SPECIAL EDITION OF THE DUTCH INDEPENDENT JOURNAL GEOTECHNIEK
16TH EUROPEAN CONFERENCE ON SOIL MECHANICS AND GEOTECHNICAL ENGINEERING
UK, EDINBURGH 13-17 SEPTEMBER 2015
What about polluted rivers and soils? What about polluted rivers and soils? What What about about polluted polluted rivers rivers and and soils? soils?
What about the rising sea level? What What about about the the rising rising sea sea level? level?
What about the rising sea level?
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What about increasing emissions? What about increasing emissions? What What about about increasing increasing emissions? emissions?
What about disappearing resources? What What about about disappearing disappearing resources? resources?
What about disappearing resources?
DEME has a leading position in a number of highly specialized and complex hydraulic disciplines. In the next decades, the DEME DEME has has aa leading leading position position in in aa number number of of highly highly specialized specialized world will be facing major challenges such as the effectsthe of and complex hydraulic disciplines. In the next decades, and complex disciplines. the next the DEME has a hydraulic leading position in aInnumber ofdecades, highly specialized climate change and scarcity of resources. Through innovative world will major challenges as effects of world will be be facing facing major disciplines. challenges such such as the the effects of the and complex In the next decades, thinking DEME and ishydraulic offering sustainable solutions ininnovative response climate change scarcity of resources. Through climate change and scarcity of resources. Through innovative world be facing challenges such as as soil the effects of to thesewill future needs major in various fields such and thinking thinking DEME DEME is is offering offering sustainable sustainable solutions solutions in in response response sediment remediation, water treatment, coastal protection, climate change and scarcity of resources. Through innovative to to these these future future needs needs in in various various fields fields such such as as soil soil and and development of green and bluetreatment, energy, offshore dredging of thinkingremediation, DEME is offering sustainable solutions in response sediment water coastal protection, sediment remediation, water treatment, coastal protection, gravel and sand, deep sea harvesting of minerals and creation to these future needs in various fields such as soil and development development of of green green and and blue blue energy, energy, offshore offshore dredging dredging of of of land in densely populated regions, ports and industries. gravel and deep of and sediment remediation, water treatment, protection, gravel and sand, sand, deep sea sea harvesting harvesting of minerals mineralscoastal and creation creation of land in densely populated regions, ports and industries. ofpopulated green andregions, blue energy, offshore dredging of ofdevelopment land in densely ports and industries.
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A coarse sand barrier as an effective piping measure N71 Voorwerk_Opmaak 1 28-08-13 12:10 Pagina 3
Van Beek, V.M. / Koelewijn, A.R. / Negrinelli, G. / Förster, U.
Interpretation of TA and DSS test on organic soft soil to derive strength parameters for dike design. Cofra BV
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A coarse sand barrier as an effective piping measure Figure 1 - The coarse sand barrier concept.
Van Beek, V.M. Deltares
Koelewijn, A.R. Deltares
Negrinelli, G. University of Brescia / Deltares
Fรถrster, U. Deltares
increase of the dike length to be reinforced. Traditional measures, such as a berm, are not attractive when large seepage lengths are required and often houses are situated closely behind the dikes. Sheet pile walls are an alternative, but are economically unfeasible when it comes to application for long dike stretches.
Introduction Backward erosion piping is a process whereby shallow pipes are formed at the interface of a cohesive layer and a sand layer, due to the removal of sand particles under the action of water flow. Ongoing pipe development can lead to severe erosion and finally failure of the water-retaining structure. The foundation that is susceptible to this mechanism, a combination of a uniform sandy layer covered by a cohesive layer is often encountered below river dikes in deltaic areas. Rise of the water level results in the formation of sand boils as a first sign of backward erosion. Numerous sand boils have been observed in the past, but failure due to backward erosion piping is not very common. Nevertheless, several dike failures in the Netherlands, China and the U.S. are attributed to this mechanism (Vrijling et al., 2010, Yao et al., 2009).
In the Netherlands, backward erosion piping is predicted with the Sellmeijer model (Sellmeijer, 1988, Sellmeijer et al. 2011, TAW, 1999). This model predicts the critical head on the basis of the groundwater flow towards the pipe, the viscous flow through the pipe and the limit-state equilibrium of particles at the pipe bottom. The model has been validated using experiments, but application in the field proves to be complex, as the required parameters are difficult to determine and show large fluctuation in the field. The uncertainty with respect to input parameters leads to the selection of conservative estimates, such that considerable reinforcements are due. The more stringent safety standards, due to a recent validation of the model, the inclusion of the length-effect and the risk approach, recently embraced in the Netherlands, lead to a further
GEOTECHNIEK - September 2015
Innovative or alternative piping measures are therefore becoming more and more popular. An example of an innovative measure is the vertical sand-retaining geotextile (Bezuijen et al., 2013, Fรถrster et al., 2015). Using this method nearby the toe of the dike a vertical geotextile is inserted into a trench. Above the sand layer the trench is refilled with clay, such that upward seepage is not possible. An optimisation of this innovative solution is proposed here: a coarse sand barrier. In this solution the pipe formation is resisted by coarse sand instead of by the geotextile. The concept The coarse sand barrier relies on the concept that coarse sand provides more resistance to pipe formation than fine sand. The coarse sand is brought into the subsurface by creating a trench and simultaneous filling with coarse sand. Once a pipe forms, it will develop along the top of the sand bed and will collide with the barrier. As the coarse sand provides more resistance to erosion and the water flow is controlled by the overall properties of the aquifer, the water forces acting on the coarse particles are not sufficient for pipe development, unless the head across the structure is raised.
Recent safety studies indicate a significant risk of backward erosion piping as a failure mechanism for dikes. Traditional measures, such as berms or sheet pile walls take a lot of space or are expensive, in particular in areas with a high infrastructural density. Consequently, there is an urgent need for alternative piping measures. An innovative solution is the vertical sand-retaining geotextile, which has been developed and tested in the
field. An optimisation of this method is a coarse sand barrier: a trench filled with coarse sand, covered by a clay layer. The coarse sand provides more resistance to erosion than the inherent sand of the aquifer with its uncertain composition with respect to d70, so that initiated pipes cannot continue to develop. Laboratory and field scale tests suggest that this is an effective and economically feasible method.
Figure 2 - Laboratory set up, showing the small-scale model with circular exit. Figure 3 - Slope type exit viewed from above (above) and circular exit covered by a sand boil (below).
