonite slurry occurred. This slurry seeped through the interconnected bedrock fractures of the Lower Mercer Limestone, from approximate elevation 940–935 ft (286.5–285 m), and emerged at the ground surface approximately 200 ft (61 m) upstream of the wall. See Figure 23 for the location of the slurry loss at panel P6467 and the approximate outbreak locations. This panel was promptly filled with sand and was then successfully re-excavated and backfilled. However, the seepage of the bentonite slurry is an indicator of

potential seepage paths that could supply groundwater from the pool to bedrock fractures in the vicinity of the dam embankment. These critical seepage paths are now interrupted by the positive cutoff provided by the seepage barrier wall.

CONCLUSIONS The drilling and grouting contractor, TerraFirm Construction, LLC, mobilized suitable equipment and materials to perform the work. The contractor and USACE field personnel were experienced and conscientious professionals, which is key to any project success.

The drilling, pressure testing, and grouting methods satisfied specified tolerances and criteria. The majority of the grout volume was placed in downstaged intervals, where significant drill water loss was encountered, which was due, in part, to open discontinuities within two thin limestone units. Pressure test values and injected grout volumes were sequentially reduced with higher-order grout holes, indicating that the discontinuity apertures within the bedrock were filling as grouting progressed. Bedrock permeability in the vicinity of the grout curtain was reduced by 1 to 1.5 orders of magnitude. This reduced permeability should decrease the velocity of the groundwater seepage downstream of the grout curtain. This decreased velocity should also reduce its ability to erode or transport loose finegrained material, which will in turn reduce the risk of dam embankment failure. To date, the grout curtain

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has not been tested with flood-event pools above elevation 943 ft (287.4 m), which is comparable to those pools that produced downstream seepage through the Lower Mercer Limestone (such as flood events in years 1991, 2005, 2008, and 2011). During future high-pool events, the condition of downstream seepage will be noted, and piezometer readings will be compared to those of historic floods to further determine the extent of dam safety risk reduction.

DISCLAIMER The views, opinions, and findings contained in this article are those of the author and should not be construed as an official Department of the Army position, policy, or decision, unless so designated by other official documentation. In allowing publication of this article, the USACE does not endorse any entity or firm associated with this work.

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REFERENCES HOULSBY, A. J., 1976, Routine interpretation of the Lugeon watertest: Quarterly Journal Engineering Geology, Vol. 9, pp. 303– 313. ˜ QUINONES -ROZO, C., 2010, Lugeon test interpretation, revisited: In Collaborative Management of Integrated Watersheds, U.S. Society of Dams, 30th Annual Conference: S. 405–414.

TERRAFIRM CONSTRUCTION, October 2016, Bolivar Dam Major Rehabilitation Project—Seepage Barrier—Final Report.

U.S. ARMY CORPS OF ENGINEERS (USACE), Huntington District, May 2010, Major Rehabilitation Design Documentation Report, Bolivar Dam, Sandy Creek of the Tuscarawas River, Ohio.

USACE, Huntington District, February 2014, Construction Plans for Seepage Barrier, Bolivar Dam, Sandy Creek of Tuscarawas River, Bolivar, OH. USACE, Huntington District, February 2014, Specifications for Seepage Barrier, Bolivar Dam, Sandy Creek of Tuscarawas River, Bolivar, OH. USACE, July 2014, Grouting Technology, Engineer Manual EM 1110-2-3506.

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