The permeable barrier will deflect excessive vertical seepage below the pipe tip, such that fluidisation of the sand bed below the pipe is less likely than for an impermeable structure like a sheet pile wall. As in the method with the geotextile, above the sand the trench is filled with clay, to prevent upward seepage. Due to the clay filling above the barrier, the method is different from a more common filter, which aims for controlled discharge of water. For the coarse sand barrier the discharge is not expected to increase, not more than would be the case for any other barrier-type solution. Clogging is not expected, as long as the barrier is continuously below the water level. Laboratory evidence Laboratory experiments have been performed to illustrate the functioning and potential of a coarse sand barrier. A small-scale box (described in more detail in Van Beek et al., 2011 and Van Beek et al., 2015) with transparent cover
(simulating the dike) was used for the experiments. Two configurations were used, one with an open exit representing a 2D exit with unconstrained flow towards the surface (also described in Van Beek et al., 2008) and one with a circular exit in the cover, representing a 3D exit with concentrated flow towards a single point (Figure 3). In both configurations the box was filled with fine sand with a band of (medium) coarse sand and a head difference was applied to the sand until pipe formation occurred. The sand types used in the slope-type configuration were Playground sand and Masonry sand (with a d50 of 0.191 and 0.454 mm respectively) and the sand types used in the hole-type configuration were Baskarp sand and Itterbeck fraction 431 Îźm sand (with a d50 of 0.132 and 0.342 mm respectively). All sand types are uniform (d60/d10 1.5 â€“ 2.6). The observed process was similar in both con-
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figurations: the head drop was increased until a pipe formed. This pipe developed up to the coarse sand and started to develop parallel to the direction of flow and adjacent to the coarse sand barrier. Upon significant increase of head the pipe passes through the coarse sand barrier and develops towards the upstream side. The critical hydraulic heads obtained in the experiments are indicated in Figure 4 for the samples with a coarse sand barrier and their homogeneous equivalents. It is noted that the upstream filter resistance causes quite some head loss in the slope-type configuration, such that the actual head loss across the sand bed is
smaller. Nevertheless, the difference between the critical heads in the experiments with and without sand barrier is very large, illustrating the potential of the method. Field evidence As the critical head of piping is affected by scale, it is relevant to also test the method at larger scale. In 2009 and 2012 large-scale experiments were performed at the location of the IJkdijk. In these tests an actual dike was built on top of a sand bed placed in a basin. The test dike was 3.5 m high, 15 m long and 15 m wide at its base. It was constructed of compacted clay. The base consisted of a 3 m thick sand layer which extended 15 m beyond the test dike both at the upstream side and at the downstream side. Homogeneous tests without measures were performed in 2009. One of the purposes of the experiments in 2012 was to test piping measures, of which one of them was a coarse sand barrier. This coarse sand barrier was applied as an obstructing bar underneath the dike, 0.5 m wide and 0.5 m deep, at about one quarter of the seepage length from the downstream toe, see Figure 5 and 6. The coarse sand filter has been applied in the same basin where the first IJkdijk piping test had been carried out (Van Beek et al., 2011). In the upper 0.5 m a new, comparable sand was placed, with d50 =0.180 mm, Cu =1.7, d70 =0.207 mm. For the selection of a suitable sand for the filter, three criteria should be met: the filter should have a sufficient permeability, it should retain the finer material of the test sand and it
should be internally stable. Based on filter criteria by Terzaghi and given by Giroud (2010) and Burenkova (1993) the coarse sand was selected, with d50=1.331 mm, Cu=1.3, d70=1.52 mm, d10=1.054 mm, d15=1.085 mm, d90=1.79 mm. The first sand boil appeared at a head of 1.60 m. The head was increased until a level of 3.49 m, which equalled the height of the levee and was therefore the maximum head that could be applied in this experiment. In the homogeneous equivalent, performed in 2009, the dike failed at a hydraulic head difference of 2.1-2.3 m. These experiments indicate that at this scale the dike with the coarse sand barrier can at least with-
stand a head that is 1.6 times that of a dike without the barrier. Discussion and Conclusions The laboratory and field experiments illustrate that the application of a coarse sand barrier as a piping measure is promising. In the small-scale experiments an increase by a factor of 3-4 was established. An increase in strength of at least 1.6 is obtained in the field experiments at which failure due to piping did not even occur. Design rules should be based on filter criteria, heave criteria and the horizontal resistance against piping, which still requires investigation. Due to the experience with the vertical geotextile as a
Figure 4 - Critical gradients obtained in the experiments.
Figure 5 - IJkdijk test schematisation showing the location of the coarse sand barrier.
GEOTECHNIEK - September 2015
A COARSE SAND BARRIER AS AN EFFECTIVE PIPING MEASURE
Figure 6 - Digging the trench for the barrier (left) and the coarse sand barrier (right). At the background the dike that is to be placed is drawn at the side of the basin.
piping measure, which is in many ways similar, practical issues for application in the field are likely to be resolvable. References - Bezuijen, A., Van Beek, V., Förster, U., (2013). Geotextiel als pipingremmend scherm, hoe werkt het? Geotechniek 18(1): 38-41, katern Geokunst. - Burenkova, V.V. (1993). Assessment of suffusion in non-cohesive and graded soils, Filters in geotechnical and hydraulic engineering, Brauns, Heibaum& Schuler, Balkema, Rotterdam, 357-360. - Förster, U., Bezuijen, A, Van den Berg, S. G., (2015), : Vertically inserted geotextile used for strengthening levees against internal erosion Category: B3 Earthworks, Dams and Dykes - Giroud, J. P. (2010). Development of criteria for geotextile and granular filters, Proceedings 9th International conference on Geosynthetics, Guaruja, Brazil, 20 pp.
- Sellmeijer, J.B. (1988). On the mechanism of piping under impervious structures. Doctoral dissertation, TU Delft, The Netherlands. - Sellmeijer J.B., Lopéz de la Cruz J., Van Beek V.M., Knoeff J.G. (2011). Fine-tuning of the piping model through small-scale, mediumscale and IJkdijk experiments. European Journal of Environmental and Civil Engineering 15(8): 1139-1154. - TAW (Technische Adviescommissie voor de Waterkeringen) (1999). Technisch rapport Zandmeevoerende wellen. Technische Adviescommissie voor de Waterkeringen, Delft, The Netherlands. - Van Beek, V.M., Koelewijn, A., Kruse, G., Sellmeijer, H., Barends, F. (2008). Piping phenomena in heterogeneous sands – experiments and simulations, Proceedings of the 4th International Conference on Scour and Erosion, p. 453-459, http://scour-and-erosion.baw.de/ conferences/icse4/. - Van Beek, V.M., Knoeff, J.G., Sellmeijer, J.B.
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(2011). Observations on the process of backward piping by underseepage in cohesionless soils in small-, medium- and full-scale experiments. European Journal of Environmental and Civil Engineering 15(8): 1115-1137. - Van Beek, V.M., Van Essen, H.M., Vandenboer, K., Bezuijen, A. (2015). Developments in modelling of backward erosion piping. Géotechnique, to be published. - Vrijling, J.K., Kok, M., Calle, E.O.F., Epema, W.G., Van der Meer, M.T., Van den Berg, P., Schweckendiek, T. (2010). Piping - Realiteit of Rekenfout? Technical report, Dutch Expertise Network on Flood Protection (ENW). - Yao, Q., Xie, J., Sun, D., Zhao, J. (2009). Data collection of dike breach cases of China. Sino-Dutch Cooperation Project Report. China Institue of Water Resources and Hydropower Research.
Interpretation of TA and DSS test on organic soft soil to derive strength parameters for dike design. Figure 1 - Test locations in the topview of the dike.
H. J. (Arny) Lengkeek Witteveen + Bos, Deventer, The Netherlands
M. (Matteo) Brunetti Witteveen + Bos, Deventer, The Netherlands
1. Introduction For a study into undrained stability calculations of flood defences on behalf of STOWA and waterboard of Delfland, Witteveen+Bos investigated the differences between drained and undrained stability calculations for a local flood defence. The reference site is a local flood defence along the Berkelsche Zweth channel located between Rotterdam and Delft in the west of the Netherlands. The Berkelsche Zweth channel functions as a discharge for the water pumped out of the surrounding polders. The studied section is the flood defence between the channel (with a water level of -0.4 m NAP) and the low lying polder Schieveen (level approximately -5.0 m NAP).
Figure 2- Soil profile, water level in channel and polder level.
The analyzed flood defence has been previously studied in 1972-1973 in the context of the systematic local flood defence study (“systematisch kade-onderzoek COW”). Based on those results it has been improved in the 1980s. In the design of that improvement, particular attention was paid to limiting the “driving” force (e.g. limiting the toe ditch) and limiting the surface subsidence. In 2006 another improvement of the flood defence has resulted in its current stable state. From archives it is known that in 1806 a slope instability occurred at about 600m from the pilot location. Further collapse have been prevented by strengthening the dike with a berm. 2. Site investigation program and soil profile An extensive geotechnical investigation program has been carried out to gain reliable data on soil profile and parameters. In a single cross-section, eight CPT’s and three boreholes with continuous sampling have been carried out (see Figure 1). The CPT’s were made to a level of -19 m NAP, while the boreholes reached a depth of 5 m in the crest of the dike and 10 m in the berm and in the polder. After the field work the following
GEOTECHNIEK - September 2015
This paper presents the findings on the behaviour of organic soft soil (clay and peat) using a variety of state-of-the-art laboratory tests. Moreover it addresses the importance of critical evaluation of the each test and the parameter determination. The tests are performed within a research program on behalf of research foundation “STOWA” to verify the stability of a local flood defence of the waterboard “Hoogheemraadschap van Delfland”. Oedometric tests (OED) and constant rate of strain tests (CRS) have been carried out to determine the preconsolidation stress, after which anisotropically K0-consolidated undrained triaxial tests (TA-ACU) and direct simple shear tests (DSS) have been executed with three different consolidation paths. All tests have been performed on undisturbed samples to obtain reliable soil parameters. The interpretation of laboratory tests is made in terms of pre-consolidation stress (s’vy), undrained shear strength (Su) and effective friction angle (φ’). This paper focuses on determination of the strength parameters at large strains, the so called critical state or ultimate state parameters, further on called ultimate state parameters (“ult”). The ultimate parameters can be used for a effective stress approach with Mohr-Coulomb strength model using effective friction angle and no cohesion. The ultimate parameters can also be used for a critical state soil mechanics (CSSM) approach with undrained shear strength related to the effective stress and overconsolidationratio (OCR).
The undrained shear strength (Su) is normalised by the vertical consolidation stress (s’vc) to derive the undrained shear strength ratio (S). By plotting S versus the OCR, both the normally consolidated value (S;nc) as the exponent (m) to describe the overconsolidated behaviour are derived. The OCR is defined as s’vy divided by s’vc. This paper presents additional criteria to define the ultimate state of Su. To obtain the ultimate state effective friction angle (φ’), the interpretation of TA-ACU is made by plotting t against s’. For the DSS tests, Su is plotted against the ultimate vertical effective stress (σ’ult), which represents the vertical effective stress acting on the sample at the moment of failure. This paper presents additional criteria to define the ultimate state of φ’. Lastly a comparison is made between two consolidation procedures to derive the undrained shear strength at in-situ stress conditions. Two identical samples have been tested, one sample is consolidated to the in-situ stress level before undrained shearing. The other samples is first consolidated to the preconsolidation stress, then consolidated to the in-situ stress before undrained shearing, the so called SHANSEP method.
Tabel 1 - Index properties of soil layers description
B002 Holland Peat B003
Tiel Clay B002
ρ (kN/ m3)
* γn is the unit of weight of the material at the natural saturation degree found in the field
GEOTECHNIEK - September 2015
laboratory tests have been performed: • Classification tests (24x) • Conventional Oedometric tests (10xOED) • Constant Rate of Strain tests (11xCRS) • Anisotropically K0-Consolidated Undrained triaxial tests (18xTA-ACU) • Direct Simple Shear tests (18xDSS) The subsoil conditions on the site are characterized by layering of natural and manmade cohesive deposits. The top layers in the dike body consist of anthropogenic clay. Three anthropogenic layers can be distinguished, which are named Clay Dike, Clay #4 and Clay 1972. The last two layers are not included in this study since it was shown that the failure circles were only just intersecting these layers and that their influence on the overall stability is therefore negligible. Below the manmade soil two natural Holocene deposits are encountered, the Holland peat underlain by the Tiel clay. Below the compressible and organic Holocene deposits a thick layer of sand is present. An overview of the index properties of the studied layers is given in Figure 2. In the table it can be observed that within the geological deposits the density (ρ) and the water content (w) vary widely. There is a clear correlation in the variability, lower density correlates to higher water contents (and higher organic content). Even though the variability within each of the layers is significant, there is still a very clear separation between the different layers.
Figure 3 - Testing methodology (test types and resulting data).
Figure 4 - In-situ effective stress and (pre-)consolidation stresses.
3. Methodology of geotechnical testing In general the methodology of the tests has been to determine the state of the soil by using CRS tests after which both the ultimate state strength parameters of the different layers could be determined by using TA and DSS tests.
3.1. CRS testing Eleven CRS tests were performed to gain reliable pre-consolidation stresses. The resulting compression parameters were not evaluated in this studies since no settlement calculations were required. Traditionally, compression
parameters and pre-consolidation stresses are determined using oedometer tests (OED). In oedometer tests, each load step is placed instantaneously. In order to improve the resolution of the data obtained, constant rate of strain tests are performed. Using this combination of tests, more reliable pre-consolidation stress and compressibility parameters are obtained. From the CRS test compression parameters and pre-consolidation stresses were obtained. The best estimate pre-consolidation stress (s’vy;A) is derived from normal Casagrande procedures. The upper bound pre-consolidation stress (s’vy;B) obtained where the compression curve starts to be linear at higher stresses showing true normally con-
GEOTECHNIEK - September 2015
solidated behaviour. This value is typically twice the value of s’vy;A. Using the results of the CRS tests and the calculated in-situ effective stress, the consolidation stresses for the laboratory testing have been determined. The in-situ stresses including the pre-consolidation stresses determined by the CRS tests have been graphically presented in Figure 4. In this figure also the consolidation stresses used for TA and DSS tests have been indicated. The initial specimen height was about 20 mm with a diameter of 66 mm, the rate of strain during the tests varied between 0.2 and 0.5 % / hour. The specimen is placed in a stainless steel ring resting on a ceramic porous stone. The specimen is loaded by means of a piston with a gear driven load frame in order to guarantee the constant rate of strain. During the test the excess pore pressure and chamber pressure are measured by means of transducers. The imposed displacement and the vertical load are also measured. Within the project the following loading schedule was followed: - Start at a low stress up to 4 kPa - Increase the load to a maximum of about 300 kPa (up to 10 times the in situ effective stress) - Series of unload, reload and relaxation. 3.2. Triaxial test Anisotropically consolidated undrained triaxial tests (ACU-TA) have been performed to determine the ultimate state strength parameters for drained and undrained calculations. A number of depths were selected for testing. For each selected depth three different consolidation procedures (levels) were used on three separate samples: • Test-I: Consolidation to the theoretical vertical effective in-situ stress. • Test-II: Consolidation to the best estimate preconsolidation stress (s’vy;A) obtained from CRS and then brought back to the theoretical vertical in-situ stress (SHANSEP). • Test-III: Consolidation to at least the upper bound pre-consolidation stress (s’vy;B) obtained from CRS in order to ensure it behaves as normally consolidated. Six samples were tested from the Dike clay and twelve from the Tiel clay. Using the in-situ stress, pre-consolidation stress and the test methodologies as described before, the resulting consolidation stresses for the different samples are shown in Table 2. No TA were performed on the layer of Holland peat, since TA on peat do not give an accurate
INTERPRETATION OF TA AND DSS TEST ON ORGANIC SOFT SOIL TO DERIVE STRENGTH PARAMETERS FOR DIKE DESIGN.
Tabel 2 - TA-ACU consolidation pressures used soil type
3 Clay Dike
20 Tiel clay 5 B003
results of the strength parameters. The initial specimen height was about 66 mm with a diameter of 33 mm, the rate of strain during the tests varied between 0.2 and 0.5 % / hour. The tests were performed to 15% axial strain (equals 22% shear strain). In line with the research program all tests were K0-consolidated, where K0 has been taken as 0.5. An OCR dependent K0 has not been applied. 3.3. DSS test DSS with constant height (constant volume) during shearing have been performed to determine the ultimate state strength parameters for drained and undrained calculations. Three samples were tested of the Dike clay, Nine of the Holland peat and six of the Tiel clay. The direct simple shear tests (DSS) are consolidated in the same way as the TA-ACU, with one slight chan-
sample level (m NAP)
38 → 20
76 → 32
38 → 16
44 → 22
22 → 11
84 → 42
42 → 21
36 → 20
18 → 10
46 → 20
18 → 10
ge. In test-III the samples are first consolidated to 10% higher stress level before consolidating to s’vy;B. This way the excess pore pressures before shearing are minimized. Furthermore the tendency to develop large creep strains compensated by vertical stress relief (due to constant height test condition) is reduced. The consolidation pressures as shown in Table 3 have been used. The DSS on de Dike clay showed some unrealistic behaviour and will not be discussed in this paper. The initial height of the sample was about 27.5 mm with a diameter of 65.9 mm. The rate of strain during the tests was between 1.2 and 1.8 mm/hour. The tests were performed to 40% shear strain. 3.4. DSS simulation and interpretation During the DSS test only the vertical effective
GEOTECHNIEK - September 2015
stress and the shear stress can be measured. This leads to the situation where the full stress condition of the sample is unknown (the directions of the principal stresses cannot be determined). In order to be able to determine a reasonable value for the critical state parameters, some assumptions on the behaviour of the material have to be made. When back-analyzing the DSS test in Plaxis, it can be shown that the principal stresses are under an angle of 45o with the vertical (Ref. ). This leads to the situation in which σ’v = σ’h = s’, and Su = τ = t. When plotting Mohr-circles using this assumption, it is clear that the mobilized friction angle can be found by using sin φ’ = τ / σv’ instead of using tan φ’ = τ / σv’. This assumption is not necessarily conservative, but especially when Plaxis is also used in the design for a situation where the material is loaded in shear, it is considered a valid and consistent approach since the back-analysis has already shown that Plaxis will in this situation represent the actual behaviour of the material correctly. 4. Results 4.1. Stress strain curves The stress strain curves of Holland peat and Tiel clay are presented in are presented in Figure 5 and 6. The graph of Dike clay is not added as it shows similar behaviour. The stress is normalised by the vertical consolidation stress. Various observations are made: • Test III with the normally consolidated samples results in the lowest stress ratio (S). Test I and II result in higher S-values due to the overconsolidation. This is in line with the critical state soil mechanics (CSSM) consideration. • Large strains are generally required to reach ultimate state, most samples show ongoing hardening. • In particular the DSS on Holland peat (Figure 6) show large strains to reach ultimate state, even for test III. • In particular the TA test III on Tiel clay (Figure 5) show small strains to reach peak strength and some softening at ultimate state. In general it can be concluded that the tests results are considered to be normal and in line with experience with organic soft soils. 4.2. Stress path Some stress path’s require careful evaluation. Two examples are presented in Figure 7 and 8. The tests are selected to address the importance of critical evaluation of the strength parameter. Normally the ultimate state is defined as the large strain value. Here this leads to unrealistic
high values for Su, in particular for overconsolidated samples (Test I and II). These samples tend dilate strongly with increasing strain.
Therefore two additional criteria are defined: • The Su;ult value is defined as the maximum secant value in the stress path graph (green line
Tabel 3 - Consolidation pressures for DSS tests soil type
s’vc;1 → s ‘vc;2 (kPa)
level (m NAP)
19.3 → 10
84.3 → 80
35.2 → 19
91.5 → 86
31.7 → 18
118.5 → 112
26.5 → 10
93 → 88
66.9 → 43
209 → 196
37.3 → 20
108.6 → 102
Figure 5 - Stress strain curves TA-ACU on Tiel clay.
in Figure 7 and 8). This is the maximum t/s’ ratio in TA and the maximum τ/s’v in DSS. Beyond this value the sample tend to dilate strongly where the normal stress increase significant more than the shear stress. • The Su;ult value is defined as the first intersection with the tensile cut-off line (Figure not added). The tensile cut-off is the state in which the σ’3 is lower than 0, i.e. that tensile stresses start to develop. In theory soil does not provide any tensile resistance , this implies that the minimum allowable effective stress (σ’3) can never be negative. Assuming that σ’3 = 0 it geometrically implies that the left limit is identified by a 1:1 line in the s’ - t plane. For DSS it is assumed the same 1:1 line can be plotted in the τ / σ’v plane. The minimum of the ultimate state and the two additional criteria is now defined as the new condition and compared to the ultimate state only. The new condition is briefly indicated with “min” compared to “ult” for the ultimate state only. 4.3. CSSM S-OCR curves The undrained shear strength can be evaluated is in line with the critical state soil mechanics (CSSM) approach, where the undrained shear strength ratio (S) is plotted versus the overconsolidationratio (OCR) and equations 1 and 2 apply to derive S and m Su;nc = S∙s’v (eq. 1) Su;oc = Su;nc∙OCRm (eq. 2) The results of both conditions are presented in Figure 9 and 10. Various observations are made: • The S-value of dike clay is considered to be very high, in particular for a low organic material. This is not fully understood. A few aspects might be of influence. The layer is manmade Figure 6 - Stress strain curves DSS on Holland peat.
TA-ACU Tiel clay
DSS Holland peat
B3M5 test I
B2M12 test I
B2M20 test I
B3M5 test II B3M8 test II
B2M12 test II B2M20 test II
S = tau / σ'vc (-)
S = t / σ'vc (-)
B3M8 test I
B3M3 test I B2M9 test I
B2M7 test I B3M3 test II
B2M9 test II B3M5 test III
B2M7 test II
B3M8 test III B2M12 test III
B3M3 test III
B2M9 test III
B2M20 test III
B2M7 test III
10 Axial strain (%)
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20 25 Shear strain (%)
INTERPRETATION OF TA AND DSS TEST ON ORGANIC SOFT SOIL TO DERIVE STRENGTH PARAMETERS FOR DIKE DESIGN.
Figure 7 - Stress path example TA-ACU on Dike clay (B001 M3 test-I).
Figure 8 - Stress path example DSS on Holland peat (B003 M3 test-I).
Figure 9 - S-OCR curves from TA.
Figure 10 - S-OCR curves from DSS.
and is mainly unsaturated under in-situ conditions. • The S-value of Tiel clay is 0.33 for TA conditions and 0.27 for DSS conditions. The DSS value is about 80% of the TA value which is in line with reported experiences of other clays. • The new condition yields generally in a better regression coefficient and less extreme values. In particular the unrealistic high values are taken out. This causes in general a lower exponent (m). The intersection at the OCR=1 axis is not significantly influenced and is in all cases similar the average of only test III (values between brackets in the legend). • The determination of the exponent (m) depends very much on a few outliers. It is concluded that these should be carefully examined
and the new conditions is an improvement. It is furthermore required to have sufficient tests and also larger ranges of overconsolidation ratio. 4.4. SHANSEP The effect of SHANSEP can be investigated by comparing test I and II. Test-I is consolidated to the theoretical vertical effective in-situ stress. Test-II is consolidated to the pre-consolidation stress and then brought back to the theoretical vertical in-situ stress (SHANSEP procedure). Various observations are made from Figure 11: • The Su value of test II is higher than test I for all three soil types and both for TA and DSS tests. Most point are above the 1:1 line and below the 2:1 line.
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• Figure 11 is based on the new condition. It should be noted that the scatter and ratio for the ultimate state condition would be larger. • Test II yields in a higher exponent (m) in the CSSM model. • Applying the SHANSEP method typically provides 5 kPa higher Su values, say 20 kPa (Test II) instead of 15 kPa (Test I) at in situ stress conditions. This increase can be significant for stability calculations. The reason of this difference should be further investigated. It might be that the actual OCR was overestimated. Another reason might be the OCR is a result from aging, and not from physical pre-loading. It should be further investigated if the effect of aging on the Su (or on the exponent
m) is less prone than for physical pre-loading. 4.5. Effective friction angle The effective friction angle can be derived from Figure 11 - Comparison Su results test I and II.
the failure points at the ultimate of new condition. In line with the CSSM approach the cohesion intercept is set to zero and equations 3 and 4 apply. sin ϕ’ = t’ult/s’ult sin ϕ’ = τult/s’v;ult
The TA results are plotted in a t-s’ diagram (see Figure 12). The DSS results are plotted in a τ-s’v diagram (see Figure 13). It should be noted that for the DSS the ultimate vertical stress (at failure instead of consolidation) is used. The following observations are made: • The ultimate condition and the new condition yield in almost identical results. Apparently the stress path moves along the failure line. • The best fit lines show a small cohesion intercept, likely to be caused by the overconsolidated test I and II. • The TA show very little scatter and the effective friction angle (φ’) can be derived accurately.
• The DSS show more scatter, also for the Tiel clay at normally consolidated stress level (Test III). • The derived ϕ’ for each layer is presented in the legend between brackets. The friction angle is derived by averaging each individual test III result per layer. • The ϕ’ derived from TA on Dike clay is considered to be high, as also concluded for the Svalue. • The ϕ’ derived from TA on Tiel clay is considered to be high, but more often seen for clays with organic content. • The ϕ’ derived from DSS on Tiel clay yields to the same value as that from TA. • The ϕ’ derived from DSS on Holland peat is high, as expected. 4.6. Resulting parameters The results of the tests show high friction angles for both TA and DSS testing. In numerous previous researches high friction angles for or-
Figure 12 - TA t-s’ diagram.
Figure 13 - DSS τ-s’v diagram.
Tabel 4 - Overview of strength parameters Condition
Minimum of requirements
t’/s’ or τ/s’v (-)
t’/s’ or τ/s’v (-)
With: S = Su ratio normalised by the consolidation stress m = OCR exponent t’/s’ = Stress ratio at failure of Test III samples in TA τ/s’v= Stress ratio at failure of Test III samples in DSS ϕ’ = Effective friction angle.
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INTERPRETATION OF TA AND DSS TEST ON ORGANIC SOFT SOIL TO DERIVE STRENGTH PARAMETERS FOR DIKE DESIGN.
ganic soils have been found in triaxial testing. This is especially the case for fibrous soils (peat) [8, 9], the fiber reinforcing effect has therefore been a main explanation for the high friction angles . However, since similar behaviour is found in organic clays, which do not contain high amounts of fibres, it has to be concluded that at this moment the cause of this behaviour is not fully understood. For direct simple shear testing high friction angles have also been reported earlier (for example [5, 11]), and recent studies on Dutch peats have shown that in the τ/σv’ plane in most cases the 1:1 line is approached [6, 7]. This leads to high friction angles for the DSS test as well, similar to the findings in this study. The undrained shear strengths found from DSS tests is approximately 80 of the undrained shear strengths from TA-ACU. The resulting internal friction angles from the DSS tests are higher than the values found from TA-ACU. Both results obtained are in accordance with literature [1; 2; 3] The resulting parameters are presented in Table 4. It is concluded that the minimum of requirements is an improvement for the determination of the undrained shear strength parameters (S, m). For the effective friction angle there is no difference. 5. Conclusions Several laboratory tests have been performed to gain reliable soil profile and parameters in order to check the stability of an existing flood defence. Table 4 provides an overview of the strength parameters obtained from the laboratory tests and used in performing stability calculations. Based on the analyzed case study it is concluded that: For the TA-ACU tests: • TA-ACU tests have been performed only in the layers of clay because they are considered not reliable when performed on peat. • The resulting Su ratio of Dike clay is about 0.50 with a power of 0.6-0.8. • The resulting Su ratio of Tiel clay is about 0.33 with a power of 0.6-0.9. • The resulting internal friction angles of Dike clay is about 37°. • The resulting internal friction angles of Tiel clay is about 37°. For the DSS tests:
• DSS tests interpreted in terms of undrained parameters provide comparable results with triaxial tests. • The resulting Su ratio of Tiel clay is about 0.27 with a power of 1.0. • The resulting Su ratio of Holland peat is about 0.50 with a power of 0.6. - The resulting internal friction angles of Tiel clay is about 37°. - The resulting internal friction angles of Holland peat is about 46°. DSS tests interpreted in term of undrained Su ratio provide lower values when compared with results from triaxial test, while DSS tests interpretation in term of drained strength parameters provide similar results. These results are in accordance with literature [1; 2; 3]. In general strength parameters evidenced in this case-study are significantly higher when compared with usual and expected cohesive soil strength parameters. A good explanation of this behaviour can be related to the high organic and water content. In organic material the fibres present within the soil are able to provide high tensile strength, thus increasing the soil shear strength. This paper presents two additional criteria to define the ultimate state, indicated with “min” compared to “ult” for the ultimate state only. It is concluded that a careful evaluation of the tests and the additional requirement are an improvement for the Su parameter determination. It is furthermore advised to have sufficient tests and also larger ranges of overconsolidation ratio. The determination of the effective friction angle is not effected or improved. The comparison between two consolidation procedures showed that the so called SHANSEP method yields in higher Su values, which can be significant for stability calculations. It is recommended to investigate this into more detail. Finally authors would like thank Jan Tigchelaar of Hoogheemraadschap van Delfland and Henk van Hemert of STOWA for their cooperation in this research project. References  Tigchelaar, J., De Feijter, J.W., Den Haan, E.J. “Shear tests on reconstituted Oostvaardersplassen clay”. Soft Ground Technology: pp. 67-81, 2001.  Hansen, L.A., and Clough, G.W. “Characterization of the Undrained Anisotropy of
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Clays”, Application of Plasticity and Generalized Stress-Strain in Geotechnical Engineering, Ed. by R.N. Yong and E.T. Selig, ASCE, New York, pp.253-276, 1980.  Jamiolkowski, M., Laad, C.C., Germaine, J.T., Lancellotta, R. “New Developments in field and laboratory testing of soils”. Proceedings of the 11th International conference on Soil Mechanics and Foundation Engineering, San Francisco, Aug. 12-16, 1985 Vol. A.A. Balkema, Boston, 1985.  Farrell, E., Jonker, S., Knibbeler, A., and Brinkgreve, R. “The use of direct simple shear test for the design of a motorway on peat”. In 12th European Conference on Soil Mechanics and Geotechnical Engineering, Rotterdam. A.A. Balkema. 1999.  Yamaguchi, H., Yamaguchi, K., and Kawano, K., “Simple shear properties of peat.” Proc., Int. Symp. on Geotechnical Engineering of Soft Soils, M. J. Mendoza and L. Montañez, eds., Sociedad Mexicana de Mecánica de Suelos, Coyoacán, Mexico, 1, 163–170, 1987.  Den Haan, E.J. “Ongedraineerde sterkte van slappe Nederlandse grond, Deel 2.” Geotechniek, 1:42–51. www.vakbladgeotechniek, 2011.  Den Haan, E.J. “Modelling peat with an anisotropic time-dependent model for clay”, Numerical Methods in Geotechnical Engineering – Hicks, Brinkgreve & Rohe (Eds), Taylor & Francis Group, London, 978-1138-00146-6, 2014  Landva, A.O., “Characterization of Escuminac peat and construction on peatland”, Characterisation and Engineering Properties of Natural Soils – Tan, Phoon, Hight & Leroueil (eds), Taylor & Francis Group, London, ISBN 978-0-415-42691-6, 2007  Mesri, G., Ajlouni, A.M., “Engineering Properties of Fibrous Peats”, Journal of Geotechnical and Geoenvironmental Engineering, Vol. 133, No. 7, 1090-0241/2007/7850–866, 2007.  Cola, S., Cortellazzo, G.,”The Shear Strength Behaviour of Two Peaty Soils”, Geotechnical and Geological Engineering, 23: 679-695, 10.1007/s10706-004-9223-9, 2005.  Lengkeek, H., Bouw, R., “Triaxial, DSS, CRS tests and numerical simulations of soft soils at river dike”, Proceedings of the 15th European Conference on Soil Mechanics and Geotechnical Engineering, Anagnostopoulos, A., Pachakis, M., Tsatsanifos, C.(eds), IOS Press, 10.3233/978-1-60750801-4-427, 2011
Working Through Water
Niall Corney Huesker Synthetic GMBH (area manager)
This landmark highway project in Turkey had many demanding site constraints. This led to innovative construction techniques being adopted by the Client, Designer and Contractor.
The detailed design called for Huesker Ringtrac® GEC sleeves of 80cm diameter with a spacing density between 17 to 20%. Additionally three layers of high strength polyester Stabilenka® reinforcement were laid above the heads of the GECs as a basal platform structure. Over 10,000 GECs were installed and in excess of 80,000m2 of Stabilenka® reinforcement. The key embankment statistics are as follows: Maximum height: 18m Base width: 90m Length: 380m Lake water depth: up to 7m Sediment depths: up to 20m
The Ringtrac sleeve for the GECs was delivered on a roll.
This embankment construction involved GECs being installed from land and advancing outwards with the subsequent sequential installation of the base reinforcement and finally the embankment fill in a progressive manner across the reservoir. A 300T crane with a boom capable of reaching some 75m was utilized to install the GEC piles and the high strength base reinforcement geotextile, all underwater.
The funnel was filled with the tube This highway project required the construction of 2km of new build road, 5km of existing road upgrading to highway standard and a 0.8km long embankment to carry the highway across an existing reservoir for approximately 380m. The completion of this section of highway had been politically complex since the proposed road embankment location corresponded with the intersection of three of the Federal Highways Authority regions. In addition to this It was considered that there was no economically feasible way to cross the
400m of open water as the recent sediments accumulated in this part of the reservoir were up to 20m in depth and were classified as ‘very’ to ‘extremely soft’. These sediments were located under approximately 7m of water which in itself added to the complexity of the construction task. The project was tendered in June 2011 and subsequently awarded in September 2011 to main contractor STY Insaat with ATLASYOL Inc. appointed as the specialist sub-contractor responsible for the general earthworks, Geotextile Encased Columns (GECs) and basal platform construction.
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It should be noted that this is the first time that GEC columns have been installed under standing water. The unique installation procedure and standard equipment adaptations which were made by the contractor, ATLASYOL, have been a fusion of their previous GEC experience, common sense and the ability of the site based engineer team to design and manufacture unique apparatus and equipment all of which facilitated the successful construction. Of particular interest in this project was that the GEC heads terminated just 50cm above the bed of the lake, up to 7m below the water surface. This meant that the subsequent laying of the basal reinforcement for the basal platform, located immediately above the GEC heads, had to be supervised by divers in a very turbid environment with limited visibility.
a.p. van den berg The CPT factory
a.p. van den berg The CPT factory
Soil investigation equipment for onshore, offshore & near shore
Digital data acquisition system
Icone click-on modules
Onshore CPT equipment
Offshore CPT equipment
Easily connected to the Icone to measure other than the four standard parameters:
• Light & compact 100 kN CPT system • Versatile & easy to transport 200 kN CPT SKID for onshore and near shore • 200 kN CPT Trailer • Most compact & lightest crawler, the 100 kN Mini CPT Crawler • Compact & manoeuvrable 200 kN Midi CPT Crawler • 200 kN CPT Crawlers for rough or soft terrain; various designs • CPT Truck with 200 kN pushing force or even more • Versatile high-performance 200 kN CPT Track-Truck
ROSON seabed systems: • Near shore ROSON for water depths up to 500 m • ROSON 50, 2x50 & 100 for water depths up to 1,500 m • DW ROSON 100 & 2x100 for water depths up to 4,000 m WISON-APB wireline CPT systems: • WISON-APB-Classic for water depths up to 550 m • WISON-APB-1000 for water depths up to 1,000 m • WISON-APB-3000 for water depths up to 3,000 m Seabed Sampler XL for large high quality seabed samples at all water depths up to 4,000 m
• a digital data logger ‘Icontrol’ • a digital cone ‘Icone’ that measures the four standard parameters: cone tip resistance (qc ), sleeve friction (fs ), pore water pressure (u) and inclination (Ix/y ) • a pressure compensated Icone for water depths from 1,000 up to 4,000 m
• Seismic for determining the ground stability, by measuring the propagation speed of sound • Conductivity for measuring variations in the electrical conductivity of the soil • Magneto for detecting objects containing magnetisable metal • Vane for determining undrained & remoulded shear strength for the stability analysis of soft soils
APB CPT Ad ECSMGE 216x138 30062015 try2.indd 1
Geotechnical foundation design for onshore wind turbines Introduction In the Netherlands, as well as internationally, many onshore large scale wind farms are being constructed or developed (Figure 1). For these wind farms the machine size and rated power continuously (Figure 2). The consequence of this development is the increase of vertical and shear loads at the tower base, along with significant overturning moments. Large foundations are needed to resist these dynamic loads. The design of the large foundations is complex compared to building foundations. Besides the significant loads, the design includes the selection out of numerous codes and standards, foundation methods, calculation methods and design models, ranging from basic to advanced. The design should also meet some very strict design criteria. For this reason the understanding of the geotechnical investigations and design are becoming more and more important to achieve a safe and economical foundation design. To provide a better understanding, some considerations regarding geotechnical design and soil investigation are described in this article. Foundation types Wind turbines can be founded on a variety of foundation types. Figure 3 shows various foundation methods for onshore wind turbines. The selection of the best applicable foundation type depends mainly on the geotechnical conditions. The subgrade strength and stiffness of the soil or rock need to be sufficient to resist the cyclic and dynamic wind loads. A spread footing, or gravity foundation, can be considered as the simplest foundation type. The spread footing is placed directly on the foundation soil or rock. The most efficient form is a circular footing with a tapered cone. For construction of the cone a maximum angle of approximately 12° is often used. The weight of the concrete and optional overburden provides resistance against the overturning moments. At locations were strong bedrock is encountered near the surface, post-tensioned rock anchors
J.E. (Jurgen) Cools M.Sc. Geotechnical specialist at Royal HaskoningDHV, the Netherlands
Figure 1 - Construction of a large scale wind farm with 40 wind turbines in South Africa .
can be applied to reduce the dimensions of the cap foundation significantly. The rock anchors must be designed for fatigue.
Battered piles are often required to increase the lateral foundation stiffness and to increase the pile bearing capacity.
In regions where competent soil or rock is found at shallow depth, the overlying weak or compressible soil can be improved. Many techniques for improvement are available and depend on the type and thickness of the soil to be improved. If the thickness is small the soil can be excavated and recompacted or replaced. For greater thicknesses the soil can be improved without (e.g. vibro-compaction) or with admixtures (e.g. soil mixing) and rigid inclusions (e.g. short piles). In all cases the strength and stiffness of the soil mass is improved. The foundation slab is not connected to the inclusions and should be designed as a spread foundation.
A few decades ago wind turbines with hub heights of 20 to 30 meters were often founded on a small diameter monopile (open ended steel pile with approx. 4m diameter). Due to transportation limitations of the steel piles and the increasing overturning forces, the monopole has long been considered as inapplicable. However, currently new techniques are available to install very large diameter (segmental) steel monopiles (see [REF.2]) or to construct continuous bored pile circular walls that can be considered as an alternative foundation method to the more common piled foundations. An advantage of these more ‘innovative’ foundation methods is the limiting foundation footprint.
Pile and cap foundations are used in regions where the competent soil or rock is encountered at much greater depth. This foundation method is most commonly used in the Netherlands. Figure 4 shows an example of the construction of a pile foundation. The overturning forces on the cap foundation are being transferred to the piles as compressive and tensile axial loads. The piles transmit these loads to the ground via a combination of friction and end bearing. Lateral loads are resisted through lateral earth pressures on the piles.
GEOTECHNIEK - September 2015
Design guidelines, codes and standards The most commonly used design codes and guidelines for wind turbine foundation design are: • IEC-61400-1 Wind turbines - Design requirements [REF.3] • DNV/Risø Guidelines for Design of Wind Turbines [REF.4] • GL Guideline for the Certification of Wind Turbines [REF.5] • Eurocode 7: Geotechnical design of structures - part 1: general rules [REF.6]
The onshore wind industry continues to grow rapidly with the construction and development of many large scale wind farms using large megawatt wind turbines. This development requires a clear understanding of the geotechnical foundation behaviour in order to achieve a safe and economical foundation design. The foundation design however includes the selection of various types of foundations, design methodologies and mathema-
tical models, ranging from basic to advanced. The geotechnical designer needs to make a choice based on numerous codes and standards, design requirements and the anticipated geotechnical soil conditions. In some cases, standards are conflicting, or just lacking specific design guidelines for wind turbines. Based on extensive design experience some design considerations are described in this article.
Figure 2 - Growth in size of typical commercial wind turbines [REF.1].
Figure 3 - Foundation methods for onshore wind turbines. The grey subgrade represents competent soil or rock, the brown subgrade represents weak soil.
Besides these guidelines local codes and annexes (e.g. the Dutch NEN-EN) are also applicable. These national (building) codes are more general in nature. Requirements for the foundation design are often specified by turbine manufacturers in technical documents. Because of the differences between standards, guidelines and specifications, it is important that the designer is aware of any conflicts or omissions. The most important standards are briefly described below.
NEN-EN-IEC The wind turbine design in Europe shall meet the requirements contained in the Safety Standard ‘IEC 61400-1, Ed. 3’. In the Netherlands, these standards have been incorporated into the NENEN-IEC 61400-series. In this standard all Design Load Cases (DLC) for wind turbine design are specified. The IEC standard deviates from the Eurocode NEN-EN1990 (Basis of structural design). Two major differences concern the load factors and the design life time: • In the NEN-EN-IEC61400-1 load factors are
GEOTECHNIEK - September 2015
given for wind classes IEC 1, 2 or 3 to derive design values for the wind loads. According to this standard a value of 1.35 for normal loads and 1.10 for abnormal loads. The Eurocode NEN-EN-1990 distinguishes permanent and transient destabilizing loads, with a value of 1.5 for wind loads. • According to the NEN-EN-1990 for buildings a design life time of 50 years shall be used. The NEN-EN-IEC61400-1 specifies a design life time of 20 years for wind turbines. These differences indicate that wind turbines should be regarded as structures, other than buildings, for which in the Netherlands specific NEN-EN-IEC standards should be applied. Based on these standards a level of structural safety can be achieved, as required by the Building Act. The IEC standard is lacking specific guidelines for geotechnical investigation and foundation design, therefore additional geotechnical standards and guidelines should be applied. Eurocode 7 (NEN-EN9997-1 including national Annex) The Eurocode 7 does not cover the specific foundation design of wind turbines. More general, this standard gives recommendations for the scope of the geotechnical site investigation. It also prescribes a limit state design method. The limit state design implies the application of partial factors load and material parameters or resistance. With regard to the partial factors for foundation design no distinction is made between the different consequence classes. For structures this distinction is commonly made according to Eurocode NEN-EN 1990. However, since the load factors for wind turbines are determined according to the IEC, a distinction in consequence classes is not relevant. DNV Guidelines The Det Norske Veritas (DNV) and Risø National Laboratory have jointly drafted guidelines for the design of wind turbines. With regard to the DLC’s the DNV guidelines are consistent with the IEC standard. One section is related to the foundation design. In this section recom-
mendations are given for ground investigation. The guidelines also provide design methods for a spread foundation and a foundation on piles. The proposed design methods for spread foundations are comparable with the Eurocode 7. Regarding the design of pile supported foundations the calculation methods comply with the API standard, and not with the Eurocode 7. GL Guidelines Germanischer Lloyd (GL) together with DNV has drafted guidelines and technical specification for the certification of wind turbines. In the ‘Guideline for the Certification of Wind Turbines’ [REF.5] requirements are given for the scope of the geotechnical site investigation. It is stated that the investigation program should comply with at least Geotechnical Category 2 according to the German DIN (DIN 4020:2010-12 and DINEN 1997-2). This geotechnical category is consistent with Geotechnical Category 2 according to part 2 of Eurocode 7 [REF.7]. The guideline also describes the methodology to be used for the foundation design. The prescribed method is consistent with part 1 of the Eurocode 7.
Limit states In accordance with the Eurocode 7 and the GL guidelines the design shall be checked for the ultimate limit state (ULS) and the service limit states (SLS). The limit states that must be checked are listed in Table 1. Geotechnical design criteria Some of the most important design criteria that are specified in codes, guidelines and technical documents are briefly described below. Rotational foundation stiffness For both spread foundations, as well as piled foundations, one of the main design criteria for foundation design is the rotational stiffness. In order to avoid excessive motion at the tower top and to provide the required damping, the turbine manufacturer always provides a minimum rotational stiffness value. The final foundation design must satisfy this minimum value. Typical minimum values of the rotational stiffness are 60 to 120 GN-m/rad. The rotational stiffness depends on the stiffness
Figure 4 - Example of the construction of a pile foundation for Wind Farm De Zuidlob in The Netherlands
of the foundation structure and the subsoil. In case of a spread foundation often an infinitely rigid foundation block is assumed, so that the rotational stiffness only depends on the dynamic shear modulus G of the subsoil, and hence of the dynamic modulus of elasticity Edyn and the Poisson’s ratio v. The DNV guidelines provide a clear overview of formulas for the rotational stiffness for various subsoil conditions and foundation methods. For pile foundations, the rotational stiffness depends very much on the cyclical spring stiffness of the piles. The cyclical spring stiffness of the piles can be determined according to the empirical method based on pile load tests (as described in [REF.8] and [REF.9]). A typical value for the minimum required cyclical spring stiffness is about 200 MN/m. Gapping The GL guidelines describe ground gap limitations for the foundation design. The ground gap criterion requires that under specific IEC normal operational design load cases, no ground gap (i.e. zero contact pressure) shall occur at the foundation-soil contact. This means that in these cases the entire foundation footprint must remain in compression. This gapping limit acknowledges that in the case of a spread foundation the rotational stiffness decreases non-linearly after foundation uplift (zero contact pressure). Besides this, in certain soil conditions, a limit on gapping will also ensure that soils subject to cyclic degradation are prevented from experiencing multiple instances of zero pressure which, in the presence of water, could lead to breakdown of the in-situ soil structure and subsequent related serviceability problems [REF.8]. For unfactored loads (SLS) under specific operational conditions it has to be proven that the eccentricity (e) of the total vertical load is less than 25% of the radius: e < R/4. Commonly, for unfactored (SLS) extreme loads, no more than 50% of the base area may be
Table 1 - Limit states. Spread foundation
Foundation on piles
Overturning (on rock) Overall stability (slopes) Rotational shear failure Bearing resistance Sliding resistance Buoyancy
Ground gap Settlement Heave (swelling, frost) Tilt Foundation stiffness
Overall stability Compressive resistance Uplift or tensile resistance Lateral resistance
Settlement Heave Tilt Rotational stiffness Lateral stiffness
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GEOTECHNICAL FOUNDATION DESIGN FOR ONSHORE WIND TURBINES
Figure 5 - Relation between overturning moment and the rated power.
without compression. It has to be proven that: e < 0.59*R. For spread foundations on competent rock with a good rock mass quality gapping can be allowed. The high stiffness and strength allow for some decrease due to the prolonged load alterations. Horizontal stiffness For a pile foundation a minimum horizontal stiffness is required by the turbine manufacturer. Often the minimum value depends on the total mass of the wind turbine and its foundation, or on the rotational stiffness. Typical minimum values of the horizontal stiffness vary between 500 and 1000 MN/m. A common procedure to calculate the lateral load resistance is by means of the horizontal subgrade reaction for the different soil layers below foundation level. The subgrade reaction can be determined according to Menard’s method, based on the actual cone resistance and soil pressure. For this analysis two dimensional calculation models (e.g. DSheetPiling – single pile) can be used, in which the stiffness of a single vertical pile is calculated. More advanced programs with three dimensional models (e.g. Plaxis3D or Ensoft Group) can be used to take into account the effect of battered piles. It should be taken into consideration that the lateral soil reaction should be reduced due to the cyclic loading. This reduction can be up to 10% for sand and 30% for clay (according to DNV guidelines). Overall stability For spread foundations it must be verified that the overall stability is sufficient. This is parti-
Figure 6 - Relation between the overturning moment and the hub height.
cularly relevant for footings placed on or near sloping ground. For spread foundations on rock the overturning stability should be checked for the extreme loads in the ULS. According the GL guidelines and Eurocode 7 the safety against is guaranteed by verification of the bearing capacity of the subsoil. Bearing capacity Generally spread footings are first proportioned for bearing capacity (and overturning stability in case of rock). The bearing capacity of the ground can be calculated from the general bearing capacity equations as provided in the Eurocode 7. The DNV guidelines provide additional equations for extreme eccentric load cases. In these equations equations for bearing capacity and load inclination factors it is important to apply negative values for the shear forces due to change in direction of the rupture.
Foundation loads Wind turbine Generally wind turbine foundations are subjected to vertical and shear forces along with significant overturning moments. These overturning moments are principal for foundation design. Figure 5 depicts the relation between the hub height and overturning moment at the tower base, based on technical data from various wind turbines. From the figure is can be seen that there is wide range of moments for a certain hub height due a large variation in weight of the nacelle and design wind load. Figure 6 shows the relation between the overturning moment and the rated power.
In calculation of the settlements it is important to differentiate static loads, cyclic wind loads and dynamic loads.
Design loads of the tower are provided by specialists of the wind turbine manufacturer in a Load Document (or technical specifications) that satisfies the load cases as outlined in the IEC 61400 standards. All design load combinations as indicated by the IEC 61400 standards are analysed. A distinction is made between operational loads, extreme normal loads, extreme abnormal loads and fatigue. The operational loads are cyclic loads such as low wind speeds of 5m/s or starting and stopping of the wind turbine generator. The extreme loads have a low probability of occurrence, but result in high design forces at the tower base. Examples of extreme winds are 1-in-50 year 3 second gust or turbine emergency stops. The extreme loads can be considered as dynamic loads. It depends on the turbine type and site specific wind conditions which DLC is governing for foundation design.
Settlement is not governing for design in case of spread foundations on dense soil or good rock, since (average) contact pressures from vertical loads are typically quite low (e.g. 50 to 75 kPa).
Seismic loading The effect of earthquake loads on the wind turbine is analysed by the turbine supplier in a number of prescribed load cases. In the Technical
Tilt The foundation shall be designed to minimize settlements and especially differential settlements. Deformation criteria are often specified by the turbine manufacturer. Generally for wind turbines the following must be kept strictly: • A maximum inclination of 3mm/m resulting from the characteristic extreme load; • A maximum unequal settlement of the foundation of 1mm/m from the characteristic operational load.
GEOTECHNIEK - September 2015
Figure 7 - Seismic survey (electrical resistivity) along 3 wind turbine locations down to 10 m depth for wind farm Cerfontaine in Belgium. The blue and green areas indicate the soft soil cover, which is overlying a karstic limestone (indicated by orange and red color).
Documents the assumed seismic ground accelerations are specified. As a foundation designer one needs to check whether the site-specific ground accelerations do not exceed the assumed values. In case of exceeding new calculations should be performed by the manufacturer. The seismic load is normally combined with operational loads and not with extreme loads. The resulting foundation loads are in many cases lower than under extreme loads without earthquake. However, it should be checked in the foundation design that, for example due to liquefaction, the bearing capacity is not exceeded. Buoyancy The foundation design shall take into account the vertical hydrostatic pressure (buoyancy). Important is how to deal with buoyancy. The upward water pressure can be regarded as a negative stabilizing load or as a positive destabilizing load. The designer should be aware of the effect of buoyancy for various limit states. Ground investigation Based on the level of complexity of design the Geotechnical Category 2 (GC2) is selected for spread foundations according to Eurocode 7. For pile foundations GC3 is selected due to the cyclic and dynamic pile loading. Eurocode 7 and the DNV guidelines provide recommendations for the scope of the ground investigation. Turbine manufacturers often provide additional specifications for ground investigation. Generally the soil investigations consist of the following parts: • Geological desk study • Geotechnical investigations • Geophysical survey (optional) A geological desk study should be performed first to establish a basis for selection of methods and extent of the site investigation. A geotechnical investigation may consist of trial pits, borings, in-situ testing, soil sampling and laboratory testing. For pile foundations in
the Netherlands at least three CPTs should be performed for each wind turbine location, combined with one borehole in the centre point of the structure. For spread foundations on soil or rock it is sufficient to perform at least two borings or CPT’s for each location, of which one is performed in the centre point of the wind turbine. The depth to be covered by the investigation is at least 5m in case of unweathered rock. For foundations on soil the depth to be covered should be at least equal to 1 to 1.5 times the largest base dimension of the footing. A typical investigation depth is 20 to 30 m. Geophysical survey can be used to extend the localized information from borings or CPTs. The results give a better understanding of the stratigraphy within the considered area. Figure 7 shows the result of a geophysical survey for wind turbines on karstic bedrock. The spatial variation of the top level of the rock could not have been investigated with only borings and CPTs. Geophysical survey (Ground Penetrating Radar) has also proven its success for wind farms in Finland to investigate the level of the bedrock and thickness of the moraine cover. Conclusions Due to the continuous developments in the wind industry it is expected that loads imposed on wind turbine foundations will also become larger. This requires a clear understanding of the driving considerations for geotechnical design. Due consideration should be given to the following matters: • The extent of the geotechnical investigation should be based on a geological desk study and comply with the Eurocode 7 and the GL guidelines. For difficult soil or rock conditions geophysical survey can is preferred to extend the localized information from borings and CPTs. • Based on the encountered ground conditions the most appropriate foundation method
GEOTECHNIEK - September 2015
should be selected, that meets the stringent design criteria. Principal design criteria for design of spread foundations are ground gapping and the maximum permissible settlement and tilt. For pile foundations the principal design criterion often is the rotational stiffness. • Wind turbines should be regarded as structures, other than buildings, for which in specific IEC 61400 standards exist. These standards are however lacking specific design guidelines for wind turbine foundations. The DNV guidelines explicitly address the design of wind turbine foundations. For spread foundations these guidelines are well applicable, especially for calculation of bearing capacity under extreme eccentricity. For pile foundations the design methods provided in the Eurocode 7 are preferred. The methods in the DNV for pile foundations comply with the API standards. References  Wiser, R., et al (2011), Wind Energy. In IPCC Special Report on Renewable Energy Sources and Climate Change Mitigation. Cambridge University Press, Cambridge, United Kingdom and New York, NY, USA.  Middendrop, P. and Dorp, R. van (2012), Van Mini- tot Giga-palen. Geotechniek, Funderingsdag 2012 Special. J.16, no.5: pages 30-33.  International Electrotechnical Commission (2005), IEC-61400-1 Wind turbines – Part 1: Design requirements. Third edition.  DNV/Risø (2002), Guidelines for Design of Wind Turbines.  GL Renewables Certification (2010), Guideline for the Certification of Wind Turbines. Hamburg, Germany.  Eurocode 7: Geotechnical design of structures - part 1: general rules  Eurocode 7: Geotechnical design of structures – part 2: Ground investigation and testing  Morgan K. and Ntambakwa, E. (2008), Wind Turbine Foundation Behaviour and Design Considerations. AWEA Windpower Conference, Houston.  ‘Fundamenteel’ in Cement 1995/9, pages 17-19  ‘Fundamenteel’ in Cement 1995/5, pages 74-76
Website: http://www.windfarmbop.com/ wind-turbines-foundation-design/
Reinforced soil walls over compressible soils
Sander Suk Terre Armee B.V.
In the early 1960’s Henri Vidal, introduced the Terre Armée (Reinforced Earth®) construction technique. Henri Vidal conceptualized this method and built the first full scale demonstration. For 50 years since, Terre Armée has set the standards for reinforced soil structures and has been involved in more than 50,000 projects all over the world, accumulating knowledge and experience in the field of engineered backfills.
Figure 1 Terre Armée structures can be built on compressible subsoil. The settlements have to be accounted for in the design. Conventional Reinforced Earth walls with precast concrete facing panels can be used in areas where the anticipated differential settlements are within the tolerable limit for the precast panel system. The limit depends of the width of the joints, but typically less than 1%, or 1 metre of differential settlement along a 100 metre wall length. At locations where greater differential settlements are expected, for example a two stage construction or an obstacle in the subsoil, special provisions like a vertical joint can be applied. In many projects limitations to the acceptable differential settlements are set by project requirements, like aesthetics or a superstructure. Most often, common geotechnical solutions can be applied: preloading, soil improvement, light backfilling materials or consolidation methods. Preloading can be done before, during or after construction of the Terre Armée walls. Over compressible or poor quality soils where pre-
loading or soil improvement cannot be used, piles combined with basal reinforcement or rigid inclusions like Menard’s Controlled Module Columns (CMC) can be used to control both stability and settlement of the structure. Menard’s Controlled Module Columns (CMC) is a technique where grout is installed in the foundation soil to both densify the soil and to provide additional support. On numerous projects over the world Terre Armée structures have been built on piles with basal reinforcement or rigid inclusions. Based on theoretical models additional soil reinforcing strips were added to account for potential additional loading due to stress concentrations caused by arching around the top of the inclusions. In 2012, in order to understand more clearly and demonstrate the behaviour of the combined system during and after construction, extended theoretical studies and a full scale instrumentation was performed on a structure, constructed along the Garden State Parkway in Bass River (USA). Data was recorded over 5 months during construction and 2 months after construction. Figure 3 shows the instrumentation. Stress gauges to measure the stress in the reinforcing strips; gauges to measure soil pressures and settlements at the bottom of the reinforced soil block and gauges to measure stresses in the rigid inclusions. Based on the analysis of the recorded instrumentation data, it was shown that the strains within the reinforcing strips did not exceed in the very lowest
GEOTECHNIEK - September 2015
levels of the MSE wall mass the conventional design values. As a result, there is no need for increasing the density of soil reinforcing strips in the lower levels of reinforcement. The location of the top of the rigid inclusions is an important aspect of the design to prevent the soil reinforcement within the Reinforced Earth being overloaded. If the distance is too small, the stresses in the reinforcing strips may increase due to differential movement around and in between the rigid inclusions. A vertical distance of 0.55 m has proven sufficient. In some cases, additional reinforcing geosynthetics are used in the LTP where the loading is further spread out and the concentration of stresses is reduced. Geosynthetics have been found to be ineffective in the LTP for steel reinforced RE walls, since steel is relatively inextensible and will engage loading before geosynthetics can be effective. Studies and build structures have proven that the combination of two techniques: Terre Armée and rigid inclusions, is an economical and efficient solution that has the potential to accelerate construction and limit long term settlement. This solution gives designers, owners and contractors a valuable tool to design Reinforced earth over compressible soils, that can be used next to more conventional methods like soil improvement, preloading, LPTs or using light backfilling materials.
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